Nuta Elena 201003 MASc Thesis

7/30/2019 Nuta Elena 201003 MASc Thesis

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SEISMICANALYSIS OF STEELWINDTURBINETOWERS

IN THE CANADIAN ENVIRONMENT

by

Elena Nuta

A thesis submitted in conformity with the requirements

for the degree of Master of Applied Science

Graduate Department of Civil Engineering

University of Toronto

Copyright by Elena Nuta (2010)


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SEISMICANALYSIS OFWINDTURBINETOWERS IN THE CANADIAN ENVIRONMENT

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Seismic Analysis of Steel Wind Turbine Towers in the Canadian Environment

Master of Applied Science (2010)

Elena Nuta

Department of Civil Engineering

University of Toronto

ABSTRACT

The seismic response of steel monopole wind turbine towers is investigated and their risk is

assessed in the Canadian seismic environment. This topic is of concern as wind turbines are

increasingly being installed in seismic areas and design codes do not clearly address this aspect of

design. An implicit finite element model of a 1.65MW tower was developed and validated.

Incremental dynamic analysis was carried out to evaluate its behaviour under seismic excitation, to

define several damage states, and to develop a framework for determining its probability of damage.

This framework was implemented in two Canadian locations, where the risk was found to be low for

the seismic hazard level prescribed for buildings. However, the design of wind turbine towers is

subject to change, as is the design spectrum. Thus, a methodology is outlined to thoroughly

investigate the probability of reaching predetermined damage states under seismic loading for future

considerations.


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ACKNOWLEDGEMENTS

I would firstly like to express my gratitude to Professor J. A. Packer and to Professor C.

Christopoulos for their guidance and the countless meetings that ensured my project was always on

track. I see now, in retrospect, how paramount this guidance was, and I thank you both.

Special thanks go out to Andrew Voth and Dr. Gilberto Martinez-Saucedo, for many hours of

help working out finite element modelling glitches, and to Lydell Wiebe and Nabil Mansour, for

their willingness to always discuss thesis concerns with me. I would also like to thank my many

officemates, research group members, and colleagues for enriching my graduate experience and

providing conversation and laughter.

Financial support has been provided by Ontario Graduate Scholarships (OGS), the National

Sciences and Engineering Research Council of Canada (NSERC), and the Steel Structures Education

Foundation (SSEF). I also gratefully acknowledge the Ontario Centres of Excellence (OCE), and

the Fraunhofer Centre Windenergie und Meerestechnik, Bremerhaven, Germany where I spent the

summer of 2008 as an intern.

Last but not least, I thank the people most important in my life. To my awesome parents,

Floarea and Mihai Nuta, thank you for teaching me to always aim high and for supporting me

always. To my beautiful sister and brother-in-law, Gabriela Nuta and Andrew Orel-Golla, thank you

for making sure I had enough distractions to stay sane and for accommodating my erratic schedule.To my loving boyfriend, Michael Colalillo, thank you for your motivation, understanding,

encouragement, time, and patience during this time; and of course, thank you for the pasta dinners.

To all my friends, thank you for the unwavering mental support and for never doubting me.


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TABLE OF CONTENTS

ABSTRACT .............................................................................................................................. ii

ACKNOWLEDGEMENTS .................................................................................................... iii

TABLE OF CONTENTS........................................................................................................ivLIST OF TABLES................................................................................................................... ix

LIST OF FIGURES................................................................................................................. xi

LIST OF SYMBOLS AND ABBREVIATIONS.....................................................................xv

CHAPTER 1:INTRODUCTION...................................................................................................1

1.1 Overview of Thesis................................................................................................................................ 1

1.2 Wind Turbine Type, Components, and Terminology .........................................................................2

CHAPTER 2:LITERATURE REVIEW..........................................................................................4

2.1 International Standards ........................................................................................................................4

2.1.1 International Electrotechnical Commission (IEC).................................................................4

2.1.2 Germanischer Lloyd (GL) ........................................................................................................5

2.1.3 Det Norske Veritas (DNV) ......................................................................................................5

2.1.4 Other European Standards.......................................................................................................7

2.2 Canadian Standards ..............................................................................................................................7

2.2.1 CAN/CSA-C61400-1:08, Wind Turbines Part 1: Design Requirements.............................. 7

2.2.2 CAN/CSA S37-01, Antennas, Towers, and Antenna-Supporting Structures..........................8

2.2.3 CAN/CSA S473-04, Steel (Fixed Offshore) Structures............................................................8

2.2.4 CAN/CSA S16-09, Design of Steel Structures .........................................................................8

2.3 Book Publications.................................................................................................................................8

2.4 Current Research on Wind Turbine Towers........................................................................................ 9

2.4.1 Comparison of Seismic Analysis Methods: Frequency-Domain vs. Time-Domain..............9

2.4.2 Shell Buckling......................................................................................................................... 10

2.4.3 Dynamic Soil-Structure Interaction Effects............................................................................11

2.5 Summary ............................................................................................................................................. 12

CHAPTER 3:FINITE ELEMENT MODEL DEVELOPMENT ANDVALIDATION ......................... 13

3.1 Geometry of Wind Turbine Towers.................................................................................................... 13

3.2 Finite Element Analysis Program ...................................................................................................... 13

3.3 Material Properties.............................................................................................................................. 14

3.4 Choice of Elements............................................................................................................. ................ 16


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3.4.1 Shell Elements ........................................................................................................................ 16

3.4.1.1 Classical Plate Theory............................................................................................ 16

3.4.2 Solid Elements........................................................................................................................ 17

3.4.2.1 Elastic Beam Theory ............................................................................................. 18

3.4.3 Solid-Shell Interaction............................................................................................................ 18

3.5 Connection Modelling ........................................................................................................................ 183.6 Tubular Members under Bending...................................................................................................... 19

3.6.1 FE Model for Pure Flexure .................................................................................................... 21

3.6.1.1 Mesh Sensitivity ..................................................................................................... 22

3.6.1.2 Refinement of Mesh...............................................................................................23

3.6.1.3 Results and Analysis of Tubular Members under Pure Flexure........................... 24

3.6.2 FE Model of a Cantilever Tower under Bending .................................................................. 27

3.6.2.1 Results and Analysis of Cantilever Tower under Bending ................................... 28

3.6.2.2 Stiffening Effect of a Flange .................................................................................. 29

3.6.2.3 Effect of Local Imperfections on Flexural Behaviour...........................................303.7 Tubular Members under Axial Compression..................................................................................... 31

3.7.1 FE Analyses for Validation of Axial Buckling Behaviour ..................................................... 32

3.7.1.1 FE Models for Validation of Axial Buckling Behaviour....................................... 32

3.7.1.2 Eigenvalue Buckling Analysis...............................................................................32

3.7.1.3 Nonlinear Buckling Analysis Geometric and Material Nonlinearities ............. 33

3.7.1.4 Geometric Imperfections for Global Buckling...................................................... 33

3.7.1.5 Geometric Imperfections for Local Buckling ....................................................... 33

3.7.1.6 Results and Analysis of Cantilever Tower under Axial Compression .................. 34

3.7.2 Summary of Modelling Decisions..........................................................................................36

3.8 Time-History Analysis under Seismic Excitation.............................................................................. 36

3.8.1 Damping in ANSYS ............................................................................................................... 37

3.8.1.1 Comparison of Effect of Damping ........................................................................ 38

3.8.1.2 Aerodynamic Damping.......................................................................................... 41

3.8.2 Incremental Nonlinear Analysis ............................................................................................ 41

3.8.2.1 Failure Mode.......................................................................................................... 42

3.9 Summary ............................................................................................................................................. 43

CHAPTER 4:PRELIMINARYANALYSIS OFVESTASWINDTURBINETOWER...........................45

4.1 Structure Characteristics.....................................................................................................................45

4.1.1 Dimensions And Details ........................................................................................................ 45

4.1.1.1 Discontinuities....................................................................................................... 45

