Tall Bldg Design

7/27/2019 Tall Bldg Design

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CONCEPT OF STRUCTURAL DESIGN AND EVALUATION OF

MULTYSTORY BUILDING UNDER WIND AND SE ISMIC LOADS

By

BASSAM BLAL

Presented to the Faculty of Civil Engineering

Technical University of Engineering Bucharest Romania

DISSERTATION

For the Degree

DOCTOR OF PHILOSOPHY

In

Structural Design Engineering

In

Technical University of Civil Engineering-Bucharest

Scientific

Prof.univ.dr,ing. Constantin PAVEL

JUNE 2010


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TABLE OF Contents

Chapter A. Multistory and Tall Buildings...................................4

A.1. INTRODUCTION.......4

Chapter 1. Wind Load .....................................................................5

1.1. DESIGN EXAMPLE UBC 1997... 6

Chapter 2. Seismic Design ............................................................... 7

2.1. INTRODUCTION .... 7

Chapter 3. Advanced Earthquake Engineering

Performance Based Design ....................................... 10


3.2. BASIC METHODOLOGY OF PBSD .... 11

3.3. PERFORMANCE OBJECTIVES .... 11

Chapter 4. Case Studies ...................................................................14


4.2. CASE STUDY I - 7 Story RC BUILDING .... 14

4.2.1. Building Description .... 14

4.2.2. Building Instrumentation .... 15

4.2.3. Description of Structural Model .... 16

4.2.4. Nonlinear Time-History and Pushover Analysis Results .... 16

4.3. CASE STUDY II - 10 Story RC BUILDING .... 234.3.1. Building Description .... 23




4.4. CASE STUDY III - 13 Story RC BUILDING .... 27


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4.4.1. Building Description .... 27




Chapter 5. Research Synthesis, Conclusions and

Recomandations...........................................31

References ..................................33

Authors CV ..................................35


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A- MULTISTORY AND TALL BUILDINGS

A-1. INTRODUCTION

While the world is full of interesting structures, large and small, old and

modern, the most eye-catching and the ones that instill the greatest sense of wonder in the

onlooker are the modern skyscrapers. They are monuments of power and prestige, supreme

achievements in engineering and design, comforting landmarks, testimonials to the human

spirit, and public relations at the highest level. When considering skyscrapers, until

recently, the observer was drawn to great cities such as New York and Chicago. Today,

after a century during which New York and Chicago went unchal- lenged as home to the

worlds tallest modern buildings, the crown has been snatched first by Kuala Lumpurs

twin Petronas Towers, then by Taipeis 101 Tower, and recently by the Burj Dubai, the 162

floor, mostly residential concrete tower in Dubai.In New York, the Empire State Building was completed in 1931 and with 102 stories

stood at no less than 1250 ft height (see Figure A.1. a). Among the many astonishing

features of the building is the fact that it has no less than 73 elevators, although not all

traverse the full height. Amazingly it was built in the relatively short period of 410 days.

After the completion of the Empire State Building, architects began to explore more

extensive use of materials such as glass and metal to replace the traditional masonry

cladding of the building. The ultimate expression of the trend were the twin towers of the

World Trade Center (WTC), as shown in Figure A.1.b, which became the worlds tallest

building in 1972 with each tower reaching up to 1368 ft and to 1727 ft with the addition

of TV and other antennas.

To turn to the worlds tallest towers, it might be surprising to find that the threehighest towers in the world lie outside of China and the United States. In Kuala Lumpur, the

capital of Malaysia, are the Petronas Towers standing side by side at a height of 1483 ft (see

Figure A.2). The accolade of the worlds tallest building, not accounting for yet to be

completed Burj Tower, goes to Taipei 101 Tower, which stands at 1671 ft in Taipei, the

capital of Taiwan.

The Burj Dubai shown in Figure A.3 is a super-tall skyscraper in Dubai, the United

Arab Emirates. The tower is composed of a Y-shaped floor plan with setbacks occurring at

each segment in an upward-spiraling pattern, decreasing the cross section of the tower as it

reaches toward the sky.

Tall buildings have fascinated humans from the beginning of civilization as

evidenced by the pyramids of Giza, Egypt; the Mayan temples of Tikal, Guatemala; and theKutub Minar of Delhi, India. The motivation behind their construction was primarily for

creating monumental rather than human habitats. By contrast, contemporary tall buildings

are primarily a response to the demand by commercial activities, often developed for

corporate organizations as prestige symbols in city centers.

The Burj Dubai shown in Figure A.3 is a super-tall skyscraper in Dubai, the United

Arab Emirates. The tower is composed of a Y-shaped floor plan with setbacks occurring at

each segment in an upward-spiraling pattern, decreasing the cross section of the tower as it


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reaches toward the sky. Four season Hotel shown as a multistory modern building in

Damascus, Syrian Arab Republic.

Four season Hotel-Damascus-Syria Burj Dubai

FIGURE A.3

CHAPTER 1 Wind Loads

1.1. DESIGN

CONSIDERATIONS

In designing for wind, a building cannot be considered independent of its surround-

ings. The influence of nearby buildings and land configuration on the sway response of

the building can be substantial. The sway at the top of a tall building caused by wind may

not be seen by a passerby, but may be of concern to those occupying its top floors. There


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is scant evidence that winds, except those due to a tornado or hurricane, have caused major

structural damage to new buildings. However, a modern skyscraper, with lightweight

curtain walls, dry partitions, and high-strength materials, is more prone to wind motion

problems than the early skyscrapers, which had the weight advantage of masonry partitions,

heavy stone facades, and massive structural members.

To be sure, all buildings sway during windstorms, but the motion in earlier tallbuildings with heavy full-height partitions has usually been imperceptible and certainly

has not been a cause for concern. Structural innovations and lightweight construction

technology have reduced the stiffness, mass, and damping characteristics of modern

buildings. In buildings experiencing wind motion problems, objects may vibrate, doors

and chandeliers may swing, pictures may lean, and books may fall off shelves. If the

building has a twisting action, its occupants may get an illusory sense that the world

outside is moving, creating symptoms of vertigo and disorientation. In more violent

storms, windows may break, creating safety problems for pedestrians below. Sometimes,

strange and frightening noises are heard by the occupants as the wind shakes elevators,

strains floors and walls, and whistles around the sides.

Following are some of the criteria that are important in designing for wind:

1. Strength and stability.

2. Fatigue in structural members and connections caused by fluctuating wind

loads.

3. Excessive lateral deflection that may cause cracking of internal partitions and

external cladding, misalignment of mechanical systems, and possible perma-

nent deformations of nonstructural elements.

4. Frequency and amplitude of sway that can cause discomfort to occupants of

tall, flexible buildings.

5. Possible buffeting that may increase the magnitude of wind velocities on

neighboring buildings.

6. Wind-induced discomfort in pedestrian areas caused by intense surface winds.

7. Annoying acoustical disturbances.

8. Resonance of building oscillations with vibrations of elevator hoist ropes.

1.1. Design Examples, UBC 1997

Eleven-Story Building: UBC 1997.

Given.

Eleven-story communication building deemed necessary for post-disaster

emergency communications,Iw = 1.15

Building height 120 ft (36.6 m) consisting of 2 bottom floors at 15 ft (4.6 m)

and 9 typical floors at 10 ft (3.05 m)

Exposure category = C

Basic wind speed V= 100 mph


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Building width = 60 ft

Required. Design wind pressures on primary wind-resisting system.

