1 INTRODUCTION
Flow slides associated with heavy rainfalls are often
devastating, causing many casualties in many parts
of the world. However, their development has not
been well understood until recently, in particular be-
cause the stress paths followed in slopes subjected to
pore pressure increase were not considered and the
concepts of instability of cohesionless soils were not
well understood. When a slope is subjected to pore
pressure increase due to infiltration or rising water
table, total stresses and shear stresses remain essen-
tially constant but effective stresses, mean effective
stress in particular, decrease. This corresponds to a
specific stress path which, examined within the con-
cepts of critical state and instability, provides a gen-
eral framework for understanding slope behaviour in
both loose and dense soil deposits (Chu et al. 2003).
The first part of the paper focuses on the onset of
slope instability (i.e. development of plastic strains),
and post-failure as such is not considered. In the
second part, evidences from physical models and
case histories supporting the framework are pre-
sented. Finally, there is a discussion on the practical
implications of the framework on the evaluation of
slope stability.
2 MECHANISMS LEADING TO SLOPE INSTA-
BILITY
Before considering slopes, it is useful to examine
some aspects of cohesionless soil behaviour. When a
soil is consolidated, isotropically or anisotropically,
in a triaxial cell and then subjected to an undrained
compression test, the stress-strain curve shows a
peak and then moves towards an ultimate state, often
called steady state or critical state. Figure 1 presents
such stress paths in a q/p’cs vs p’/p’cs diagram, in
which p’cs is the mean effective stress at the critical
state. Sladen et al. (1985b) called the line joining the
peaks obtained in CIU tests to the critical state (CS)
the Collapse surface. At a given void ratio and in a
p’ vs q diagram, that would be a collapse line; at a
different void ratio, the collapse line is different.
Lade (1993) defines these lines as instability lines
(IL). The zone bounded by the IL and the critical
state line (CSL) is the zone of instability in which
loose sand becomes unstable when an undrained
condition is imposed (Chu et al., 2003; Wanatowski
et al., 2009a). The IL also appears to be a state
boundary surface that is followed by soil elements
that reach it to move towards the critical state. This
was evidenced by Sasitharan et al. (1993) who per-
formed constant shear drained (CSD) tests, with
constant deviatoric stress and decreasing mean effec-
tive stress from initial stress conditions at point I, on
loose Ottawa sand (Fig. 2). The specimen collapsed
at point Y, well below the (CSL), at a void ratio of
0.809. The collapse (or instability) line correspond-
ing to the same void ratio is also shown on the fig-
ure. It can be seen that yielding in the CSD test has
been obtained when the stress path reached the in-
stability line, indicating that instability is associated
with the same effective stresses and void ratio, re-
gardless of drainage conditions.
Slope instability due to pore water pressure increase
S. Leroueil Laval University, Quebec City, Quebec, Canada
J. Chu Nanyang Technological University, Singapore
D. Wanatowski The University of Nottingham, Nottingham, United Kingdom
ABSTRACT: Chu et al. (2003) developed a framework based on specific stress paths followed in slopes sub-
jected to pore water pressure increase and concepts of soil instability and critical state. This framework applies
to loose as well as dense sands. In this paper, these concepts are slightly extended and confirmed on the basis
of physical model observations and case studies. Practical implications related to the evaluation of slope sta-
bility and the significance of calculated factor of safety are then discussed.
From Sladen et al. (1985b)’s data (see Fig. 1), it
turns out that, if stress conditions on the collapse
surface are described by an angle of strength mobili-
zation φ’mob (e.g. corresponding to M6.9 for the test
consolidated at 6.9 p’cs), this angle is lower than the
critical state friction angle φ’cs and increases towards
its value when the consolidation stress p’/p’cs de-
creases towards one.
Been & Jefferies (1985) proposed the state pa-
rameter Ψ for characterising the behaviour of sands.
Ψ is defined in a e vs p’ diagram (Fig. 3) as the dif-
ference between the current void ratio eo (at a point
such as I) and the void ratio on the steady state line
(or CSL), ess(I), under the same mean effective stress,
Ψ = eo – ess. Positive values of Ψ are associated with
contractant soil behaviour whereas negative values
of Ψ are associated with dilatant soil behaviour.