4.1.2 Mass........................................................................................................................................48

4.1.3 Mode Shapes .......................................................................................................................... 49

4.1.4 Damping.................................................................................................................................49


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4.2 Finite Element Model of Vestas Wind Turbine Tower ..................................................................... 49

4.3 Pushover Analysis ............................................................................................................................... 52

4.3.1 Background ............................................................................................................................ 52

4.3.2 Pushover Analysis of Wind Turbine Tower...........................................................................53

4.3.2.1 Imposed Imperfections..........................................................................................54

4.3.3 Results of Pushover Analysis ................................................................................................. 554.3.3.1 Interpretation of Pushover Analysis Results......................................................... 57

4.4 Summary ............................................................................................................................................. 57

CHAPTER 5:NONLINEARTIME-HISTORYANALYSIS OF THEVESTASWINDTURBINETOWER........................................................................................................................................58

5.1 Earthquake Suite.................................................................................................................................58

5.1.1 Earthquake Input in Time-History Analyses........................................................................62

5.1.2 Scaling of Earthquake Records..............................................................................................62

5.2 Results of LA01 & LA02 (Imperial Valley, 1940, Elcentro)................................................................ 635.2.1 Intensity and Damage Measures ........................................................................................... 63

5.2.1.1 Peak Displacement ................................................................................................ 64

5.2.1.2 Peak Rotation.........................................................................................................65

5.2.1.3 Peak Stress .............................................................................................................66

5.2.1.4 Residual Deformation............................................................................................ 67

5.2.2 Displaced Shape ..................................................................................................................... 67

5.2.3 Time-History Displacement Response ................................................................................. 69

5.2.4 Orbit Plots .............................................................................................................................. 70

5.2.5 Definition of Damage States for Wind Turbine Towers ....................................................... 72

5.2.5.1 0.2% Residual Out-of-Straightness........................................................................72

5.2.5.2 First Yield...............................................................................................................72

5.2.5.3 1.0% Residual Out-of-Straightness........................................................................72

5.2.5.4 First Buckle / Loss of Tower.................................................................................72

5.3 Summary of Results for LA Earthquake Suite ................................................................................... 73

5.3.1 Incremental Dynamic Analysis Curves..................................................................................74

5.3.1.1 Assessment of Damage Measures ......................................................................... 74

5.3.1.2 Average Damage Measures ................................................................................... 75

5.3.2 Location of Buckle for 4th Damage State...............................................................................75

5.3.3 Definition of Fragility Curves ................................................................................................ 76

5.3.4 Effect of Vertical Earthquake Component ............................................................................ 79

5.3.5 Effect of Damping..................................................................................................................80

5.3.6 Validation of Connection Modelling...................................................................................... 81

5.4 Summary ............................................................................................................................................. 83


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CHAPTER 6:INCREMENTALANALYSIS FORTWO CANADIAN SITES ......................................85

6.1 Eastern Canada Site............................................................................................................................85

6.1.1 Simulated Time-History Records for the Eastern Canada Site ............................................ 86

6.1.2 Earthquake Suite for the Eastern Canada Site ...................................................................... 86

6.2 Western Canada Site ........................................................................................................................... 89

6.2.1 Simulated Time-History Records for the Western Canada Site............................................896.2.2 Earthquake Suite for the Western Canada Site ..................................................................... 90

6.3 Methodology for Scaling Records for IDA.........................................................................................93

6.3.1 Efficiency of Method..............................................................................................................94

6.4 Results for Eastern Canada Site ......................................................................................................... 96

6.5 Results of Time-History Analysis for Western Canada Site .............................................................. 97

6.5.1 Incremental Dynamic Analysis Curves..................................................................................97

6.5.1.1 Average Damage Measures ................................................................................... 98

6.5.2 Fragility Curves ...................................................................................................................... 99

6.6 Summary ........................................................................................................................................... 100

CHAPTER 7:CONCLUSIONS AND RECOMMENDATIONS........................................................101

REFERENCES..................................................................................................................... 103

APPENDIX A: RESULTS OF INCREMENTALTIME-HISTORYANALYSIS FORLA

EARTHQUAKE SUITE.................................................................................. 107

A.1 LA03 & LA04 (Imperial Valley, 1979, Array #05)............................................................................ 108

A.2 LA05 & LA06 (Imperial Valley, 1979, Array #06)............................................................................. 111

A.3 LA07 & LA08 (Landers, 1992, Barstow)...........................................................................................114

A.4 LA09 & LA10 (Landers, 1992, Yermo)..............................................................................................117

A.5 LA11 & LA12 (Loma Prieta, 1989, Gilroy) ....................................................................................... 120

A.6 LA13 & LA14 (Northridge, 1994, Newhill)...................................................................................... 123

A.7 LA15 & LA16 (Northridge, 1994, Rinaldi RS) ................................................................................. 126

A.8 LA17 & LA18 (Northridge, 1994, Sylmar)........................................................................................ 129

A.9 LA19 & LA20 (North Palm Springs, 1986) ...................................................................................... 132

A.10 IDA Curves for Investigated Intensity Measures ........................................................................... 135

APPENDIX B: SEISMIC HAZARD FORTWO CANADIAN SITES............................................ 138

APPENDIX C: RESULTS OF THEWESTERN CANADAEARTHQUAKE SUITE........................141

C.1 WCan01 (Magnitude 6, 8 13 km) .................................................................................................. 142

C.2 WCan02 (Magnitude 7, 10 26 km) ................................................................................................ 144



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C.6 WCan06 (Magnitude 7, 30 100 km) .............................................................................................. 152

C.7 WCan07 (Cascadia Record, Magnitude 9, 112 201 km)................................................................ 154

C.8 Fragility Curves for Additional Intensity Measures........................................................................ 156


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LIST OFTABLES

Table 3.1: Cross-sectional slenderness limits (D/t) of circular hollow sections in bending 20

Table 3.2: Slenderness limits (D/t) of circular hollow sections in axial compression for non-slenderbehaviour, using E = 200000 MPa 31

Table 3.3: Summary of results of time-history analyses for the UCSD tower comparing different dampingvalues 38

Table 3.4: Summary of results of incremental time-history analyses for the UCSD tower 42

Table 4.1: Mass of Vestas wind turbine tower 48

Table 5.1: Properties of LA earthquake suite records 59

Table 5.2: Summary of displacement results of time-history analyses subjected to LA01 & LA02 (ImperialValley, 1940, Elcentro) 63

Table 5.3: Minimum, average, and maximum values of damage measures at each damage state 75

Table 5.4: Location of buckle for the LA earthquake suite 76

Table 5.5: Intensity measures (magnification factors) of each earthquake analysis and statistics for all thedamage states for the LA earthquake suite 77

Table 5.6: Properties of earthquake records for analyses that included a vertical component 79

Table 5.7: Variation of peak displacement compared to 1% damping for LA11 and LA12 (Loma Prieta,1989, Gilroy) 80

Table 5.8: Properties of bolts used in intermediate flanges of Vestas wind turbine tower 82

Table 5.9: Characteristics of wind turbine tower flanges 83

Table 6.1: Spectral hazard values (Sa(T)) and peak ground acceleration (PGA) for the Eastern Canada site,2% probability of exceedance in 50 years 85

Table 6.2: Scale factors and PGA of earthquake records chosen for the Eastern Canada site 87

Table 6.3: Spectral hazard values (Sa(T)) and peak ground acceleration (PGA) for the Western Canada site,

2% probability of exceedance in 50 years 89

Table 6.4: Scale factors and PGA of earthquake records chosen for the Western Canada site 91

Table 6.5: Summary of results of time-history analyses for the Eastern Canada suite 96

Table 6.6: Summary of results of time-history analyses for the Eastern Canada suite with magnificationfactor of 10 96

Table 6.7: Minimum, average, and maximum values of damage measures at each damage state for theWestern Canada site 99