Solution. The design pressure is given by the chain equation

p =CeCq qsIw

The values of Cethe combined height, exposure, and gust factor coefficienttabulated in Table 1.4are taken directly from Table 1.2. Note that for suction on theleeward face, Ce is at the roof hight, and is constant for the full height of the building. Thewind pressure qs corresponding to basic wind speed of 100 mph is given by

qs = 0.00256V2

qs = 0.002561002 = 25.6psfThe values of pressure coefficient Cq, obtained using the normal force method

(Method 1), are 0.8 for the inward pressure on the windward face, and 0.5 for the suction on

the leeward face. Because the building is less than 200 ft (61 m), the combined value of0.8 + 0.5 = 1.3 may be used throughout the height to calculate the wind load on theprimary wind-resisting system. Observe that Method 2 (projected area method) yields thesame value ofCq = 1.3.

Design pressures and floor-by-floor wind loads are shown in Table 1.4. Notice thatthe wind pressure and suction on the lower half of the first story (between the ground and 7.5

ft aboveground) is commonly considered to be transmitted directly into the ground. Thewind load at each level is obtained by multiplying the tributary area for the level by theaverage of design pressures above and below that level.

Analytical Procedure: Step-by-Step Process

Design wind pressure or suction on a building surface is given by the equation:

Pz =qzGfCp (1.2.5.3)

CHAPTER 2 Seismic Design

2.1. INTRODUCTION

The seismic analysis and design of buildings has traditionally focused on reducingthe risk of loss of life in the largest expected earthquake. Building codes have based their

provisions on the historic performance of buildings and their deficiencies and have developedprovisions around life safety concerns, i.e., to prevent collapse under the most intense earth-quake expected at a site during the life of a structure. These provisions are based on the

concept that the successful performance of buildings in areas of high seismicity depends on acombination of strength, ductility manifested in the details of construction, and the presence


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of a fully interconnected, balanced, and complete lateral-force-resisting system. In regions of

low seismicity, the need for ductility reduces substantially. In fact, in some instances,strength may even substitute for a lack of ductility. Very brittle lateral-force-resisting systems

can be excellent performers as long as they are never pushed beyond their elastic strength.

In general, most earthquake code provisions implicity require that structures be able to resist

1. Minor earthquakes without any damage.

2. Moderate earthquakes with negligible structural damage and some nonstruc-

tural damage.

3. Major earthquakes with some structural and nonstructural damage but without

collapse. The structure is expected to undergo fairly large deformations by

yielding in some structural members.

It is important to distinguish between forces due to wind and those induced by

earthquakes. Earthquake forces result directly from the distortions induced by the motion ofthe ground on which the structure rests. The magnitude and distribution of forces and

displacements resulting from ground motion is influenced by the properties of the structureand its foundation, as well as the character of the ground motion.

An idea of the behavior of a building during an earthquake may be grasped by

considering the simplified response shape shown in figure 2.2.1. As the ground on which

the building rests is displaced, the base of the building moves with it. However, the

building above the base is reluctant to move with it because the inertia of the building mass

resists motion and causes the building to distort. This distortion wave travels along the

height of the structure, and with continued shaking of the base, causes the building to

undergo a complex series of oscillations.

Although both wind and seismic forces are essentially dynamic, there is a funda-

mental difference in the manner in which they are induced in a structure. Wind loads,

applied as external loads, are characteristically proportional to the exposed surface of astructure, while the earthquake forces are principally internal forces resulting from the

distortion produced by the inertial resistance of the structure to earthquake motions.

The magnitude of earthquake forces is a function of the mass of the structure rather

than its exposed surface. Whereas in wind design, one would feel greater assurance about

the safety of a structure made up of heavy sections, in seismic design, this does not

necessarily produce a safer design.

Structures that are built into the ground and extended vertically some distance above-

ground respond as either simple or complex oscillators when subjected to seismic ground

motions. Simple oscillators are represented by single-degree-of-freedom systems (SDOF),and complex oscillators are represented by multidegree-of-freedom (MDOF) systems. A

simple oscillator is represented by a single lump of mass on the upper end of a vertically

cantilevered pole or by a mass supported by two columns, as shown in figure 2.11.2.

The idealized system represents two kinds of structures:

1) a single-column structure with a relatively large mass at its top;

2) a single-story frame with flexible columns and a rigid beam.


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Figure 2.11.2 Idealized single-degree-of-freedom system

The stiffness K of the system is the force F divided by the corresponding

displacement

. If the mass is defl

ected and then suddenly released, frequency of freevibrations is called natural or fundamental frequency of vibration. The reciprocal of frequency

is the period of vibration. It represents the time for the mass to move through one

complete cycle. The period Tis given by the relation:

K

M2T = (2.11.1)

In an ideal system having no damping, the system would vibrate forever (figure

2.11.3). In a real system, where there is always some damping, the amplitude of motion will

gradually decrease for each cycle until the structure comes to a complete stop (figure 2.11.4).

The system responds in a similar manner if, instead of displacing the mass at the top, a

sudden impulse is applied to the base.

Figure 2.11.3 Undamped free vibrations of a single-degree-of-freedom system

Buildings may be analyzed as multidegree-of-freedom (MDOF) systems by lumpingstory-masses at intervals along the length of a vertically cantilevered pole. During vibration,

each mass will deflect in one direction or another. For higher modes of vibration, somemasses may move in opposite directions. Or all masses may simultaneously deflect in the


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same direction as in the fundamental mode. An idealized MDOF system has a number of

modes equal to the number of masses. Each mode has its own natural period of vibration witha unique mode shaped by a line connecting the deflected masses. When ground motion is

applied to the base of the multimass system, the deflected shape of the system is acombination of all mode shaped, but modes having periods near predominant periods of the

base motion will be excited more than the other modes. Each mode of a multimass systemcan be represented by an equivalent single-mass system having gener alized values M and K

for mass and stiffness. The generalized values represent the equivalent combined effects ofstory masses mi and stiffness ki. This concept, shown in figure 2.11.4, provides a

computational basis for using response spectra based on single-mass systems for analyzingmultistoried buildings. Given the period, mode shape, and mass distribution of a multistoriedbuilding, we can use the response spectra of a single-degree-of-freedom (SDOF) system for

computing the deflected shape.

Figure 2.11.4Damped free vibration of a single-degree-of-freedom system

CHAPTER 3 Advanced Earthquake EngineeringPerformance Based Design

3.1. INTRODUCTION

Recently, researchers have adopted and developed a new approach in the earthquakeresistant design of the structures instead the code-to-follow empirical formulations, thismodern tendency is the performance-based seismic design PBSD or PBEE, which is an

attempt to design buildings with an expected degree of damages, or in other words, to predict

the response of a building with a predictable seismic performance. The PBEE was developedand used in the design and construction of vehicles, after was adopted by earthquakeengineering researchers.