In the context of slopes subjected to pore water
pressure increase, the stress paths followed are at an
essentially constant shear stress with decreasing ef-
fective stresses. In the laboratory, such stress paths
can be simulated by drained tests with constant q
value and decreasing p’ value called constant shear
drained (CSD) tests (Brand, 1981; Anderson & Rie-
mer, 1995). Such tests had been performed by
Anderson & Riemer (1995) on colluvial soil and by
Santos et al. (1996) on residual soil. On the basis of
the results, Leroueil (2001) indicated that instability
is not controlled by initial stress conditions at point I
(Fig. 3) and the corresponding Ψ value but rather by
the relative position of the point at yielding Y (Fig.
3) with respect to the CSL. Chu et al. (2003) sug-
gested calling this difference the modified state pa-
rameter __
Ψ (Fig. 3). In fact, the value of Ψ in a slope
varies with seasons whereas slope instability and
possibly failure depends on __
Ψ .
Chu et al. (2003) performed a series of triaxial
tests, including CSD tests, on Changi sand having a
mean grain size of 0.30-0.35 mm and a uniformity
coefficient of 2.0. In the interpretation of the tests,
Chu et al. (2003) defined the term instability as be-
haviour in which large plastic strains are generated
rapidly due to the inability of a soil element to sus-
tain a given stress or load. Figure 4 illustrates insta-
bility under CSD tests. Two specimens of loose sand
(ec = 0.94), DR7 and DR10, were anisotropically
consolidated at p’ = 200 kPa (point A for DR7, Fig.
4a) and then sheared at an essentially constant devia-
toric stress at a rate not allowing the development of
pore pressure during the test that was thus fully
drained. From A to B, there were little axial and
volumetric strains (Fig. 4b). However, at point B,
these strains started developing at a faster rate, indi-
cating unstable behaviour. As shown in Figure 4a,
the instability point B is well below the critical state
line, at an angle of strength mobilization φ’mob lower
than φ’cs.
Two specimens of the same sand, but dense (ec =
0.65), DR39 and DR40, were anisotropically con-
solidated at about p’ = 260 kPa (point D for DR39,
Fig. 4a) and then sheared at an essentially constant
deviatoric stress. As shown on Figure 4b, from D to
E, the axial and volumetric strains were small. How-
ever, at point E, the axial strain started increasing,
indicating the onset of instability; as shown in Fig-
ures 4b and 4c, this latter instability was associated
with dilation of the dense soil specimens. It can be
seen that in these cases, instability is reached at an
angle of strength mobilization φ’mob larger than φ’cs.
Figure 4c shows the location of the instability
points for the previously mentioned tests in an e vs
log p’ diagram. The critical state line is also shown
on the figure. It can be seen that the modified state
parameter __
Ψ is equal to about 0.060 for the loose
specimens and about -0.186 for the dense specimens.
From these tests and other triaxial tests performed
on the same sand, Chu et al. (2003) defined the rela-
tionship between the effective stress ratio at instabil-
ity, ηIL = q/p’, and __
Ψ (Fig. 5a). It can be seen that
ηIL, equal to 1.35 at critical state, varies from about
0.70 for __
Ψ > 0.10 to about 1.50 for __
Ψ < - 0.15.
Wanatowski & Chu (2007) and Wanatowski et al.
(2009b) performed plane strain tests on the same
Changi sand and observed a general behaviour simi-
lar to that obtained in triaxial conditions. However,
the critical state lines obtained in both tests were dif-
ferent with, in particular, M values equal to 1.35 for
triaxial tests and 1.16 for plane strain tests. The rela-
tionships between the effective stress ratio at insta-
bility, ηIL, and __
Ψ were also different. However,
Wanatowski et al. (2009a) showed that, when nor-
malised with respect to the M value obtained in the
relevant type of test, triaxial or plane strain, both
curves come on a unique one (Fig. 5b).
From these results, Chu et al. (2003) developed a
general framework for understanding the instability
of slopes in loose or dense sand subjected to pore
water pressure increase. Leroueil (2004) summarised
it by using a q/p’cs vs p’/p’cs diagram in which the
CSL and IL (linear for simplicity) were drawn (Fig.
6). C is the normalised critical state, at the intersec-
tion of the CSL and IL. For normalised q values lar-
ger than the one at C (generally relatively loose
sands for slopes of precarious stability), initial con-
ditions will be at a point such as Ils in Figure 6 and
instability will be reached at a point such as Yls, be-
low the CSL. At Yls, the soil will have a tendency to
move towards its critical state C. As the deviatoric
stress at C is smaller than that due to gravity forces
in the slope (q at Ils), there will be static liquefaction
of the soil and collapse of the slope. Major conse-
quences of this phenomenon are that failure is trig-
gered at an angle of strength mobilization smaller
than the critical state friction angle and that instabil-
ity (at Yls) is followed by an increase in pore water
pressure since p’ decreases. If soil in a slope is at Ids,
at a normalised q value smaller than the one at C
(generally relatively dense sands), and is subjected to
pore water pressure increase, the stress path will
move towards Yds where there is development of
plastic strains and then have a tendency to go to-
wards its critical state C. However, this can only be
achieved if the soil dilates and q/p’cs increases,
which takes time.
