Steel Against Earthquake

8/8/2019 Steel Against Earthquake

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1 INTRODUCTIONFrom the seismic point of view, Europe can be con-sidered as subdivided into two main parts. Thenorthern part is geologically stable and the southern

part around the Mediterranean basin is generallyearthquake prone with different intensities from oneside to another.

The northern Countries use steel and composite

structures more than the southern Countries mainlyfor economical reasons, but also for a more consoli-dated tradition. They design seismic resistant steeland composite structures just for exportation to theearthquake prone Countries of the third World.

Contrary, the southern Countries are less richthan the northern ones and they traditionally use

both masonry and reinforced concrete as the mostcommon constructional materials.

The lesson learned in the last decades from themore severe earthquakes in south Europe (Friuli

Italy, 1976; Bucarest Romania, 1977; Campaniaand Basilicata Italy, 1980; Banat Romania, 1991;Erzincan Turkey, 1992; Dinar Turkey, 1995;Umbria Italy, 1997; Adana Turkey, 1998; Izmitand Duzce Turkey, 1999; Athens Greece, 1999)have shown that the behaviour of reinforced con-crete structures, which are the majority there, has

been generally very poor for many unquestionablereasons, i.e. non-seismically designed structures orformally seismically designed but characterised bywrong detailing and bad execution with poor materi-als or seismically designed but with load lower than

the actual one. The collapse of r.c. buildings is veryoften a tragic reality in the Mediterranean Countries(Figure 1). On the contrary, the performance of steeland composite structures, even if they are not verynumerous, has been satisfactorily good everywhere,

with very limited number of damage and total ab-sence of collapsed buildings.

Figure 1. The global collapse of a r.c. building.

This evidence is going to produce a slow, but con-tinuous, increase of the use of steelwork in seismicresistant structures also in south Europe (Mazzolani& Gioncu, 2000).

In fact, steel structures have been always consid-ered as a suitable solution for constructions in highseismicity areas, due to the very good strength andductility exhibited by the structural material, thehigh quality assurance guaranteed by the industrial

production of steel shapes and plates and the reli-ability of connections built up both in workshop andin field (Mazzolani & Piluso, 1996). In spite of thesenatural advantages, researchers are concerned aboutthe necessity that, in order to ensure ductile struc-tural behaviour, special care must be paid mainly inconceiving dissipative zones, which have to be

properly detailed, assuring stable hysteresis loops,

able to dissipate the earthquake input energy withhigh efficiency (Bruneau et al., 1998; Gioncu &Mazzolani, 2002). As a confirmation, during the re-cent seismic events of Northridge (Los Angeles, 17January 1994) and Hyogoken-Nanbu (Kobe, 17

Steel against earthquake

F.M. MazzolaniUniversity of Naples Federico II, Naples, Italy

ABSTRACT: The use of steelwork for seismic protection purposes is now-a-day a reliable solution whose ef-fectiveness is proved by the evidence of the catastrophic earthquakes in the World. This aspect is widely illus-trated in this paper, considering different aspects, from the development of the design methodologies to thecurrent codification, form the structural typologies to the damage analysis and consequent lessons. Some sig-nificant examples of seismic resistant buildings, erected in the European earthquake prone Countries during

the last decades, are also illustrated.


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January 1995), even if the cases of collapse of steel buildings have been extremely rare, steel momentframe buildings, considered as highly ductile sys-tems, exhibited an unexpected fragile behaviour.They presented many failures located at the beam-to-column connections, challenging the assumptionof high ductility and demonstrating that the knowl-edge on steel moment frames was not yet complete

(Mazzolani, 2000). Hence, in order to improve con-structional details and to propose new design solu-tions for achieving a correspondence between thedesign requirements and the actual structural re-sponse, the scientific community began to deepenthe reasons of this poor behaviour: does it dependson the material quality, on the design concept, on thestructural scheme, on the constructional details, onthe code provision, or on the seismic input occurred(Mazzolani, 1998)? Most of these questions are still

being analysed, but much more has been understoodon the seismic behaviour of steel structures. Conse-

quently, during the last years most of the recentknowledge has been already or is going to be intro-duced into the structural design provisions for seis-mic resistant design in all the earthquake proneCountries, giving rise to a new generation of seismiccodes (Gioncu & Mazzolani, 2002).

2 SEISMIC DESIGN METHODOLOGIESThe earthquake-resistant design of steel structures is

based on the same principles which all civil engi-neering constructions in seismic zones are basedupon.

The traditional philosophy (Mazzolani, 1998) ba-sically identifies three performance levels, whichshould be achieved for increasing the intensity ofearthquake actions: structural and non structuraldamage in frequent minor ground motions has to be

prevented; the minimisation of both structural andnon structural damage during occasional moderateseismic events has to be attained; finally collapse orserious damage in rare major earthquakes has to beavoided (Mazzolani and Piluso, 1994, and Maz-

zolani et al. 1995a). In some codes, the two first lev-els are concentrated in one, corresponding to a dam-age limitation level, which can be calledserviceability limit state.