Table 6.8: Probability of exceedance of particular damage states for varying seismic event intensities 100

Table A. 1: Summary of displacement results of time-history analyses subjected to LA03 & LA04 (ImperialValley, 1979, Array #05) 108

Table A.2: Summary of displacement results of time-history analyses subjected to LA05 & LA06 (ImperialValley, 1979, Array #06) 111

Table A.3: Summary of displacement results of time-history analyses subjected to LA07 & LA08 (Landers,1992, Barstow) 114

Table A.4: Summary of displacement results of time-history analyses subjected to LA09 & LA10 (Landers,1992, Yermo) 117


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Table A.5: Summary of displacement results of time-history analyses subjected to LA11 & LA12 (LomaPrieta, 1989, Gilroy) 120

Table A.6: Summary of displacement results of time-history analyses subjected to LA13 & LA14(Northridge, 1994, Newhill) 123

Table A.7: Summary of displacement results of time-history analyses subjected to LA15 & LA16(Northridge, 1994, Rinaldi RS) 126

Table A.8: Summary of displacement results of time-history analyses subjected to LA17 & LA18(Northridge, 1994, Sylmar) 129

Table A.9: Summary of displacement results of time-history analyses subjected to LA19 & LA20 (NorthPalm Springs, 1986) 132

Table C.1: Summary of results of time-history analyses for WCan01 142




Table C.5: Summary of results of time-history analyses for WCan05 150Table C.6: Summary of results of time-history analyses for WCan06 152



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LIST OF FIGURES

Figure 1.1: Typical horizontal-axis wind turbine 2

Figure 3.1: Engineering and true stress-strain curve from Voth (2010) for cold-formed circular HSS 15

Figure 3.2: Stress-strain curves used in subsequent analyses 15

Figure 3.3: Geometry of shell element used to represent tower walls (ANSYS, 2007) 16

Figure 3.4: Geometry of 20-noded solid element used to represent flanges (ANSYS, 2007) 17

Figure 3.5: Geometry of wind turbine tower ring flanges 18

Figure 3.6: Bolted flange connections of wind turbine tower (Vestas, 2006) 19

Figure 3.7: Schematic and descriptions of FE models for validation of pure flexure 21

Figure 3.8: Normalised moment-curvature response of FEA of VF-el for various mesh sizes using uniformmesh and perfect geometry/loading 22

Figure 3.9: Refined mesh configuration 23

Figure 3.10: Incorrect buckling configuration due to perfectly symmetrical model and loading 24

Figure 3.11: Normalised moment-curvature response of FE model VF-el compared with experimental resultsfrom Elchalakani et al. (2002) 25

Figure 3.12: Local buckling failure of CHS under pure bending, D/t = 111 25

Figure 3.13: Normalised moment-curvature response of FE models VF-1 and VF-el (D/t = 111) 26

Figure 3.14: Normalised moment-curvature response of FE models VF-2 (D/t = 286) 27

Figure 3.15: Local buckle failure of very slender CHS under pure bending (D/t = 286) 27

Figure 3.16: Schematic and descriptions of FE models for validation of bending behaviour 28

Figure 3.17: Normalised moment-curvature response of FE models VB-1 (D/t = 111), VB-2 (D/t = 276),and VBS-1 (stiffened) 29

Figure 3.18: Investigation of location of buckle 30

Figure 3.19: Effect of local imperfections on cantilever tower under bending 31Figure 3.20: Schematic and descriptions of FE models for validation of axial compression 32

Figure 3.21:Assessment of influence of geometric imperfections for local buckling 34

Figure 3.22:Axial loading analysis FE results of VA-1 (D/t = 111) with and without local imperfections 35

Figure 3.23:Axial loading analysis FE results of VA-2 (D/t = 286) with and without local imperfections 35

Figure 3.24: Details of small wind turbine tested at UCSD 37

Figure 3.25:Acceleration at top of nacelle for the reference earthquake for various damping ratios 39

Figure 3.26:Acceleration at upper joint for the reference earthquake for various damping ratios 40

Figure 3.27: Displacement response of incremental time-history analysis of small wind turbine 42

Figure 3.28: Buckled shape of UCSD wind turbine tower analysis at a magnification factor of 10 43Figure 4.1: Details at base of Vestas wind turbine tower 46

Figure 4.2: Wind turbine tower dimensions and layout (Vestas, 2006) 47

Figure 4.3: D/t ratio of Vestas wind turbine tower sections along the height with CSA (2009b) cross-sectionclassification in bending 48

Figure 4.4: Mode shapes of Vestas tower in horizontal direction 49

Figure 4.5: Mesh of Vestas wind turbine tower 51

Figure 4.6: Direction of pushover analyses 54


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LIST OF FIGURES


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Figure 4.7: Multimode load pattern for pushover analysis of Vestas wind turbine tower 54

Figure 4.8: Load-displacement curves for pushover analysis at 0 for material properties with gradualyielding and with yield plateau 55

Figure 4.9: Buckled failure of Vestas wind turbine tower subjected to pushover analysis at 0 56

Figure 4.10: Peak load for pushover analysis acting at various angles 56

Figure 4.11:Top displacement at peak load for pushover analysis acting at various angles 56

Figure 5.1: Elastic acceleration response spectra for earthquake suite considered 60

Figure 5.2: Elastic displacement response spectra for earthquake suite considered 60

Figure 5.3: Accelerograms of 20 scaled ground motion records of the LA earthquake suite 61

Figure 5.4: Top view of tower showing definition of angle in plan view 64

Figure 5.5: Displaced shape of wind turbine tower used to determine the peak rotation for LA01 & LA02(Imperial Valley, 1940, Elcentro) 66

Figure 5.6: Displaced shape of wind turbine tower at various magnification factors for LA01 & LA02(Imperial Valley, 1940, Elcentro) 68

Figure 5.7: Bucked shape of Vestas wind turbine tower analysis for LA01 & LA02 (Imperial Valley, 1940,Elcentro) 69

Figure 5.8: Incremental time-history displacement response of Vestas wind turbine tower subjected toLA01 & LA02 (Imperial Valley, 1940, Elcentro) at hub height 70

Figure 5.9: Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA01 & LA02 (ImperialValley, 1940, Elcentro) 71

Figure 5.10: Incremental dynamic analysis curves for three damage measures: peak displacement, peakrotation, and residual displacement 74

Figure 5.11: Fragility curves for LA earthquake suite for magnification factor intensity measure 78

Figure 5.12: Fragility curves for LA earthquake suite for PGV intensity measure 78

Figure 5.13: Fragility curves for LA earthquake suite for PGA intensity measure 79

Figure 5.14: Orbit in x-z plane (in mm) for varying damping values of wind turbine tower: 0.5%, 1.0% and

1.5% of critical 81Figure 5.15: Geometry of bolted connection of tower flange 82

Figure 6.1: 2005 NBCC UHS for the Eastern Canada site for 2% in 50 years and average spectra of 4record sets of simulated earthquakes 86

Figure 6.2: Accelerograms of 14 scaled ground motion records for the Eastern Canada site 88

Figure 6.3: Acceleration response spectra for the Eastern Canada earthquake suite for 2% in 50 yearsprobability of exceedance 89

Figure 6.4: 2005 NBCC UHS for the Western Canada site for 2% in 50 years and average spectra of 4record sets of simulated earthquakes 90

Figure 6.5: Accelerograms of 14 scaled ground motion records for the Western Canada site 92

Figure 6.6: Acceleration response spectra for the Western Canada earthquake suite for 2% in 50 yearsprobability of exceedance 93

Figure 6.7: Flowchart of scaling procedure for the Canadian earthquake suites 95

Figure 6.8: Incremental dynamic analysis curves for Western Canada suite 97

Figure 6.9: Fragility curves for Western Canada site for the magnification factor intensity measure 99

Figure A.1: Peak displaced shape of wind turbine tower subjected to LA03 & LA04 (Imperial Valley, 1979,Array #05) 108