PBSD implies design, evaluation, and construction of engineered facilities whoseperformance under common and extreme loads responds to the diverse needs and objectives

of owner-users and society. PBSD is based on the premise that performance can bepredicted and evaluated with sufficient confidence for the engineer and client jointly to

make intelligent and informed decisions based on building life-cycle considerations rather


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Figure 3.3.1 Performance-based design process (source: Ronald O. HAMBURGER) [2-2])

Table 3-1 of ATC-40 gives the recommended combinations of SP and NP to frombuilding performance levels.Now the ATC-58 [2] project (will be done in the near future) hasan important goal, is to utilize performance objectives that are quantifiable and predictable,

as well as meaningful and useful for the decision makers who must select or approve theobjectives used as a basis for design (Hamburger) [2]. However, now the decision makers

view seismic performance from a different perspective and select desired performance usingdifferent decision making processes.

Figure 3.3.2 Structural Performance level and Non-Structural performance level


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1. For the global building response, verify:

The lateral force resistance has not degraded by more than 20% of the peak

resistance.The lateral drift limits satisfy the limits given in the table 3.7.2.

2. Identify and classify the different elements in the building in the following types:

beam-column frames, slab-column frames, solid walls, coupled walls, perforatedwalls, punched walls, floor diaphragms and foundations.

3. Identify all primary and secondary elements.

4. For each element type, identify the critical components and actions to check asdetailed in Chapter 11 of ATC-40 [4].

5. The strength and deformation demands at the performance point should be equal to orless than the capacities detailed in Chapter 11 of ATC-40.

6. The performance of secondary elements (such as gravity load carrying members not

part of the lateral load resisting system) are reviewed for acceptability for thespecified performance level.

7. Non-structural elements are checked for the specified performance level.

Table 3.7.2

To account for the problem of higher mode effects, the pushover analysis should be

performed using corresponding force distribution (ATC-40). Another method, as mentioned

above, which is the Modal Pushover Analysis (MPA) developed by Chopra and Goel [8] canbe used with, which gives a good accuracy and accounts for higher modes effects.

Furthermore, by the development of the CSM, Chopra and Goel proposed also to usethe inelastic spectrum developed by Newmark and Hall (see [9]) which opens the challenge

in this field.It should be noted that in this section an insight is given on the ATC-40 methodology,

this is doesnt exclude the existence of other methods that are cited in the FEMA-273/274, or

in another technical reports.


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CHAPTER 4 Case Studies

4.1. INTRODUCTION

This chapter shall address the analysis of three multi-story reinforced concretestructures, designed and detailed according to U.S. design codes from 60s. All three

structures are part of seismic instrumentation programs for civil buildings NSMIN(National Strong Motion Instrumentation Network), CSMIP (California Strong Motion

Instrumentation Program), operated by CDMG (California Division of Mines and Geology)and USGS (U.S. Geological Survey). All three constructions are placed in Californian

seismic zone from the south-west of USA.For all three structures seismic response records ofNorthridge 1994 and Whittier Narrows 1987 earthquakes are available.

The structures analyzed in this chapter are as it follows:

I. 7 story RC frame structure, without basement.

II.

Dual moment-resisting frame coupled with shear walls elevenstories (S+10 above ground stories) structure. Along direction of

analysis, horizontal loads are undertaken by moment-resistingframes.

III. 15 story moment-resisting frames (2 basements + 13 stories above

ground).The available seismic response records, allow calibration and validation of dynamical

characteristics and strength capacity of the three structures, by comparing calculated andrecorded response.

For a better and more precise evaluation of seismic performance of the three structures,their seismic response to Vrancea 1977 and Al Aqaba 1995 earthquakes will be analyzed. For

Vrancea 1977 earthquake, INCERC - Bucharest N-S component is considered. For Al Aqaba

1995 earthquake, N-S component is considered. The accelerograms of Vrancea 1977 and AlAqaba 1995 earthquake were gradually scaled up to the level corresponding to first columnfailure (max > 4pl in a plastic hinge), or to a local failure mechanism appearance.

One of the aims of numerical simulations in this chapter is to asses the exactity of highcomplexity analyses methods (pushover and nonlinear dynamic analyses) in describing

earthquake structural response of structures. Nevertheless, the possibility of the abovementioned methods to detect seismic response pecularities and structural design faults is

investigated as well.

4.2. CASE STUDY I 7 Story RC Building

4.2.1. Building description

The construction Van Nuys 7 story hotel, is located in the district of Van Nuys (SanFernando Valley), 8244 Orion Avenue at latitude 34.321 and longitude 118.471 W. The

building was designed (according Los Angeles City Building Code 1964 edition) in 1965and constructed in 1966. The structure is cast-in-place. A photograph of the building is

shown in figure 4.2.1.1.


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In-plane dimensions of the building are about 160 x 62 (48.77 x 18.90 m). Total

height of the building is 65.7 (20 m). The buildings has three spans (201 2010 201)and eight bays (8 x 18-9). The first story height is 4.12 m, the height of the rest of the

stories is 2.65 m.

(a) (b)

Figure 4.2.1.1 General view of the Holliday Inn 7 story hotel (a),

3D view of structural model (b)

The structural system comprises moment-resisting frames and a flat slab-columnsystem. The moment resisting frame is designed to carry most of lateral loads. It is placed at

the exterior of the building. Exterior frames consist of 14 x 20 (356 x 508 mm) inchesconstant in cross-section columns, 16 x 30 inches (406 x 762 mm) beams at second story and

16 x 22 inch (406 x 0.559 mm) beams at the rest of the stories. The interior columns aresquared - 20 inches in dimension at first story and 18 inches (406 mm) in dimension at the

rest of the stories. The slab thickness is 10 inches at second story, 8.5 inches at 3 rd to 7thstories and 8 inches at the roof. An important observation that can be done about structural

system configuration concerns the disadvantageous orientation of column cross-sections of

lateral longitudinal frames, i.e. with minor axis perpendicular with plane of lateral frame.The foundation of the building consists of isolated footings interconnected by fundationbeams.

4.2.2. Instrumentation

The building is part of California Strong Motion Instrumentation Program (CSMIP Station 24386) operated by California Division of Mines and Geology (CDMG). It is located

approximately 4.5 miles east of the epicenter of Northridge earthquake. The damageexperienced during Northridge 1994 earthquake was heavy, building passing near collapse.

The building instrumentation consists of sixteen sensors, recording motion on E-W, N-S andUp-Down directions.

Acceleration spectra for Northridge 1994 is represented in figure 4.2.2.1 Vrancea 1977and Al Aqaba 1995 earthquake acceleration spectra, as-recorded and scaled with 0.68 and2.25 respectively, are represented as well. On the very same figure, the first three periods of

the structure in longitudinal direction are indicated using vertical bars. Such kind ofrepresentation allow a quick and synthetic comparison of efects induced by the three

considered earthquakes (Northridge 1994, Vrancea 1977 and Al Aqaba 1995) upon structure.


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0

2

4

6

8

10

12

14

16

0.0 0.5 1.0 1.5 2.0 2.5 3.0Period [sec]

SpectralAcc.

[m/s/s]

Northrid ge - ch.16

Vrancea sf = 1.00

Vrancea sf = 0.68

Al Aqaba sf = 1.00

Al Aqaba sf = 2.25

T 1T 2

T 3

Figure 4.2.2.1 Acceleration spectra for Northridge 1994, Vrancea 1977 and Al Aqaba

1995. T1, T2 and T3 first three periods of the building in longitudinal (E-W) direction.

4.2.3. Description of Structural Model

The building was analyzed in the E-W (longitudinal) direction. The building was

modeled as three-dimensional structure. The floor diaphragms were assumed to be infinitelyrigid. The vertical elements are assumed fixed at the ground level.