This framework can be described in more details
by referring to Figure 7. When a loose sand is
sheared along a q = cst path starting from point I
(Fig. 7a), instability occurs at point Y, on the IL at
the corresponding __
Ψ value. If the pore water pressure
can dissipate freely (i.e. under drained condition),
the stress path will eventually reach the failure state
at point C1, on the critical state line that is also the
failure line for loose sands. During this process,
large axial and volumetric strains will develop and
the void ratio of the soil will decrease (Fig. 7b). If
failure is reached at C1, as there is then equilibrium
between the applied shear stress and the critical state
strength, the available kinetic energy will be small
and the rate of movement should be small (Leroueil
et al., 1996). On the other hand, if the pore-water
pressure cannot dissipate freely, the stress path will
move towards the critical state associated with its
current water content, i.e. C2 under undrained condi-
tions. As the corresponding strength is smaller than
the applied shear stress, the kinetic energy available
at yielding will be large. There will thus be runaway
failure and development of flow.
For dense sand, the stress path moves from point I
to point Y, where the soil becomes unstable (Fig.
7b). For dense sand, instability is reached above the
CSL but below the failure line. If the pore water
pressure can dissipate freely (i.e. under drained con-
ditions), both axial strain and volumetric strain (dila-
tive in that case) will start increasing at point Y, and
the stress state will move towards the failure line at a
point such as F1. If failure is reached, post-failure
could be essentially undrained, with a stress path
moving towards point C4 in Figure 7b. As the corre-
sponding strength may be smaller than the shear
stress applied by gravity forces in the slope, there
may be development of flow at the post-failure
stage. If at point Y, dilation cannot be accommo-
dated, such as for undrained conditions, negative
pore water pressure will develop (see Wanatowski et
al., 2009a) and the stress state will move towards
point C3 (Fig. 7b). The soil will then remain stable;
it could however become unstable and the slope
could possibly fail when pore water pressure will
dissipate. In such a case, failure is delayed.
Figure 7c shows a particular case in which insta-
bility is reached at the same time as the CSL, at
point C5. In that case, instability is associated with
failure. However, as there is then equilibrium be-
tween the applied shear stress and the critical state
strength, the available kinetic energy will be small
and the rate of movement should also be small.
3 EVIDENCE FROM PHYSICAL MODEL TESTS
AND FIELD OBSERVATIONS
The framework previously described has important
implications for slopes: In loose sand, the onset of
failure can be obtained at an angle of strength mobi-
lization, φ’mob, smaller than φ’cs; in such case, and if
essentially undrained the onset of failure is followed
by an increase in pore pressure. In dense sand, the
onset of instability is reached at a mobilized friction
angle, φ’mob, larger than φ’cs; however, instability is
then associated with a tendency of the soil to dilate
before failure can be reached.
That framework has been established for satu-
rated conditions. It is more complex for unsaturated
soils since, with infiltration, matric suction (ua – uw)
decreases, (p – uw) decreases and the strength enve-
lope is lowered. In addition, there does not seem to
have detailed information on what happens to insta-
bility lines with changing suction. However, and as
indicated by several reported case histories, the con-
cepts previously described apply. Olivares &
Damiano (2007) specify that when the soil is suscep-
tible to static liquefaction and is essentially saturated
at the onset of slope failure, post-failure will evolve
into a flow slide as for saturated soils. These authors
also mention that if the soil is not saturated at the
onset of failure, then post-failure may not evolve
into a flow slide but rather into a slide or debris ava-
lanche, with smaller runout distance.
As suggested by the US National Research Coun-
cil (NRC 1985), failure may also result from re-
distribution of void ratio within a globally undrained
sand layer or spreading of excess pore water pressure
in a slope. These latter possibilities are not examined
here. Also, it is tried to avoid in this paper cases
where failure could be associated with erosion or ex-
cavation at the toe of slopes. Such cases involve dif-
ferent stress paths and possibly a quasi-undrained or
partly drained behaviour (e.g. failure at the Jamuna
Bridge, Bangladesh, (Hight et al., 1999) and failure
of the Mississippi riverbanks (Torrey & Weaver,
1984)).