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Figure A.2: Incremental time-history displacement response of Vestas wind turbine tower subjected toLA03 & LA04 (Imperial Valley, 1979, Array #05) at hub height (80m) 109

Figure A.3: Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA03 & LA04 (ImperialValley, 1979, Array #05) 110

Figure A.4: Peak displaced shape of wind turbine tower subjected to LA05 & LA06 (Imperial Valley, 1979,Array #06) 111

Figure A.5: Incremental time-history displacement response of Vestas wind turbine tower subjected toLA05 & LA06 (Imperial Valley, 1979, Array #06) at hub height (80m) 112

Figure A.6: Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA05 & LA06 (ImperialValley, 1979, Array #06) 113

Figure A.7: Peak displaced shape of wind turbine tower subjected to LA07 & LA08 (Landers, 1992,Barstow) 114

Figure A.8: Incremental time-history displacement response of Vestas wind turbine tower subjected toLA07 & LA08 (Landers, 1992, Barstow) at hub height (80m) 115

Figure A.9: Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA07 & LA08 (Landers,1992, Barstow) 116

Figure A.10:Peak displaced shape of wind turbine tower subjected to LA09 & LA10 (Landers, 1992, Yermo)

117Figure A.11:Incremental time-history displacement response of Vestas wind turbine tower subjected to

LA09 & LA10 (Landers, 1992, Yermo) at hub height (80m) 118

Figure A.12:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA09 & LA10 (Landers,1992, Yermo) 119

Figure A.13:Peak displaced shape of wind turbine tower subjected to LA11 & LA12 (Loma Prieta, 1989,Gilroy) 120

Figure A.14:Incremental time-history displacement response of Vestas wind turbine tower subjected toLA11 & LA12 (Loma Prieta, 1989, Gilroy) at hub height (80m) 121

Figure A.15:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA11 & LA12 (LomaPrieta, 1989, Gilroy) 122

Figure A.16:Peak displaced shape of wind turbine tower subjected to LA13 & LA14 (Northridge, 1994,Newhill) 123

Figure A.17:Incremental time-history displacement response of Vestas wind turbine tower subjected toLA13 & LA14 (Northridge, 1994, Newhill) at hub height (80m) 124

Figure A.18:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA13 & LA14(Northridge, 1994, Newhill) 125

Figure A.19:Peak displaced shape of wind turbine tower subjected to LA15 & LA16 (Northridge, 1994,Rinaldi RS) 126

Figure A.20:Incremental time-history displacement response of Vestas wind turbine tower subjected toLA15 & LA16 (Northridge, 1994, Rinaldi RS) at hub height (80m) 127

Figure A.21:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA15 & LA16(Northridge, 1994, Rinaldi RS) 128

Figure A.22:Peak displaced shape of wind turbine tower subjected to LA17 & LA18 (Northridge, 1994,Sylmar) 129

Figure A.23:Incremental time-history displacement response of Vestas wind turbine tower subjected toLA17 & LA18 (Northridge, 1994, Sylmar) at hub height (80m) 130

Figure A.24:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA17 & LA18(Northridge, 1994, Sylmar) 131


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Figure A.25:Peak displaced shape of wind turbine tower subjected to LA19 & LA20 (North Palm Springs,1986) 132

Figure A.26:Incremental time-history displacement response of Vestas wind turbine tower subjected toLA19 & LA20 (North Palm Springs, 1986) at hub height (80m) 133

Figure A.27:Orbit in x-z plane (in mm) for Vestas wind turbine tower subjected to LA19 & LA20 (NorthPalm Springs, 1986) 134

Figure A.28:IDA curves for various intensity measures and peak displacement damage measure 135Figure A.29:IDA curves for various intensity measures and residual displacement damage measure 136

Figure A.30:IDA curves for various intensity measures and peak rotation damage measure 137

Figure C.1: Peak displaced shape of wind turbine tower subjected to WCan01 142

Figure C.2: Incremental time-history displacement response of Vestas wind turbine tower subjected toWCan01 at hub height (80m) 143


Figure C.4: Incremental time-history displacement response of Vestas wind turbine tower subjected to

WCan02 at hub height (80m) 145Figure C.5: Peak displaced shape of wind turbine tower subjected to WCan03 146





Figure C.10:Incremental time-history displacement response of Vestas wind turbine tower subjected toWCan05 at hub height (80m) 151

Figure C.11:Peak displaced shape of wind turbine tower subjected to WCan06 152


Figure C.13:Peak displaced shape of wind turbine tower subjected to WCan07 154


Figure C.15:Fragility curves for the Western Canada earthquake suite for PGV intensity measure 156

Figure C.16:Fragility curves for the Western Canada earthquake suite for PGA intensity measure 156


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LIST OF SYMBOLS ANDABBREVIATIONS

AISC American Institute of Steel Construction

BSI British Standards Institution

CEN European Committee for Standardization (Comit Europen de Normalisation)

CSA Canadian Standards Association

GL Germanischer Lloyd

IEC International Electrotechnical Commission

NBCC National Building Code of Canada

CHS circular hollow section

DBE design-based earthquake

DM damage measure

DOF degree-of-freedom

DS damage state

FE finite element

FEA finite element analysis

FEM finite element method

HSS hollow structural section

IDA incremental dynamic analysis

IM intensity measure

LP load pattern

MCE maximum considered earthquake when referring to the Los Angeles area

MF magnification factor

PGA peak ground acceleration

PGV peak ground velocity


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SRSS square root of the sum of the squares

SSI soil-structure interation

UHS uniform hazard spectra

A cross-sectional area

Ab nominal area of bolt

]c[ damping matrix

D outside diameter of a CHS

Dm centreline diameter of a CHS

E Youngs modulus of elasticity

I moment of inertia

Fc elastic compressive buckling stress defined based on Fyand D/t

( ))t(F dynamic load vector

Fu ultimate tensile stress

Fy yield tensile stress

Fy,eff effective yield tensile stress calculated to meet Class 3 D/t limits

g acceleration due to gravity, 9.81m/s2

h height above base

H hub height

hf height of stiffening flange from base of tower

]k[ stiffness matrix

t thickness

tf thickness of stiffening flange

Ti period of mode i

Lr length of member having refined finite element size

]m[ mass matrix


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M moment

MFDS1 magnification factor at damage state 1

Mp plastic moment of a cross-section

Mu moment at ultimate capacity

My yield moment of a cross-section

n factor used in predicting scale factors that account for nonlinearity of response

P load

Pe Euler buckling load

Pu load at ultimate capacity

S elastic section modulus

SAtarg target spectral acceleration

SAsim spectral acceleration of simulated ground motion record

se element size

Tu ultimate bolt capacity

wf width of stiffening flange

)x( nodal displacement vector

)x(& nodal velocity vector

)x(&& nodal acceleration vector

Z plastic section modulus

coefficient used in Rayleigh damping

i modal participation factor of mode i

coefficient used in Rayleigh damping

lateral deflection

max peak lateral deflection

max,avg,DS1 average peak lateral deflection for the first damage state


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res residual lateral deflection

top lateral deflection at top of tower

true-strain; parameter used in slenderness limits

nom engineering strain

u ultimate strain at fracture

i modal damping ratio of mode i

rotation of tower as defined in Section 5.2.1.2

max maximum rotation of tower as defined in Section 5.2.1.2

plan angle of tower in plan view (or top view, or x-z plane)

curvature

p curvature at plastic moment: Mp/EI

average/mean, used in defining fragility curves

true-stress; standard deviation used in defining fragility curves

nom engineering stress

mises Von Mises stress

i mode shape of mode i, normalized to mass matrix

i circular frequency of a mode i


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CHAPTER 1: INTRODUCTION

Wind energy has gained popularity worldwide as many countries aim to increase the production

of clean energy. As Canada follows suit, Canadian design codes are largely adopting international

standards for the design of wind turbine components. However, several gaps are evident due to

Canadas unique environment. One such gap is the assessment of seismic risk pertaining to wind

turbine towers, as the major developments of wind turbines have been in non-seismic areas. The

seismic risk is of particular importance to owners of wind turbine developments, especially wind

turbine farms, since all the towers are identical. This means that a seismic event would affect all the

towers in the same manner if one fails, they all fail. Such a failure would result in severe financial

losses, as well as social implications if wind energy takes over more of the energy production in

Canada.