After several trials, the effective stiffnes of the other structural elements is as follows:0.45EI for columns and 0.30EI for beams. The damping level considered in modelling is 5

% from critical (Reileigh type damping). Actual material strength is 1.50 times design valuesfor steel and concrete. Total mass of structure considered in nonlinear time-history and

pushover analyses is about 4548.4 kNs2/m. The fundamental period of structural model in

longitudinal direction is practicly identical with the one obtained using seismic records, T1 =1.534 sec.

4.2.4. Nonlinear Time-history and Pushover Analysis Results

The comparison between recorded and calculated longitudinal displacement responseat roof for 1994 Northridge earthquake is presented in figure 4.2.4.1. Up to second 9.00,

structural model aproximates with sufficient exactity recorded structural rsponse. Subsequentdifferences appear after the 4th story columns fail in shear.


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-0.25

-0.2

-0.15

-0.1

-0.05

0

0.05

0.1

0.15

0.2

0.25

0 2 4 6 8 10 12 14

Time [sec]

RoofDisp

lacement[m]

RecordedCalculated

Figure 4.2.4.1 Comparison between calculated and recordedroof displacement time-history plot

In figure 4.2.4.2, deformed shape of structure and plastic hinges configuration at 12.00

seconds are presented. Although number of plastic hinges is important, no excessive plastic

rotation occurred (max / yield < 4). Considering the great number of hinges appeard duringanalysis, it is easy to figure out that energy dissipated through plastic deformations in eachplastic hinges is small, which lead to relatively small plastic rotations. The plastic hingeconfiguration is close to the one corresonding to the ideal plastic mechanism (plastic

hinges at beam ends and at the bottom end of first story columns). However, the actualconfiguration differs from the ideal one by plastic hinges appeared at the 4 th story (figure

4.2.4.2). Total number of hinges sorted by plastic rotation is displayed in table 4.2.4.1.Because of the difficulties in formulating constitutive laws for M-N plastic hinges that

take into account shear failure, certain quantitative errors present in the analysis might be

explained by the level of aproximation admited in describing plastic hinge behavior.Besides the important number of plastic hinges, Northridge 1994 earthquake nonlinear

time-history analysis shows that top ends of 4th

story columns of lateral frames (whichundertake almost entire lateral load) are, with little exceptions, plastified. This effect is due

strength capacity reducement as longitudinal reinforcement reduces from 6 #9 la 6 #7 at 3rd

floor.

Figure 4.2.4.2 Deformed shape and plastic hinges configuration for Northridgeearthquake at 12.00 seconds

In figure 4.2.4.3 base shear vs. top displacement for Northridge 1994 earthquake time-history and pushover analyses are presented. On the plot, absolute values for base shear force


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and top displacement are represented. The two pushover loading patterns are as it follows:

mode 1 loading pattern; pattern 2 lateral loads with distribution along height simmillar toabsolute maximum base shear force distribution. One can observe pushover curve

coresponding to pattern 1 enclose most of nonlinear time-history base shear force values withone exception a relatively narrow time interval centered on 8.40 second, in which superior

modes influence in structural response is more significant.

0

2000

4000

6000

8000

0 0.15 0.3 0.45

Roof disp. [m]

B

aseShear[kN]

Northridge

Pushover - pattern 1

Pushover - mod1

Figure 4.2.4.3 Absolute base shear force vs. roof displacement for Northridge 1994earthquake time-history analysis and pushover analyses considering two loading patterns

Pushover loading patterns used for 7 story building analysis (other loading patterns than the

one corresponding to mode 1) are indicated bellow:- Pattern 1 horizontal loading with distribution simmilar to absolute maximum base

shear force distribution for Northridge 1994 earthquake.- Pattern 2 horizontal loading with distribution simmilar to absolute maximum base

shear force distribution for Vrancea 1977 earthquake.- Pattern 4 horizontal loading with distribution simmilar to absolute maximum base

shear force distribution for Al Aqaba 1995 earthquake.

0.00 0.50 1.00 1.50 2.00 2.50

1

2

3

4

5

6

7

Story

DRIFT [%]

Northridge

0.0 1.0 2.0 3.0 4.0 5.0

1

2

3

4

5

6

7

Roof

Story

Abs .Acc .[m /s/s]

Northridge

(a) (b)

Figure 4.2.4.4 Drift and absolute acceleration envelope for Northridge 1994 earthquake


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Further on, some of the results obtained for time history analyses with Vrancea 1977

and Al Aqaba 1995 are presented. Vrancea 1977 and Al Aqaba 1995 earthquakeaccelerograms were gradually scaled, up to the level corresponding to first column failure

(max > 4pl in a plastic hinge), or to a local failure mechanism appearance.

-0.1

0

0.1

0.2

0.3

0.4

0.5

0 5 10 15 20

Time [sec]

Displacem

ent[m

]3rd story

Roof

-0.1

0

0.1

0.2

0.3

0.4

0.5

0 5 10 15 20

Time [sec]

Displacem

ent[m

]3rd story

Roof

(a) (b)

Figure 4.2.4.5 3rd floor and roof displacement time-history plots for

Vrancea 1977 unscaled (a) and scaled with 1.25 (b)

In figure 4.2.4.5 (a) and (b) it can be seen for Vrancea 1977 earthquake (as-recorded andscaled with 1.25) that after 6.80 second important plastic deformations appear.

Figure 4.2.4.6 Deformed shape and plastic hinges configuration for Vranceaearthquake at 6.86

Figure 4.2.4.7 Deformed shape and plastic hinges configuration for Vranceaearthquake scaled with 1.25, at 6.82 seconds


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As can be seen in figures 4.2.4.5, 4.2.4.6, and 4.2.4.7, for Vrancea 1977 earthquakethe structure is at its strength limit.

0

2000

4000

6000

8000

0 0.15 0.3 0.45

Roof d isp. [m]

BaseShear[kN]

Vrancea

Pushov er - pattern 2

Pushover - mod1

Figure 4.2.4.8 Absolute base shear force vs. roof displacement for Vranceaearthquake time-history and pushover analyses

Obs:

pattern mod 1 similar to mode 1 imposed displacements loading pattern;load pattern 2 simmilar to lateral load distribution at second 7.46 of Vrancea

earthquake time-history analysis (at sec. 7.46 base shear is maximum).

For Al Aqaba 1995 earthquake, structural system has linear-elastic behavior. For a

more eficient use of Al Aqaba 1995 earthquake in , the accelerogram will be scaled with

4.66. The value of scalling factor (4.66) was chosen so that maximum plastic rotation dontexceed 4pl.

0.00 1.50 3.00 4.50

1

2

3

4

5

6

7

Story

DRIFT [%]

Vranc ea f = 1.00

Vranc ea f = 1.25

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5

1

2

3

4

5

6

7

Roof

Story

Abs .Acc .[m/s/s]

Vrancea f = 1.25Vrancea f = 1.00

(a) (b)

Figure 4.2.4.9 Drift and absolute acceleration envelope for Vrancea earthquake


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-0.06

-0.04

-0.02

0

0.02

0.04

0.06

5 10 15 20 25 30

Time [sec]

Displacem

ent[m

]3rd story

Roof

-0.3

-0.2

-0.1

0

0.1

0.2

0.3

5 10 15 20 25 30

Time [sec]

Disp

lacem

ent[m

]3th story

Roof

(a) (b)

Figure 4.2.4.10 3rd story and roof displacement time-history plot for Al Aqaba

1995. As-recorded (a) and scaled with a factor of 4.66 (b)

Nonlinear time-history analysis for Al Aqaba 1995 earthquake scaled with 4.66 reveal the

appearance of important plastic deformations. As figures 4.2.4.10 (b) si 4.2.4.11 show, thesedeformations are mainly concentrated at stories 4 and 5.