Several cases from the literature are examined
hereunder in comparison with the implications of
this framework.
Nerlerk berm. The Nerlerk berm was constructed
over two seasons in the Beaufort Sea, where the
depth of water was approximately 45 m. The berm
consisted of Ukaler sand core overlaid by Nerlerk
sand. The slope angle of the berm was approxi-
mately 13°. At the turn of July and August 1983,
several slides occurred, involving only Nerlerk sand.
This sand was assumed to have a relative density of
30% (Lade, 1993) and a friction angle at critical state
of 31°. Back analysing the failures, Sladen et al.
(1985b) found a mobilized angle of strength of 13-
16°, indicating instability considerably below the
CSL at a friction angle of 31°.
Coking coal stockpile physical model. Eckersley
(1990) examined flow slides in coking coal stock-
piles. The coal particles ranged from fine sand and
silt sizes to gravel with a critical state friction angle
of 40°. Instability was induced in 1 m high stock-
piles by raising the water level within the slope (Fig.
8a). For the experiment considered here (Experiment
7), the coal was placed for the bottom 400 mm at 9%
water content with no compaction (dry density of 0.7
Mg/m3); the remaining 600 mm was placed dry at a
dry density of 1.0 Mg/m3.
Failure occurred in three stages as indicated in
Figure 8a. Stage 1 comprised two shallow slides
over a 2 s period. It was followed by Stages 2 and 3
that occurred in the 4 following seconds along the
shear zones shown in Figure 8a. Pore water pres-
sures observed during experiment 7 at locations in-
dicated in Figure 8a are shown in Figure 8b. These
pore water pressures became positive during raising
of the water table and were slowly increasing at the
time of failure. What can be seen is that pore water
pressures mostly increased after the onset of failure
defined on the basis of video camera pictures (ar-
rows in Fig. 8b). Eckersley (1990) concluded: “Ex-
cess pore pressures are a consequence of failure ini-
tiation rather than a cause, and static liquefaction is
therefore a post-failure phenomenon”. He also back
calculated a mobilized angle of strength at the onset
of failure of 24-27°, much less than the critical fric-
tion angle of 40° obtained from laboratory tests.
Centrifuge tests. Zhang & Ng (2003; also reported
by Ng, 2008) have performed centrifuge tests for ex-
amining the failure mechanisms of sandy slopes sub-
jected to rainfall and rising water table. The material
used was Leighton Buzzard fine sand that shows
pronounced strain-softening in undrained triaxial
shear tests performed on loose specimens. The
model was 305 mm high with a slope of 29.4° and
built with the soil at a relative compaction of 68%.
However, when the model was subjected to an ac-
celeration of 60 g, the slope was densified to 80% of
the maximum relative compaction and flattened to
24°. At 60 g, the slope was de-stabilised by rising
water level and the soil liquefied statically and
flowed. Unfortunately, the pore water pressures re-
ported by Ng (2008) are not very detailed.
Flume tests. Wang & Sassa (2001) and Damiano
(2003) examined rainfall-induced flow slides in
laboratory flume tests. The two materials tested were
silica sandy silt and pyroclastic sand respectively.
Failure was induced by sprinkling water on the sur-
face of the soil models. Figures 9a and b show typi-
cal results obtained on these materials in loose con-
ditions. It can be seen in both cases that pore water
pressures slowly increased before the onset of failure
and rapidly increased after, similarly to Eckersley
(1990) observation (Fig. 8b).
Wachusett Dam. The construction of the dam was
completed in 1907 and failure occurred in the up-
stream slope during the first reservoir filling. Ac-
cording to Olson et al. (2000), the upstream fill con-
sisted primarily of fine sands that were placed
without compaction. Failure occurred in drained
conditions as filling of the reservoir was very slow
and was followed by a flow of the upstream fill soils
over a distance of about 100 m into the reservoir.
Back analysis of the failure performed by Olson et
al. (2000) indicate that failure occurred at average
shear strength between 37.6 and 41.9 kPa, corre-
sponding to a mobilized angle of strength that was
close to the critical state friction angle of the sand
(30°). Kinetics analysis also indicated a post-failure
strength of approximately 16 kPa, slightly less than
half of the strength at failure. A practical conclusion
from Olson et al. (2000) is that sandy fills that sub-
sequently will be saturated should not be placed
without compaction.