Thus, the need for research in this area has become evident. Wind turbine towers are differentfrom other structures because they are characterized by a very tall and slender tubular tower. This

geometry results in a structure that cannot respond in a ductile manner, thus the wind turbine

towers capacity when subjected to dynamic loads must be characterized.

1.1 OVERVIEW OFTHESISThis thesis begins with a review of existing literature on wind turbine towers, specifically

pertaining to seismic provisions and behaviour under seismic loads. International design codes for

wind turbines are presented, along with Canadian design codes for steel structures that may be

applicable to wind turbine towers. Some of the current research on seismic response of wind

turbine towers is also presented, noting that none of the existing research has evaluated the seismic

event that may cause failure of a typical wind turbine tower.

Chapter 3 describes the development and validation of a finite element model and methods

employed. Most of the validation analyses are carried out on a simple tubular member of constant

cross-section. Chapter 4 provides details of the typical wind turbine tower analysed in this thesis

and describes a preliminary analysis of the tower that was carried out using pushover analysis. The

tower was then subjected to incremental dynamic analyses, based on an earthquake suite for the Los

Angeles area in California, USA. These analyses were used to derive a methodology for determining

the seismic hazard of steel wind turbine towers. Damage states of the wind turbine tower were

defined and fragility curves were created for each damage state, indicating the probability that a


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given intensity measure will cause exceedance of a particular damage state. Thus, a framework was

set up to assess the seismic hazard for wind turbine towers in any location.

The culmination of this project was the incremental dynamic analysis and seismic risk evaluation

at two Canadian locations. One location was in Western Canada, representing the most severe

seismic hazard in the country, and one was in Eastern Canada, representing a milder seismic hazard

but one where several wind farm developments are underway.

1.2 WINDTURBINETYPE,COMPONENTS, ANDTERMINOLOGYSeveral types of wind turbines exist, but the most prevalent have a horizontal-axis rotor with

three blades and are supported by a thin-walled steel tower. This type of wind turbine is depicted in

Figure 1.1 and the main components of the wind turbine are labeled.

Figure 1.1: Typical horizontal-axis wind turbine

The rotor is made up of blades that are attached to a hub. Wind turbines are often referred to

by their hub height, which represents the height from the base to the centre of the rotor. The hub

height of the wind turbine tower analysed in this thesis is 80 m, which is a typical height. The

nacelle is behind the hub, and it contains the gearbox, generator, shafts, and other machinery. The

Main Wind TurbineComponents:

Rotor

Nacelle

Tower

Foundation


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tower is made of a thin-walled tubular steel monopole and the foundation for onshore wind turbines

is typically an octagonal reinforced concrete slab. As previously mentioned, this thesis focuses on

the steel tower of the wind turbine.


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CHAPTER 2: LITERATURE REVIEW

2.1 INTERNATIONAL STANDARDSSeveral standards for the design and safety requirements of wind turbines exist. The most

significant ones are discussed in this section, and particular attention is given to any seismic design

or analysis provisions.

2.1.1 INTERNATIONAL ELECTROTECHNICAL COMMISSION (IEC)The International Electrotechnical Commission (IEC) is the leading organization that compiles

international standards for electrical technologies. The IEC documents act as a basis for national

standardization and also as a reference for international contracts. Founded in 1906, the IEC did

not become involved in the wind turbine industry until 1988, when a technical committee, TC 88,

was formed to compile guidelines for wind turbines. This technical committee has developed theIEC 61400 series, which is comprised of 10 guidelines that cover various topics related to wind

turbine generators. The bulk of the design process for onshore wind turbines is addressed by Part 1,

Design Requirements.

Furthermore, the IEC specifies project and type certification schemes for wind turbines in the

IEC WT 01 document IEC System for Conformity Testing and Certification of Wind Turbines,

Rules and Procedures. This document refers to all of the IEC 61400 series technical standards,

while also referring to several standards from the International Organization for Standardization

(ISO) (IEC, 2001).

IEC61400-1, Wind Turbines Part 1: Design Requirements

This part of IEC61400 specifies minimum design requirements to assure the engineering

integrity of wind turbines (IEC, 2005). Wind turbine classes are defined based on the reference

wind speed and the turbulence intensity that the wind turbine is expected to experience. The primary

consideration is wind loading, for which several wind conditions are described. Other

environmental conditions are also specified, wherein earthquakes are considered as one of the

extreme other environmental conditions (IEC, 2005). The standard wind turbine classes have no

minimum earthquake requirements, but assessment of earthquake conditions is outlined in Clause

11.6. Seismic analysis may be required depending on site-specific conditions, and earthquake

assessment is not required in locations that are excluded by the local seismic codes due to weak

seismic action. In locations where seismicity may be critical, the seismic loading must be combined


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with a specified operational loading that occurs frequently during the turbines lifetime and that is

considered to be significant enough (IEC, 2005).

IEC 61400 specifies that the seismic loading be based on the ground acceleration for a 475-year

recurrence period and that response spectrum requirements be defined by the local building codes.

The evaluation of the seismic loads may be carried out either in the frequency-domain or in the

time-domain. Furthermore, a simplified conservative approach to calculate the seismic loads is

provided in Annex C, but this approach is only recommended if the tower is the only part of the

wind turbine that will experience significant loading due to seismic action (IEC, 2005).

2.1.2 GERMANISCHERLLOYD (GL)Germanischer Lloyd (GL) is a certification organization based in Germany. They use their own

guidelines, as discussed below, in addition to the IEC standards and the German Institute for

Standardization (DIN) standards to certify wind turbines and their components. Their services are

offered worldwide.

GL Wind 2003, IV Part 1, Guideline for the Certification of Wind Turbine Towers

This guideline is used in the design and certification of wind turbines. It is largely similar to

IEC61400-1, but it also describes the design process for each component of the wind turbine

separately. This guideline outlines the national requirements of several countries: Germany,

Denmark, France, the Netherlands, and India (GL, 2003).

The earthquake requirements in this guideline are very similar to those of IEC 61400-1.

Earthquakes are included in the list of design load cases, with a few minimum load cases specified.

If there are no local regulations regarding earthquake analysis, designers are referred to Eurocode 8

or the earthquake chapter in the American Petroleum Institute (API) recommended practice

document RP 2A (GL, 2003). Similarly to IEC 61400-1, the analysis may be either carried out in the

frequency-domain or the time-domain. The minimum number of modes that must be considered is

three. For analysis carried out in the time-domain, a minimum number of six simulations must be

performed per load case.

2.1.3 DET NORSKEVERITAS (DNV)Det Norske Veritas (DNV) is an independent foundation that was established in Norway, but is

now considered an international body. DNV works with the IEC and other European standards

organizations to provide project certification, type certification, and risk management for the wind


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turbine industry. These certification services are based on IEC WT 01. Aside from its involvement

with the IEC, the DNV develops its own standards, which are used to provide a link between

standards for wind turbines, standards for offshore structures, and several other building codes.