Figure 4.2.4.11 Deformed shape and plastic hinges configuration for Al Aqaba

earthquake, scaled with 4.66, at 23.1 seconds

0

2000

4000

6000

8000

10000

0 0.15 0.3R o o f d i s p . [ m ]

B

aseShear[kN]

Al Aqaba f = 4.66

Pusho ver - pattern 4

Pushover - mod1

Figure 4.2.4.12 Absolute base shear force vs. roof displacement for Al Aqabaearthquake time-history and pushover analyses


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Similarities in response spectra of Northridge 1994 and Al Aqaba 1995 (scaled with 4.66)

can be seen in structural response as well maximum drift and maximum absoluteacceleration for the two earthquakes are quite similar.

0.00 1.00 2.00 3.00

1

2

3

4

5

6

7

Story

DRIFT [%]

Al Aqab a f = 1.00

Al Aqab a f = 4.66

0.0 1.5 3.0 4.5

1

2

3

4

5

6

7

Roof

Story

Abs .Acc .[m/s /s]

Al Aqaba f = 4.66Al Aqaba f = 1.00

Figure 4.2.4.17 Drift and absolute acceleration envelope for Al Aqaba earthquake

In table 4.2.4.1 column plastic rotation informations are centralized. Due to importantnumber of plastic hinges opened in structural elements during the three earthquake analyses

(and by that low energy dissipation requirement in plastic hinges), maximum plasticrotations in column hinges dont exceed 4pl.

Table 4.2.4.1 Total number of column plastic hinges

Number of column hingeswith max / Y betweenEarthquake

Acc. scalefactor

1..4 4..6 > 6

Whittier Narrows 1 73 0 0

Vrancea '77 1 184 24 23

Vrancea '77 0.68 163 24 0

Al Aqaba 1 0 0 0

Al Aqaba 2.25 87 0 0

For the other two structures simmilar type of calculations were performed. A few of them shall be

presented bellow.


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4.3. CASE STUDY II 10 Story RC Building

4.3.1 Building description

The construction Lutheran Whittier Tower Building, is located in Los Angelesmetropolitan area (City of Whittier, 7215 Bright Avenue) at latitude 30.98 and longitude118.04 W. The building was designed (according UBC 1970) and constructed in 1972. Aphotograph of the building and of structural model is shown in figure 4.3.1.1.

(a) (b)

Figure 4.3.1.1 General view of the Lutheran Whittier Tower (a);

3D view of structural model (b)

The structural system in longitudinal (north-south) direction comprises a moment-resisting frames designed to carry most of lateral loads in the exterior frames. Exterior frames

consist of 20 x 20 (508 x 508 mm) in columns and 24 x 24 inch (610 x 610 mm) beams. Theinterior frames consist of columns with squared 16 inch (406 mm) cross-section and a 6

inch (165 mm) thick cast-in-place flat plate. In transverse direction, the structural systemconsist in shear walls coupled with moment resisting frames. The shear walls are disposed at

the ends of the building axes 2 and 13, and in the vicinity of elevator shaft (axes 7 and 8).The frames in the remaining lines are very flexibile column flat plate, so that the most

important part of transverse lateral loading is undertaken by walls.The foundation of the building consists of spread footings. The bassement has in-plane

extents greater than the building and strong perimetral walls.

In-plane dimensions of the building are 1834 x 642 (55.92 x 19.57 m) for first 1-2story and 1834 x 528 (55.92 x 16.06 m) for the rest of the stories. The total height of the

building is 90 (27.51 m). The first story height is 12 (3.66 m), the height of the rest of the

stories is 88 (2.65 m).

4.3.2 Instrumentation

The building is part of National Strong-Motion Instrumentation Network (NSMIN)

operated by U. S. Geological Survey (USGS). The building instrumentation consists of threeSMA analog accelerographs, each capable to record three components of the motion. Thelocation of instruments can be seen in figure 4.3.2.1 (bassement, 5th floor and 10th floor).


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Figure 4.3.2.1 Building instrumentation scheme

In figure 4.3.2.2 the following acceleration response spectra are represented:

- Whittier Narrows acceleration response spectra;- Vrancea 1977 as-recorded and scaled with 0.68;

- Al Aqaba 1995 as-recorded and scaled with 2.25.Beside the five response spectra, the first three natural periods are marked on the plot by

vertical bars.

02

4

6

8

10

12

14

0.0 0.5 1.0 1.5 2.0 2.5 3.0Period [sec]

SpectralAcc.

[m/s/s]

Whitt.Narrows

Vrancea sf = 1.00

Vrancea sf = 0.68

Al Aqaba sf = 1.00

Al Aqaba sf = 2.25

T 1

T 2

T 3

Figure 4.3.2.2 Acceleration spectra for Northridge 1994, Vrancea 1977 and Al Aqaba1995. T1, T2 and T3 first three periods of the building in longitudinal (E-W)

direction.

4.3.3 Description of Structural Model

The building was analyzed in the N-S (longitudinal) direction. The building wasmodeled as three-dimensional structure. The floor diaphragms were assumed to be infinitelyrigid. The vertical elements are assumed fixed at the bassement level. In-plane translation ofcolumn ends at first story is restrained (due to bassement perimetral strong shear walls).

The width of slab interacting with interior columns was established according to ACI 318-08:C+3h, where C column width and h slab thickness. To account rapid bending stiffness

degradation of the slab during earthquake, the effective stiffness of slab strips is taken 0.3EI.Shear walls are modeled with linear-elastic frame elements. This assumption is reasonable


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because of the little influence exerted by walls working along minor inertia axis in N-S

directional response.After several trials, the effective stiffnes of the other structural elements is as follows:

0.75EI for columns and 0.6EI for beams. The bending stiffness of shear walls working inlongitudinal direction (along minor axis) is 0.6 EI. The damping level considered in

modelling is 5 % from critical (Reileigh type damping). Actual material strength is 1.35/1.75times design values for steel and concrete respectively. Total mass of structure considered for

time-history analyses is 8969.4 kNs2/m. The first three natural periods of structure inlongitudinal direction are in good accordance with the ones determined using seismic

response records: T1 = 1.430 sec, T2 = 0.485 sec si T3 = 0.286 sec.

4.3.4 Nonlinear Time-history and Pushover Analysis Results

The comparison between recorded and calculated longitudinal displacement response

at stories 5 and 10 for Whittier Narrows 1987 earthquake is presented in figures 4.3.4.1 and4.3.4.2. The corelation between calculated and recorded is good.