Cernivara landslide. The Cervinara landslide oc-
curred on 16 December 1999 along a steep slope of
about 40° covered by about 2.5 m of pyroclastic soils
and developed into a flow slide that travelled over
several kilometres, stroke several houses and killed 5
persons. The volcanic soils include layers of pumice
and volcanic ash classified as sand. Detailed studies
of the mechanical behaviour of volcanic ash have
been performed at the Seconda Università di Napoli.
Olivares & Picarelli (2001, 2003) showed that: (a)
strength significantly increases with matric suction;
(b) critical friction angle is of 38°; (c) the material is
highly susceptible to liquefaction when saturated
(see Fig. 10).
Suction measurements made on the same site give
values of 20 to 50 kPa during the dry season and 4 to
8 kPa during the wet season. The stability of the
slope was thus insured by matric suction. However,
it is thought by Olivares & Picarelli (2003) that on
16 December 1999 slope failed because suction van-
ished. The mobilized angle of strength was then
probably very close to the critical state friction angle
of the soil. In addition, as volcanic ash is very sus-
ceptible to liquefaction (Fig. 10), the landslide
turned into a flow slide.
Sau Mau Ping landslides. Two landslides involving
man-made fill slopes occurred after heavy rainfall on
18 June 1972 and 25 August 1976 at Sau Mau Ping
in Hong Kong (Ho & Sun, 2009). For the 1972 land-
slide, the slope was 40 m high and inclined at 34°
with the horizontal, and the debris slid down at high
velocity, killing 71 persons; the 1976 landslide took
place in a 35 m high and 33° steep slope, and turned
into mud flow, killing 18 persons. The slopes were
made up of decomposed granite. Investigation of the
1976 landslide showed that the fill was extremely
loose (ρd = 1.35 Mg/m3, corresponding to about 75%
standard compaction) to a depth of at least 2 m be-
low the slope; beyond the crest, ρd was also low and
variable, between 1.65 and 1.2 Mg/m3 (90% and
70% relative compaction), to a depth of 7 m and
about 1.5 Mg/m3 down to 20 m. Laboratory direct
shear box tests carried out under a vertical consoli-
dation stress of 25 kPa showed a contractant behav-
iour for dry densities lower than 1.5 Mg/m3; other
laboratory tests indicated a critical state friction an-
gle of 36.8°.
It was estimated that the fill could have been satu-
rated to depths between 2 m and 6 m under the 25
August 1976 rainstorm. Also, numerical analyses
showed that failure could have been triggered if 3 m
of loose fill became saturated, but for strength condi-
tions below the critical state strength envelope. The
recommendation following this investigation was
that the soil has to be compacted to not less than
95% of standard maximum dry density for man-
made fills (Ho & Sun, 2009).
Evidences of dilatant behaviours. The possibility of
dilatant behaviour of soil masses prior to some post
failure movements is also supported by observations:
(a) Casagrande (1975) indicates that prior to lique-
faction and flow of large masses of rather dense
granular talus in the alps, brooks emerging from the
toe of the talus stopped flowing; (b) Fleming et al.
(1989) report observations of time lags between the
beginning of landslide movements and the initiation
of debris flows; (c) in three slides that Harp et al.
(1990) triggered by artificial subsurface irrigation,
they observed abrupt decreases in pore pressure 5 to
50 minutes before failure.
4 DISCUSSION
The cases presented in the previous section confirm
the soil model and its implications for slopes. In
loose sandy soils, the onset of failure can be reached
at an angle of strength mobilization lower than the
critical state friction angle (Nerlerk berm; Coking
coal stockpile; probably Sau Mau Ping slope) or
close to the critical state friction angle (Wachusett
dam; Cervinara slope); where measured (and if the
soil is not too pervious), instability is followed by
pore water pressure increase (coking coal stockpile;
flume tests); in all cases, there was strain-softening
of the soil and development of post-failure flow
slide; in all cases also, development of flow slide is
very rapid. In the cases of dense soils, there is evi-
dence of dilatant behaviour and delay between the
onset of instability (development of plastic strains)
and failure as such. For the two slope cases reported
here that were generally unsaturated (Cervinara and
Sau Mau Ping), it seems that failure was initiated
when the matric suction was close to zero. It can be
thought however that some slopes failure can be
reached when they are still unsaturated (in loess in
particular); Olivares & Damiano (2007) indicate
however that in these conditions, the possibility to
have a flow slide is smaller.
4.1 Practical application of the framework
The framework and its implications being accepted,
the two main practical questions are as follows:
Considering instability, what are the mobilizable
strength parameters that should be considered in sta-
bility analyses of a slope subjected to pore water
pressure increase? What is the representativeness of
the factor of safety calculated by effective stress
analyses? Even if extremely important, post-failure
behaviour is not considered here.