Several documents have been published by the DNV. The one that is most comprehensive for

onshore wind turbines is a guideline written by the DNV and Ris National Laboratory in Denmark,

Guidelines for Design of Wind Turbines (DNV and Ris, 2002). The DNV also publishes

standards, recommended practice documents, and classification notes. The DNV documents that

may be helpful in wind turbine tower design include:

DNV-OS-J101, Design of Offshore Wind Turbine Structures

DNV-OS-J102, Design and Manufacture of Wind Turbine Blades

DNV-OS-C101, Design of Offshore Wind Turbine Structures, General (LFRD Method)

DNV-OS-C201, Structural Design of Offshore Units (WSD Method)

DNV-RP-C201, Bucking Strength of Plated Structures

DNV-RP-C202, Bucking Strength of Shells

DNV-RP-C203, Fatigue Design of Offshore Steel Structures

DNV/Ris, Guidelines for Design of Wind Turbines

These guidelines were created through the cooperation of DNV and Ris National Laboratory

to provide a unified basis for the design of wind turbines. The book provides fairly detailed

guidance on all technical items that need to be covered. It is mostly based on meeting the

requirements of the IEC, and also some Danish, Dutch, and German codes (DNV and Ris, 2002).

The earthquake requirements discussed in these guidelines are very similar to those from the

IEC. Pseudo response spectra are suggested as the method of determining the earthquake loads.

Although accelerations in one vertical and two horizontal directions generally need to be analysed,

the guideline suggests some simplifying assumptions. Since the vertical acceleration is not expectedto create much of a dynamic response, the tower may be analysed using the load created by the

maximum vertical acceleration to determine if buckling will be critical. Furthermore, the two

horizontal directions can be simplified to one horizontal direction, because the dynamic system is

fairly symmetrical. A simple model of the wind turbine is suggested as a vertical rod with a

concentrated mass on top. The mass consists of the nacelle and rotor mass and of the tower

mass (DNV and Ris, 2002). This simplified analysis could be used as a preliminary analysis for


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designers to determine if earthquake loading might be critical and thus if a more detailed analysis is

necessary.

2.1.4 OTHEREUROPEAN STANDARDSOther European standards have been fairly harmonized with the IEC codes and thus further

discussion of these is unnecessary.

2.2 CANADIAN STANDARDSAs the number of wind turbines being constructed in Canada increases, there has been much

discussion regarding which codes are applicable to the design of Canadian wind turbine towers.

Hatch Acres (2006) carried out a code review and gap analysis for wind turbines, assessing several

aspects of design of international and Canadian codes. The seismic provisions of relevant Canadian

design codes are discussed in this section.

2.2.1 CAN/CSA-C61400-1:08,WINDTURBINESPART 1: DESIGN REQUIREMENTSThis standard is almost identical to the IEC standard of the same name, with a few Canadian

deviations. It was adopted by the Canadian Standards Association (CSA) in March 2008.

Furthermore, it replaces the 1987 standard, CAN/CSA-F416:87, Wind Energy Conversion Systems

(WECS) Safety, Design, and Operation Criteria.

Canadian Modifications to IEC61400-1

The CSA-C61400-1 has introduced a few changes to make the IEC61400-1 suitable for Canada.

The main changes are due to the external conditions that the wind turbine will experience. More

severe icing and temperature conditions are acknowledged. The CSA has added several notes to the

earthquake-related clauses of the IEC, instructing designers how to obtain seismic loads, design

spectral accelerations, and seismic design data (CSA, 2008). Additionally, the National Building

Code of Canada (NBCC) is referenced in several instances, one of which is for the determination of

the seismic loads.

The CSA acknowledges that the NBCC does not address earthquake forces acting vertically, and

identifies this as a problem because wind turbines may have vibration modes with significant mass

participation factors in the vertical directions. However, the vertical component of a seismic event

is most likely not significant, but is investigated and discussed in Section 5.3.4. Furthermore, there is

a discrepancy between the recurrence period of the seismic event to be used in design. The


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IEC61400-1, and thus the new CSA-C61400-1, suggests a 475-year recurrence period, whereas the

NBCC requires a 2500-year return period.

2.2.2 CAN/CSAS37-01,ANTENNAS,TOWERS, ANDANTENNA-SUPPORTINGSTRUCTURES

This standard applies to structural antennas and towers. It does have a few requirements

regarding the effects of earthquakes and the dynamic effects of wind, but this codes applicability to

this thesis is mostly related to determining the resistance of the tower (CSA, 2001).

2.2.3 CAN/CSAS473-04,STEEL (FIXED OFFSHORE)STRUCTURESThis is a standard that specifies the requirements for the design and fabrication of fixed steel

offshore structures, but is in the process of being replaced by an adopted ISO standard (CSA,

2009a). It acknowledges that supplementary requirements may be necessary for unusual structures,which would be the case for wind turbine towers. It is more applicable to offshore wind turbines,

although some design information is applicable to onshore wind turbine towers as well, such as the

resistance of large, fabricated slender cross-section tubes under compression and bending (CSA,

2004). It also provides significant information about fatigue details relating to tubular joints and

various connection details.

2.2.4 CAN/CSAS16-09,DESIGN OF STEEL STRUCTURESThis standard provides rules and requirements for the design, fabrication, and erection of steel

structures based on limit states design. It specifically defines steel structures as structural

members and frames, and it is apparent that it is principally intended for buildings (CSA, 2009b).

Although this standard is frequently referenced by other CSA Structural Standards, it is not very

useful for the design and analysis of wind turbine structures. The one area where it may be useful is

for fatigue design, as it provides information regarding several fatigue details that are present in wind

turbine towers.

2.3 BOOKPUBLICATIONSIn recent years, several books about wind turbines have been published, most of which are very

detailed and valuable to designers of wind turbines. However, most also have little or no mention of

the effects of earthquakes on wind turbines. A few books of note are listed here:

Wind Energy Handbook (2001), by T. Burton, D. Sharpe, N. Jenkins, E. Bossanyi


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Wind Energy Explained Theory, Design and Application (2002), by .J.F. Manwell, J.G.

McGowan, A.L. Rogers

Wind Turbines: Fundamentals, Technologies, Application, Economics(2006), by Erich Hau

Aerodynamics of Wind Turbines(2000), by M.O.L. Hansen

2.4 CURRENT RESEARCH ONWINDTURBINETOWERSThe majority of recent publications on wind turbine towers originates from universities and

research centres, with few contributions from the private sector. Some of this research that is

related to seismic behaviour of wind turbines is presented in this section.

2.4.1 COMPARISON OF SEISMICANALYSIS METHODS: FREQUENCY-DOMAIN VS.TIME-DOMAIN

Frequency-domain methods are typically favoured in design due to their ease of implementation.

Time-domain analyses have a higher computational demand and are often used in analysis of

structures, rather than in their design. Time-domain analyses are increasingly being used in the wind

turbine industry.

Currently, several wind turbine simulation software packages exist. The purpose of such

software is to analyse wind turbines under several loading cases to determine the design loads. They

range from basic to very sophisticated and are generally proprietary to companies, which carry out

the analyses and only provide the results. The more sophisticated packages can create a full

aeroelastic model of the wind turbine, including the blades, and subject it to turbulent wind loading.

The newest addition to most of these packages is wave and current loading, as offshore turbines are

becoming commonplace (van Wingerde et al., 2006; Lddecke et al., 2008). A few companies also

recognize the need to incorporate earthquake loading into these software packages, as more wind

turbines are being erected on seismically active sites. Garrad Hassan in the UK is one such

company. Their software, GH Bladed, can apply an accelerogram (real or synthesized) to a model

along with other normal loading (Witcher, 2005). The ground motion is applied in any direction anda secondary ground motion may be applied at 90 to the first. The structural dynamics of the wind

turbine are represented using a limited-degree-of-freedom modal model, and all forces and moments

at specified locations are output, as well as torques at critical locations (Witcher, 2005).

This time-domain method was validated against the frequency-domain, which is more

commonly employed. Witcher concluded that both methods were adequate, but discrepancies arose


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when the system damping was not close to that of the design spectra, which is typically 5%. For

operating wind turbines, the aerodynamic damping is close to 5% (Witcher, 2005), and thus both

methods yield very similar results. For turbines that are not operating, the aerodynamic damping is

much lower. Most building codes do not provide a method to correct the level of damping when

using the frequency-domain method, so the time-domain method provides an advantage over thefrequency method because the correct level of damping can be applied (Witcher, 2005). Therefore,

Witcher (2005) concluded that conducting seismic analysis in the time-domain is acceptable, and in

fact preferred, because the correct aeroelastic interaction can be modelled.