-0.05

-0.04

-0.03

-0.02

-0.01

0

0.01

0.02

0.03

0.04

0.05

0 2 4 6 8 10

Time [sec]

Displacement[m]

Recorded

Calculated

Figure 4.3.4.1 Comparison between calculated and recorded

5th story relative displacement time-history plot

-0.08

-0.06

-0.04-0.02

0

0.02

0.04

0.06

0.08

0 2 4 6 8 10

Time [sec]

Disp

lacement[m]

Recorded

Calculated

Figure 4.3.4.2 Comparison between calculated and recorded10th story relative displacement time-history plot


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In figure 4.3.4.3, deformed shape of structure and plastic hinges configuration at second

4.48 (the time instant characterized by maximum deformaations) are presented. Concurentlywith field observation, degradation level is small - total number of plastic hinges is small and

no excessive plastic rotation occurred (max / yield < 4). Yet, the wide majority of plastichinges are formed in upper levels columns(this fact could be explained by flexural capacity

reduction caused by axial stress reduction). The configuration of lateral frames determines astrong beam week column type behavior.

Figure 4.3.4.3 Deformed shape and plastic hinges configuration at sec 4.48

0

4000

8000

12000

16000

20000

0 0.1 0.2 0.3

10th story disp . [m]

BaseShear[kN]

Whittier Narrows


Figure 4.3.4.5 Absolute base shear force vs. roof displacement for Whittier Narrowsearthquake analyses and linear load pattern pushover analyses

Note in pushover analysis the influence of shear walls was not considered (their bending

and shear stiffnes was afected with very small coefficients).


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4.4. CASE STUDY III 13 Story RC Building

4.4.1. Building description

The building, is located in Los Angeles metropolitan area (District of Sherman Oaks,

Ventura Boulevard no.24322) at latitude 34.154 N and longitude 118.465 W. The buildingwas designed (according to Los Angeles Civil Building Code 1964) and constructed in1965. A photograph of the building and a 3D view of structural model is shown in figure

4.4.1.1. The building is a 15-story RC building, with two stories bellow ground level andthirteen above. The stories bellow ground are enclosed by structural walls. The building

orientation is E-W in longitudinal direction.The lateral load resisting system is a frame structure with two spans of 27 ft (10.97

m) and seven bays of 27 ft (8.23 m). The height of transversal beams is 50 in (1.27 m) at the

2nd floor (above ground) and 33 in (0.84 m) for the rest of stories. Longitudinal beams have

24 x 32 rectangular cross-section. The exterior columns (except corner columns which areL shaped) are rectangular in cross-section 24 x 36 in (0.61 x 0.91 m) and the interior ones

square, with 36 inches (0.91 m) in dimension. Transverse frames along lines F, G, H, J, K, Land E, M are identical to each other. The longitudinal frames 1A and 4 are identical as well(and symmetrical about 2A line). In longitudinal direction there are two secondary floor

beams for every width.

(a) (b)

Figure 4.4.1.1 General view of the building

The foundation of the building consists of concrete 34-ft-deep piles. The soil is recentalluvial deposit and is generally only moderately firm with soft layers at varying depth.


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4.4.2. Instrumentation

The building was part of instrumentation program conducted by California Division

of Mines and Geology (CDMG). The building instrumentation consists of 15 installedaccelerometers mounted at different locations (see figure 4.4.2.1).

Before instrumentation, the building had experienced 1971 San Fernando earthquakewhen little damage occurred to structural system. The corner columns at stories 1 and 2,

spandrel beams at 2nd story and basement walls suffered some damage (cracks and anchorageslippage).

During 1994 Northridge earthquake the structural and non-structural elementssustained moderate damage as well. Epicentral distance was about 9.7 Km. Al 15

accelerometers triggered during earthquake.

Figure 4.4.2.1 Building instrumentation scheme

0

5

10

15

20

0.0 0.5 1.0 1.5 2.0 2.5 3.0Period [sec]

S

pectralAcc.

[m/s/s]

NorthridgeVrancea s f = 1.00

Vrancea s f = 1.2Al Aqaba sf = 1.00

Al Aqaba sf = 5.00T 1 = 2.94 sec

T 2 = 0.98 secT 3 = 0.55 sec

Figure 4.4.2.4 Acceleration spectra for Northridge 1994, Vrancea 1977 and Al Aqaba1995. T1, T2 and T3 first three periods of the building in longitudinal (E-W) direction.


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4.4.3. Description of Structural Model

The building was analyzed in the N-S (transversal) direction. The building was

modeled as three-dimensional structure. The floor diaphragms were assumed to be infinitelyrigid. The vertical elements are assumed fixed at the ground level.

The width of slab interacting with interior columns was established according to ACI318-08: C+3h, where C column width and h slab thickness. To account rapid bending

stiffness degradation of the slab during earthquake, the effective stiffness of slab strips is

taken 0.3EI. Shear walls are modeled with linear-elastic frame elements. This assumption isreasonable because of the little influence exerted by walls working along minor inertia axis inN-S directional response.

After several trials, the effective stiffnes of the other structural elements is as follows:0.75EI for columns and 0.6EI for beams. The bending stiffness of shear walls working in

longitudinal direction (along minor axis) is 0.6 EI. The damping level considered inmodelling is 5 % from critical (Reileigh type damping). Actual material strength is 1.35/1.75

times design values for steel and concrete respectively. Total massof structure during seismic

action was considered to be 20946 kNs2/m. The first three periods of structural model in

transverse direction are in good accordance with the ones determined from seismic responserecords: T1 = 2.945 sec, T2 = 0.985 sec and T3 = 0.546 sec.

4.4.4. Nonlinear Time-history and Pushover Analysis Results

The comparison between recorded and calculated longitudinal displacement responseat 8th story and roof for Northridge 1994 earthquake are presented in figures 4.4.4.1 and4.4.4.2. As can be seen in the figures, the corelation between calculated and recorded

response is good.

-0.3

-0.2

-0.1

0

0.1

0.2

0.3

0 3 6 9 12 15

Time [sec]

RoofDisplacement[m]

Recorded

Caculated

Figure 4.4.4.1 Roof relative displacement time-history plot

In figure 4.4.4.3 deformed shape and plastic hinge configuration for marginal (a) andinterior (b) transversal frame at 11.60 seconds of Northridge 1994 earthquake. It can be

observed from figure 4.4.4.3 the column plastic hinges are formed at the upper stories whereaxial load level (and so, the flexural capacity) is low.


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-0.20

-0.15

-0.10

-0.05

0.00

0.05

0.10

0.15

0.20

0 3 6 9 12 15

Time [sec]

8thSto

ryDisp.

[m]

Recorded

Caculated

Figure 4.4.4.2 8th story relative displacement time-history plot

(a) (b)

Figure 4.4.4.3 Deformed shape and plastic hinges configuration for Northridge 1994

earthquake at sec 11.60 (a) lateral transverse frame (b) interior transverse frame

0

4000

8000

12000

16000

20000

24000

28000

0 0.1 0.2 0.3 0.4 0.5

Roof disp. [m]

Base

Shear[kN]

Northridge


Figure 4.4.4.4 Absolute base shear force vs. roof displacement for Northridge time-historyanalysis and mode 1 load pattern pushover analysis


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CHAPTER 5 Research Synthesis, Conclusions and

Recomandations

Thesis objective is to analyze the concordance between the results of high complexitystructural analysis methods (dynamic nonlinear and pushover analysis) and response recordscorroborated with field observations for reinforced concrete structures.