A major practical question concerns the location
of the instability line relative to the critical state line
or, in other words, what is the angle of strength mo-
bilization to be considered. The answer is given by a
diagram such as Figure 5a and the ratio between ηIL
at relatively large __
Ψ values and M. It seems how-
ever that the ηIL/M vs __
Ψ relationship is soil specific.
ηIL/M at large __
Ψ values is about 0.5 for Changi sand
(Fig. 5b), about 0.55 for Leighton Buzzard sand (Fig.
1) and about 0.9 for Cervinara volcanic ash (Fig.
10). Testing 3 different Japanese sands, Orense et al.
(2004) concluded that instability was reached at mo-
bilized friction angle φ’mob such that tan φ’mob = 0.73
to 0.83 tan φ’cs. ηIL/M thus appears to be variable
from soil to soil and also with__
Ψ . Also, for the time
being, there is not enough data available to be able to
correlate ηIL/M with some physical characteristics of
the soils.
A simplified approach of the problem does not
seem to be accessible for the time being. On the
other hand, it appears difficult to develop a rational
and practical methodology for the approach pro-
posed by Chu et al. (2003) and described here. There
are several reasons for that: (a) the number of tests
that is necessary for determining the CSL and the ILs
at different void ratios is important; (b) natural soil
variability may also be a difficulty; (c) if reconsti-
tuted soil is used, the mode of preparation of the
specimens may significantly influence the test results
(e.g see Vaid et al., 1995). It is thus recommended,
when possible, to take undisturbed samples, recon-
solidate them under stresses close to in situ stresses
and then shear them at constant deviatoric stress and
decreasing mean effective stress, in CSD tests. It is
thought that the test results are the best indicators of
the behaviour of a slope under increasing pore water
pressure and can be directly used. For comparison, it
is also suggested to define φ’cs for the considered
soil.
4.2 Representativeness of a calculated factor of
safety
Another practical aspect is the representativeness of
a calculated factor of safety for a given slope. As
previously indicated and as illustrated by the stress
paths IY in Figs. 2, 4 and 7a, the initiation of failure
may be obtained for an angle of strength mobiliza-
tion smaller than φ’cs in loose sandy soil. This phe-
nomenon is amplified by conventional limit equilib-
rium stability analyses. In these analyses, the factor
of safety is calculated by comparing the applied
shear stress to the shear stress at failure under the
same normal effective stress, which implicitly as-
sumes an effective stress path such as IG in Figure
11 (see Tavenas et al., 1980), thus very different
from the stress path leading to failure, i.e. IY. The
calculated factor of safety for a slope in loose sandy
soil may thus significantly overestimate the real sta-
bility; for the case schematised in Figure 11, that
would mean a calculated factor of safety of about 2
(from I to G) whereas the slope is in fact close to
collapse (from I to Y).
4.3 Solutions to decrease the possibility of instability
and flow slides
As suggested by Olson et al. (2000) and Ho & Sun
(2009), a solution for decreasing the possibility of
soil instability and liquefaction is to increase the an-
gle of strength mobilization by compaction. Ng
(2008) and other authors suggest reinforcement of
the slope by methods such as nailing.
5 CONCLUSION
The framework established by Chu et al. (2003) and
re-examined here provides a unified way to study the
instability, failure and post-failure mechanisms of
loose and dense sandy slopes. For loose cohesionless
soils, instability is reached at an angle of strength
mobilization that is smaller than the critical state
friction angle. If perfectly drained, failure is reached
at the critical state with a rate of movement that
should be small. If not perfectly drained, soil insta-
bility is followed by pore water pressure increase,
flow and runaway failure. For dense soils, instability
(development of plastic strains) is reached at an an-
gle of strength mobilization slightly larger than the
critical state friction angle. If perfectly drained and
dilation allowed, failure is reached on the failure line
and may be followed by runaway post-failure. If di-
lation cannot be accommodated, negative pore pres-
sures develop and the slope will not fail.
Observations made in physical model tests and in-
terpretations of case histories confirm this frame-
work. The practical application of these concepts is
however difficult to apply and it is suggested, when
possible, to take undisturbed soil samples, recon-
solidate them under in situ stresses and then shear
them in CSD tests in order to evaluate their behav-
iour. It is also shown that calculated factor of safety
of slopes in loose sand can strongly overestimate real
stability.
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