A similar investigation was carried out by Windrad Engineering GmbH and Nordex Energy

GmbH (Ritschel et al., 2003). The method of modal approximation was compared with a time-

domain approach using the simulation program Flex5. The main reason for investigating this

comparison is because they believe modal approximation is not an adequate method for obtaining

design loads, especially the rotor and nacelle loads, as modal approximation ignores any action above

the tower top (Ritschel et al., 2003). Thus, any system modes which might take into account the

interaction of the tower and the blades are not considered. The results of this investigation suggest

that the modal approach is very conservative for the lower part of the tower (Ritschel et al., 2003).

Regarding the machine loads on the nacelle and the rotor, the time-domain method predicts high

vertical forces, which are not predicted by the modal approximation method because the vertical

component is ignored (Ritschel et al., 2003).

Both the time-domain method and the frequency-domain method were deemed to be adequate.

In this thesis, the time-domain method is used as the intent is to obtain information about the

response of the wind turbine tower, rather than to obtain design loads.

2.4.2 SHELL BUCKLINGLocal bucking in the shell of the wind turbine tower using static, buckling, and seismic analyses

was investigated by Bazeos et al. (2002). They also assessed the influence of the door opening on

the overall behaviour of the tower. Furthermore, the effects of soil-structure interaction were also

investigated and are discussed in the next section.

Bazeos et al. (2002) found that the static analysis yielded positive results. The wind turbine was

subjected to pseudo-aerodynamic loads corresponding to survival conditions along with gravity load,

and the maximum stresses were found to be well below yield (Bazeos et al., 2002). The static

analysis also showed acceptable stress values throughout the tower and a maximum horizontal

deflection less than 1% of the total height, which is acceptable. The buckling analysis predicted local


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buckling would occur at 1.33 times the static load (Bazeos et al., 2002). Lastly, seismic analyses were

carried out and the first mode was found to dominate the response to the seismic excitation, as was

expected. The maximum stresses were found to be very low for this analysis as well. Thus, Bazeos

et al. (2002) concluded that seismic analysis does not produce a critical response for this type of

structure. However, the magnitude of the design earthquake was not specified, making it difficult toassess the results of these studies.

Shell buckling was also investigated by Lavassas et al. (2003), where the design of a prototype

steel wind turbine tower was evaluated. Shell buckling was not assessed under seismic loading. It

was concluded that assessment of shell buckling according to design codes is somewhat ambiguous.

Additionally, a simplified linear model was deemed insufficient because the stress concentrations at

the base of the tower are ignored.

2.4.3 DYNAMIC SOIL-STRUCTURE INTERACTION EFFECTSResearchers have identified the importance of analyzing the soil-structure interaction (SSI) when

assessing the seismic resistance of wind turbines. Although the wind turbine tower was identified as

the most important structural component when analyzing dynamic response (Zhao and Maisser,

2006), the interaction between the structure, the foundation, and the soil around was also considered

to be very significant (Bazeos et al., 2002; Zhao and Maisser, 2006).

Time-history analysis was used for the seismic analysis to analyse the soil-structure interaction

effects, as it is applicable to both elastic linear and non-linear analysis. A weak earthquake was used

in the analysis by Zhao and Maisser (2006). The wind speeds, also as a time-history, were used to

determine the thrust and the torque on the tower top. For these loads, Zhao and Maisser (2006)

found that the peak tower displacement was dominated by the wind forces. The inclusion of SSI

resulted in reduced fundamental frequencies of the wind turbine. Thus, it was concluded that soil-

structure interaction has a large influence on the dynamic characteristics of the wind turbine tower,

particularly in areas with flexible soil, and that this interaction should be included in dynamic analysis

of wind turbines (Zhao and Maisser, 2006). This conclusion was also reached by Bazeos et al.

(2002).

Design codes generally specify response spectra depending on the soil characteristics. Thus it

may not be necessary to include SSI effects if seismic analysis is carried out in the frequency-domain.

For time-domain analyses, the soil-structure interaction should be assessed if the wind turbine

structure is erected on flexible soil.


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2.5 SUMMARYThe existing codes offer some guidelines for seismic analysis of wind turbine towers, but

generally more guidance is needed, especially for areas of high seismicity. Some private companies

have identified this need and are incorporating seismic analysis in their wind turbine analysis

software. However, most existing research is concerned with verifying that a given wind turbine can

sustain low or moderate seismic loadings without assessing the limits of the wind turbine towers

seismic capabilities. Therefore, this thesis will focus on characterizing these limits for Canadian

locations by employing the finite element method (FEM).

The following chapter describes an essential part of any project that employs FEM: the

development and validation of the finite element model.


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CHAPTER 3: FINITE ELEMENT MODEL DEVELOPMENT ANDVALIDATION

Numerical models are useful for predicting the behaviour of complex structures or structures

with unusual loading, for which analytical methods are difficult to employ, and can be used to assess

the seismic capabilities of such structures. However, any numerical model must first be deemed

reliable. Material models, element formulations, and failure mechanisms must be verified and shown

to represent realistic behaviours.

Numerical models can be validated using a variety of methods. If experimental results are

available, those results are often used to calibrate the model. Otherwise, the validation must rely on

comparison with values calculated from various theoretical formulations. It must also be shown that

the post-peak behaviour is consistent with experimental results and that the expected failure

mechanism can be captured.

When dealing with wind turbines, very few experimental tests have been performed on the

supporting structure. The validation of any numerical study must hence be segmented, yet must

demonstrate accuracy and reliability.

3.1 GEOMETRY OFWINDTURBINETOWERSTubular steel wind turbine towers are typically very tall and slender. The particular tower that is

discussed and analysed in detail in this thesis is 78 m tall, with a centreline diameter, Dm, of

3650 mm for almost the entire bottom half of the tower. The diameter then tapers down to

2800 mm at the top. The thickness of the tower varies along the height, from 35 mm at the base to

10 mm at the top. Details of the Vestas wind turbine tower are provided in Chapter 4.

For several of the validation analyses, the model was of a simpler member having a constant

diameter and thickness, so that the theoretical closed-form solution for each analysis could be

calculated and compared to the finite element analysis (FEA) result.

3.2 FINITE ELEMENTANALYSIS PROGRAMFor this thesis, ANSYS was chosen to carry out the numerical analyses, as it offers the non-

linear capabilities that are deemed necessary to capture all the aspects of the response of the wind

turbine tower that is being studied.


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3.3 MATERIAL PROPERTIESTypical steel wind turbine towers are made from flat steel plates which are rolled into cylindrical

or conical pieces, and then welded longitudinally (Danish Wind Industry Association, 2003). The

cylindrical or conical pieces are then welded together circumferentially into sections of 20 to 30 m in

height, generally a length that is easily transportable. Each of these sections has flanges at the ends,

and the sections are bolted together on site as the tower is erected. Due to this fabrication process,

the material properties of the tower are similar to cold-formed tubular members. The stress-strain

curve of the material shows a low proportional limit, followed by gradual yielding, no clear yield

plateau, and significant strain hardening.

The material properties used for the analyses in this thesis come from the average of several

coupon tests of cold-formed circular HSS sections performed by Voth (2010). No material data

from an actual wind turbine tower was available. In the ANSYS analyses herein, the true-stress ()vs. true-strain () curve has been employed. This was obtained by modifying the engineering stress-

strain curve (nom, nom), as obtained from a tensile coupon test, in the following manner:

)1ln( nom+= (Equation 3.1)

)1( nomnom += (Equation 3.2)

These equations are only valid until necking of the coupon test occurs. After that point, the

stress distribution is no longer a simple uniaxial case, but a complex triaxial case (Aronofsky, 1951).