For the study, three multi-story reinforced concrete structures from suthern Californiawere considered. The considered structures were designed according to midd 60s and early70s design codes. For the analyzed structures response records to Northridge 1994earthquake and Whittier Narrows 1987 earthquake are available. The three structures are partof NSMIN - National Strong Motion Instrumentation Network and CSMIP (CaliforniaStrong Motion Instrumentation Program) network. Some field observations establishedimmediately after the above mentioned earthqakes are available too. The three reinforcedstructures are briefly described bellow.

1. 7 story reinforced concrete structure without basement, The structural system

comprises moment-resisting frames and a flat slab-column system. The momentresisting frame is designed to carry most of lateral loads. It is placed at the exterior of

the building. In-plane dimensions of the building are about 48.77 x 18.90 m. Totalheight of the building is 20 m. The buildings has three spans (6.10 m 6.70 m 6.10 m)

and eight bays (8 x 5.70 m). The first story height is 4.12 m, the height of the rest of thestories is 2.65 m.

2. 11-story RC building, with one story bellow ground level and ten above. The

structural system in the analyzed (longitudinal) direction comprises a lateral moment-resisting frame (designed to carry most of lateral loads) asociated with columns withsquared cross-section with a thick cast-in-place flat plate. In transverse direction, the

structural system consist in shear walls coupled with moment resisting frames. In-planedimensions of the building are 55.92 x 19.57 m for first 1-2 story and 55.92 x 16.06 m

for the rest of the stories. The total height of the building is 27.51 m. The first storyheight is 3.66 m, the height of the rest of the stories is 2.65 m.

3. 15-story RC building, with two stories bellow ground level and thirteen above. Thestories bellow ground are enclosed by structural walls. The lateral load resisting systemis a frame structure with two spans of 10.97 m and seven bays of 8.23 m. Typical storyheight is 3.58 m and for the first (above ground) story 7.16 m.

For the analyzed structures, static (pushover) and dynamic nonlinear (time-history)

analyses were conducted. For dynamic nonlinear analyses, the following input accelerogramwere considered: Northridge 1994 (for 7-story and 13-story RC buildings) and Whittier

Narrows 1987 (for 10-story building) as accelerograms of earthquakes experienced by thestructures and Vrancea 1977 and Al Aqaba 1995 to better describe the structural behavior and

performances of the three structures. Vrancea 1977 (Romania) and Al Aqaba 1995 (Jordania)were used as-recorded and scaled with factors that correspond to maximum plastic rotations

in column plastic hinges of about 4pl or to the appearnce of local failure mechanism in thestructure.

Static nonlinear anlyses were conducted considering imposed lateral displacementsproportional with fundamental mode deformed shape in the analyzed direction and


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considering horizontal forces distributed identically with maximum instantaneous lateral load

distribution obtained in nonlinear time-history analyses.

The results of nonlinear time-history analyses of the three structures for earthquakesthey had experienced are generaly in good accordance with their recorded response. The

position and magnitude of structural degradation (plastic hinges) broad into the light bynonlinear dynamic analyses are concurent with field observations.

However, for the 7-story structure certain quantitative errors in calculated responseare observed. These are mainly produced because of the level of simplification admited in

describing column plastic hinge behavior. The constitutive model does not consider thepossibility of shear failure for the columns. It has been observed during Northridge 1994earthquake shear failure at 4th story columns.

Nonlinear dynamic analyses conducted with Vrancea 1977 and Al Aqaba 1995 showsome design faults. For 13-story structure, the value of scale factor considered for Al Aqaba

1995 earthquake (5.00), correspond to a local failure mechanism formation at first story. Thisfact reveal an important disadvantage in structural design of the structure, because no matterthe type and intensity of a given earthquake, no local failure mechanisms are admitted.

CONCLUSIONS

The good accordance of calculated and recorded seismic response for the analyzedstructures show the advanced structural analysis methods (nonlinear static and dynamic) as

offering a high-fidelity image of the structural behavior during seismic action, especially forstructures that can be modeled as 3D frame structures. In this case, the fidellity level of

structural model is maximum, especially in post-elastic domain. The failure mechanisms canbe defined accurately and precise. This is much more difficult to achieve in the case of dual

or shear walls structures.

RECOMANDATIONS

1. For the constructions designed in seismic zone of Al Aqaba Jordania, nonlineartime-history analyses with local accelerograms must be performed. At the same time,amplified intensity local ground motions must be used in order to broad into lightpotential design faults like local failure mechanisms. Such failure mechanisms must beavoided for any intensity the earthquake might have. The overall colaps must take placeby maximum plastic rotations exeedance in plastic hinges from beam ends. Failure bysoft story mechanisms is totally restricted.

2. It is necessary to increase the number of seismic records and to create a groundmotion database as a base for seismic design in Syria. Adoptingseismic design codes

without take into account the local conditions may conduct to potential faults instructural design.


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References

1. Seismology Committee, Structural Engineers Association of California, "Recommended LateralForce Requirements and Commentary", San Francisco, California, 1999 Vol 1 - SEAOC.

2. Earthquake Engineering Research Institute, "Northridge Earthquake January 17, 1994", Oakland, California,

1994. Available from PEER: http://peer.berkeley.edu/

3. Computers & Structures, Inc., ETABS Integrated Building Analysis & Design, User Interface

Manual,January 20064. Naeim Farzad 1997 Performance of extensively instrumented buildings during the January 1997, Northridgeearthquake. Los Angeles: John A. Martin & Associates, Inc

5. ATC-33.03, "Guidelines for the Seismic Rehabilitation of Buildings (75% complete draft), Volume I:

Guidelines", Applied Technology Council, Redwood City, California, October 1995.6. International Conference of Building Officials (ICBO), Uniform Building Code, 1997.

7. P100-1_2006 Code for A Seismic Design of Residential Buildings, Agro-zoo-technical and Industrial

structures", Ministry of Public Works and Territory Planning.8. Southern Building Code Congress International (SBCCI), Standard Building Code, Birmingham, AL, 1993.

9. Building Seismic Safety Council, NEHRP Recommended Provisions for the Development of Seismic

Regulations for Buildings, FEMA-222,Washington, DC, 1994.

10. Applied Technology Council, Tentative Provisions for the Development of Seismic Regulations forBuildings, ATC 3-06,Washington, DC, 1978.

11. Earthquake Engineering Research Institute, Loma Prieta Earthquake Reconnaissance Report,

Earthquake Spectra, Supplement to Volume 6, May 1990.

12. Clough, R. W., and Benuska, K. L., "Earthquake Performance of High-Rise Buildings", Report to theFederal Housing Administration, T. Y. Lin and Associates, Van Nuys, CA, October 1965.

13. Jordan, R. M., "Evaluation of Strengthening Schemes for Reinforced Concrete Moment Resisting Frame

Structures Subjected to Seismic Loads", Ph.D. Dissertation, The University of Texas at Austin, Austin, TX,May 1991.

14. Eduardo Miranda 1987 Seismic Evaluation and Upgrading of Existing Buildings, Dissertation, University of

California at Berkeley15. Chopra, A. K., "Dynamics of Structures, A Primer", Monograph Series, Earthquake Engineering Research

Institute, Berkeley, CA, 1981.16. Clough, R.W., and Penzien, J., "Dynamics of Structures", McGraw-Hill Book Co., New York, NY, 1975.