The method used by Voth (2010) to determine the post-necking material behaviour was developed

by Matic (1985). It was refined by Martinez-Saucedo et al. (2006), who suggested that the Matic

material properties should only be used in the post-necked region of the stress-strain curve. The

finite element (FE) material properties were thus determined through an iterative process, wherein

several FE analyses of a coupon with the Matic material properties were carried out and compared

with the experimental stress-strain behaviour and rupture. The resulting true stress-true strain curve

is shown in Figure 3.1, as is the experimental stress-strain curve and the onset of necking.


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0

100

200

300

400

500

600

700

800

0 0.1 0.2 0.3 0.4 0.5

Strain (mm/mm)

Stress

(MPa)

True stress-strain curve

Engineering stress-strain curve

Figure 3.1: Engineering and true stress-strain curve from Voth (2010) for cold-formed circular HSS

For the subsequent analyses, three sets of material properties were used, shown in Figure 3.2.

The first, gradual yielding, is the aforementioned true stress-strain curve from Voth (2010). The

second, having a yield plateau, was adapted from Voth (2010) by modifications as shown in Figure

3.2. The third curve is bilinear, was obtained from Elchalakani et al. (2002), and was only employed

in a few analyses that were geometrically comparable to an experimental specimen.

0

100

200

300

400

500

600

700

800

0 0.05 0.1 0.15 0.2 0.25 0.3 0.35 0.4

Strain (mm/mm)

Stress

(MPa)

Gradual Yielding (from Voth, 2010)

Stress-Strain Curve with Yield Plateau

Bilinear (from Elchalakani et al., 2002)

Figure 3.2: Stress-strain curves used in subsequent analyses

E = 211449 MPaFy = 389 MPa

Fu = 833 MPau = 1.1

E = 190900 MPaFy = 408 MPaFu = 510 MPau = 0.27

Yield plateau:0.02 mm/mm

0

100

200

300

400

500

0 0.002 0.004 0.006 0.008

E

1

after necking


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3.4 CHOICE OF ELEMENTSThe elements selected for this analysis are 8-noded shell elements and 20-noded solid elements,

which are further described in the following subsections. Furthermore, mass elements and rigid link

elements were employed, and it was verified that these elements exhibit the desired behaviour.

3.4.1 SHELL ELEMENTSThe wall of the tower was represented with 8-noded shell elements (SHELL281 in ANSYS), as it

was deemed that this element could likely reflect the behaviour of the tower, essentially a thin

conical shell structure. Due to the large diameter-to-thickness ratio along the tower height, the

tower wall acts fairly independently from the rest of the tower and more like a thin shell, with

potential for local buckling. This element has six degrees-of-freedom (DOFs) at each node

translations in the x, y, and z directions, and rotations about the x, y, and z axes. A schematic of the

element and its DOFs is shown in Figure 3.3. The deformation shapes are quadratic, making this

shell element well-suited to model curved shells (ANSYS, 2007). The out-of-plane stress varies

linearly through the thickness and the transverse shear stresses are assumed to be constant through

the thickness.

Figure 3.3: Geometry of shell element used to represent tower walls (ANSYS, 2007)

Another element was also investigated (SHELL93 in ANSYS). Its geometry is identical to that

of the chosen shell element, but has fewer capabilities, albeit still adequate for modelling a tubular

tower. However, the SHELL281 element was chosen, as it has nonlinear stabilization properties,

which improves the stability of local buckling during static analyses (ANSYS, 2007). This element

was found to be more stable during transient analyses as well.

3.4.1.1 CLASSICAL PLATETHEORY

The elastic behaviour of thin plates is described by classical plate theory, also known as

Kirchhoffs plate theory. This theory has several assumptions and limitations (Szilard, 2004).

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Several of the assumptions are analogous to the properties of the shell element previously described.

Classical plate theory is a small-deflection theory, so the transverse deflections are assumed to be

small. The deflection limit is considered to be 1/10th of the thickness. When the deflection exceeds

this limit, some of the assumptions are violated and the theory is no longer as reliable the

behaviour of the plate begins to be governed by membrane action, rather than plate bending action.

A simple numerical analysis showed good agreement of the shell element with the classical plate

theory. The displacements obtained from the FE analyses were slightly higher than those predicted

by classical plate theory. The difference between the FEA and classical plate theory became more

evident when material nonlinearities were included in the analysis. However, once the difference

was significant, the classical plate theory was past its small-deflection limit, which is expected

because the theory does not account for any nonlinearity.

3.4.2 SOLID ELEMENTSThe wind turbine tower has flanges at the base of the tower and at several locations along the

height of the tower, as well as at the top. These allow the tower to be more easily erected and also

stiffen the tower. The flanges were modelled using a 3-dimensional solid element that has 20 nodes

(SOLID95 in ANSYS). The geometry of this element is shown in Figure 3.4. This solid element

was chosen for two reasons: it is well suited to model curved boundaries, as it has a mid-side node;

and it is directly compatible with the shell element that was chosen. One face of the solid element

has the same nodes and node placement as the shell element, resulting in easy and clean meshing ofthe flanges. Due to the connectivity to the shell wall, each flange only has one element through the

thickness, which is also a typical feature of 20-noded solid element modelling.

Figure 3.4: Geometry of 20-noded solid element used to represent flanges (ANSYS, 2007)

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3.4.2.1 ELASTIC BEAMTHEORY

As the main effect of the flanges is to stiffen the shell and thus prevent it from moving in the

circumferential direction, a beam analysis was carried out to evaluate the solid elements flexural

capabilities. The dimensions of the beam were similar to the width and depth of the flanges on a

wind turbine tower, and the length was about 1/6th

of the circumference of a tower having a 3 mdiameter. It was found that even a very coarse mesh captured the Von Mises stress distribution well.

Thus, one element through the thickness of the flange and two elements through the width were

deemed adequate and were used in the modelling of the flanges.

3.4.3 SOLID-SHELL INTERACTIONThere is one discrepancy between the shell and solid elements which arises from the rotational

degree of freedom that does not exist in the solid element. To ensure full connectivity between the

shell and the solid, an overlap of the two elements was used, as shown in Figure 3.5. The increased

mass due to this overlap is not significant, as the wall is quite thin.

(a) (b) (c)Figure 3.5: Geometry of wind turbine tower ring flanges

(a) dimensions(b) node locations(c) area of overlap

3.5 CONNECTION MODELLINGAs discussed in the previous section, the tower is made up of sections that are bolted together

using flanges. The flanges of the Vestas wind turbine tower are shown inFigure 3.6. The flanges

are stocky and stiff, so prying of the connection is not likely to occur. The bolted connections are

thus not modelled. However, the flanges are modelled to simulate the stiffness they lend to the

tower, but are assumed to be fully connected and monolithic. The bolt holes are not modelled, as

the bolt material almost entirely fills the bolt hole. The stiffening effect of these flanges is discussed

further in 3.6.2.2. The maximum stresses at the flange connection during seismic analysis are later

verified to ensure that the assumptions stated here are not violated (Section 5.3.6).


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Figure 3.6: Bolted flange connections of wind turbine tower (Vestas, 2006)

3.6 TUBULARMEMBERS UNDERBENDINGFlexural member cross-sections are classified by many codes based on their cross-sectional

slenderness, which governs a sections ability to carry moment. If elements of the cross-section that

are in compression are too slender, the flexural member may not reach its global flexural capacity,

but may instead buckle locally. Many codes describe the slenderness of tubular elements in flexural

compression in terms of classes, where the limits are defined based on the diameter-to-thickness

ratio (BSI, 2000; CEN, 2005; CSA, 2009b). Other codes also base their limits on the diameter-to-

thickness ratio, but describe flexural members as compact, non-compact, or slender (AISC, 2005;

Standards Australia, 1998).

Class 1 sections, also known as compact, are capable of reaching and maintaining a plastic

moment, and can thus provide sufficient rotation for plastic design. Class 2 sections, sometimes


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Nuta Elena 201003 MASc Thesis

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