17. Biot, M. A., "A Mechanical Analyzer for Prediction of Earthquake Stresses", Bulletin of the Seismological

Society of America, Vol. 31, pp.151-171, 1941.

18. Biot, M. A., "Analytical and Experimental Methods in Engineering Seismology", Proceedings ofASCE, Vol. 68, pp.49-69, 1942.

19. Housner, G. W., "An Investigation of the Effects of Earthquakes on Buildings", Ph.D. Thesis,California Institute of Technology, Pasadena, CA, 1941.

20. Miranda, E., Bertero, V., Evaluation of Seismic Performance of a Ten-Story RC Building During the Whittier

Narrows Earthquake, Report EERC 91-10, October 1991, University of California at Berkeley.

21. US Nuclear Regulatory Commission, "Combination of Modes and Spatial Components in Seismic

Response Analysis", Regulatory Guide 1.92, Washington, DC, 1976.22. Newmark, N. M. and Hall, W. M., "Earthquake Spectra and Design", Monograph Series, Earthquake

Engineering Research Institute, Berkeley, CA, 1982.

23. Saiidi, M. and Sozen, M. A., "Simple and Complex Models for Nonlinear Seismic Response of Reinforced

Concrete Structures", Structural Research Series No. 465, University of Illinois, Urbana, IL, August 1979.24. Kaazem Moslem and Mihailo D. Trifunac EFFECTS OF SOIL STRUCTURE INTERACTION ON THE

RESPONSE OF BUILDINGS DURING THE STRONG EARTHQUAKE GROUND MOTIONDEPARTMENT OF CIVIL ENGINEERING Report No. 86-0 March 198625. Abdel-Ghaffar, A. M., "Engineering Data and Analyses of the Whittier, California, Earthquake of January

1, 1976", Report No. EERL 77-05, Nov. 1977, Caltech, Pasadena, California.

26. Earthquake Engineering Research Institute. Available from PEER: http://peer.berkeley.edu/

27. Applied Technology Council, "Post-Earthquake Safety Evaluation of Buildings", ATC-20, Redwood City,

CA, 1989.28. Vision 2000 Committee, "VISION 2000, Performance Based Seismic Engineering of Buildings", Structural

Engineers Association of California (SEAOC), Sacramento, CA, April 1995.

29. Evaluation of Seismic Capacity and Retrofitting of a Typical 12 Stories RC Frame Structure in Bucharest,


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Romanian by DR. Mihai PAVEL* (For the course of Earthquake Engineering 2003 2004)

30. Sasani, M., Bertero, V., Anderson, J., Rehabilitation of Nonductile RC Frame Building Using Encasement

Plates and Energy Dissipating Devices, Report PEER 12/1992, University at California at Berkeley

31. INCERC seismic station ACCELEROGRAMS RECORDEDIN THE INCERC NETWORK DURING

THE VRANCEA EARTHQUAKES OF 1977 (Mw=7.5, h=109km)http://www.incerc2004.ro/accelerograms.htm

32. Vlado Gicev and Mihailo D. Trifunac Non-linesr Earthquake Waves in Seven-story Reinforced Concrete

HotelReport CE 06-03 November, 2006 Los Angeles, California.

33. William T. Holmes, Helen M. Ferner, Margaret Longstreth Buildings Case Studies Project1994Northridge Earthquake Proposition 122: Product 3.2 Edited by Rutherford & Chekene Seismic Safety

Commission 1900 K Street, Suite 100 Sacramento, California 95814 SSC 94-06

34. M.D. Trifunac and T.Y. Hao 7-Story Reinforced Concrete Building in Van Nuys, Photographs of the

Damage from 1994 Northridge Earthquake Report CE 01-05 July, 2001 Los Angeles, California.http://www.usc.edu/dept/civil_eng/Earthquake_eng/

35. Wilson E., "Three-Dimensional Static and Dynamic Analysis of Structures A Physical Approach with

Emphasis on Earthquake Engineering" - Computers and Structures Inc., Berkeley, California, USA, ThirdEdition - January 2002.

36. Li, Y.R. (1996), Non-Linear Time History And Pushover Analyses for Seismic Design and Evaluation,

Ph.D. Dissertation University of Texas, Austin, TX.

37. Ukaji, Kenichi Analysis of soil-foundation-structure interaction during earthquakesBLUME-018, TheJohn A. Blume Earthquake Engineering Center Stanford, California, 1975-03, 195 pages (535U3941975)

38. Melbourne F. Giberson THE RESPONSE OF NONLINEAR MULTI-STORY STRUCTURESSUBJECTED TO EARTHQUAKE EXCITATION A report on research conducted under a grant from the

National Science Foundation Pasadena, California 1967

39. Toshikazu Takeda, Mete A. Sozen and N. Norby Nielsen, Reinforced Concrete Response to SimulatedSrthquakesMembers, ASCE STRUCTURAL DIVISION Proceedings of the American Society of Civil

Engineers

40. Building Seismic Safety Council, NEHRP FEMA_222 FEMA _273 GUIDELINES FOR THE SEISMICREHABILITATION OF BUILDINGS

41. Russel Martio Nonlinear Push-Over Analysis of Reinforced Concrete StructuresFAINAL REPORT

Colorado Advanced Software Institute.

42. Kent, D. C., and Park, R., "Flexural Members with Confined Concrete", Journal of the Structural Division,ASCE, Vol. 97,(7), 1971, pp. 1969-1989.

43. Response-2000 Load-Deformation Response of Reinforced Concrete Sections by Evan C. Bentz and

Michael P. Collins. Available from http://www.ecf.utoronto.ca/~bentz/r2k.htm44. ATC40 Seismic evaluation and retrofit of concrete buildings Volume 1 ATC Applied Technology CouncilCALIFORNIA SEISMIC SAFETY COMMISSION Proposition 122 Seismic Retrofit Practices

Improvement Program Report SSC 96-01

45. James A. Mahaney, Terrence F. Paret, Brian E. Kehoe, Sigmund A. Freeman THE CAPACITYSPECTRUM METHOD FOR EVALUATING STRUCTURAL RESPONSE DURING THE LOMA

PRIETA EARTHQUAKE NATIONAL EARTHQUAKE CONFERENCE 1993.

46. SeismoSignal [2003] "Computer program for signal processing of strong motion data" [online]. Available

from URL: http://www.seismosoft.com.47. SIMQKE-II Erik H. Vanmarcke Gordon A. Fenton Ernesto Heredia-Zavoni Conditioned Earthquake

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All rights reserved


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CURRICULUM VITAE (Biography)

BASSAM BLAL was born in Swaida City , and originally from Tartous

City . Syrian Arab Republic on February 20,1962, the son of Mahmoud and

Manar. After completion of his study in Tartous Sheikh Saad High School

in 1979.

He was admitted to The Technical University of Civil Engineering

Bucharest Romania.

He was graduated the Degree of Master of Science in Civil Engineering with

Honour in 1985.

Latest projects designed

1. KIDZANIA DUBAI MALL( Kids city 14 Buildings)

2. 12 Cinemas DUBAI MALL (REEL CINEMAS)

This Dissertation was typed by the Author.

Tall Bldg Design

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