In this context, seismic resistant dissipative struc-tures are usually designed to withstand severe earth-quakes by asking for a proper combination ofstrength and energy dissipation capacity. This designgoal is pursued in current seismic codes by means of

provisions aiming at assuring a minimum level ofstrength and by means of design and detail rules for

obtaining the required energy dissipation capacity(Mazzolani, 1991a). The minimum required level ofstrength is strictly dependent on the required energydissipation capacity and, for buildings, it is com-monly expressed by a coefficient, namely q-factor in

Eurocode 8 or reduction factor R in the Americancodes UBC and AISC, which reduces the elastic de-sign spectrum. Different values of the reduction fac-tor are given by the seismic codes for each structuraltypology (EC8, 2004, UBC 1997, and AISC 1997).

At the same time a minimum level of lateral stiff-ness is required in order to reduce the discomfortoccurring during frequent minor earthquakes.

It can be pointed out that the traditional philosophyleads to a great simplification in the assessment ofthe design method, because it allows to perform in asingle shot both the check against the serviceabilitylimit state and the collapse limit state. But as it oftenhappens, simplifications induce some limitations. Infact, certain structural typologies, dimensioned withcurrent design procedures, present lesser amount ofenergy dissipation capacity than that required in or-der to prevent the collapse under the most severe de-sign earthquake (Mazzolani and Piluso, 1995). Fur-thermore even in modern seismic codes the need to

check the actual seismic inelastic behaviour of thestructure is not specified, because it requires verycumbersome calculations. But the necessity to set-upgeneral provisions, which are always on the safeside, is unquestionable even if not yet completelyfulfilled in the codes, specially for steel structures.

According to a more recent philosophy, a com-prehensive design procedure should correlate the re-sistance of a structure at various limit states to the

probability that the earthquake action can reach theintensity required to induce the corresponding fail-

ure modes (Mazzolani, 1998). In this way, theminimisation of the cost/benefit ratio, which takesinto account the construction cost and the expectedlosses, can be attained for all the limit states occur-ring during the service life of the structure (Maz-zolani and Piluso, 1996). The complete achievementof the design objectives needs the use of multi-leveldesign criteria, such as the so-called performance

based earthquake resistant design. It can be statedthat also the traditional design approach qualita-tively and partially agrees with this concept (Berteroet al. 1996, and Bertero,1996). A more detailed defi-nition of the performance based earthquake resistantdesign has been supplied within the activities ofSEAOC Vision 2000 Committee (SEAOC Vision2000, 1995).

The aim of this new approach is to provide thedesigners with the criteria for selecting the appropri-ate structural system and its layout and for propor-tioning and detailing both structural and non struc-tural components, so that for specified levels ofearthquake intensity the structural damage will beconstrained within given limits. The coupling of a

performance level with a specific level of groundmotion provides a performance design objective. Inthis perspective, four performance levels have been

proposed: fully operational, operational, life safeand near collapse. The correspondent requirements


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for different civil engineering facilities and construc-tional materials are also suggested. In addition, fourearthquake design levels are specified:frequent, oc-casional, rare and very rare, which are characterised

by a return period equal to 43, 72, 475 and 970years, respectively.

According to this framework, it can be recognisedthat the complete knowledge of the seismic perform-

ances of a structure should need sophisticated nu-merical procedure, because the quantitative evalua-tion of the structural damage requires non-lineardynamic analyses. As such analyses cannot be com-

pulsory requested in common design practice, simplified procedures have to be suggested, leading tothe evaluation of the seismic inelastic performanceat different design criteria.

In this perspective a simplified method based on atrilinear modelling of the load-displacement curvewhich requires only elastic and rigid-plastic analyseshas been proposed by Mazzolani and Piluso (1997),

as a useful tool to check the design assumptions. Ithas been also used to predict the value of the peakground acceleration corresponding to the attainmentof the above performance levels. The approximateresults have been successfully checked by means ofdynamic inelastic analyses, leading to the evaluationof the ductility demand of multi-story MR steelframes (Figure 2).

Figure 2. Results of a simplified method based on a trilinearmodelling of the equivalent SDOF system (a) and its validationby means of inelastic dynamic analyses of global (b) MR steelmulti-storey frames.

3 THE EUROPEAN SEISMIC CODIFICATIONCHRONICLE

The principles of design for the use of steel in seis-mic resistant structures represent a quite new sub-

ject, because steel is not the most popular structuralmaterial in European Countries (Mazzolani, 1999a).

The lack of direct experience has been compensatedby several research activities developed all over theWorld in the last decade (as reported in STESSA94, STESSA 97, STESSA 2000, STESSA 2003and the next STESSA 2006 Conferences).

In this direction, the Committee TC13 on Seis-mic Design of the European Convention for Con-structional Steelworks (ECCS) played a leader role

of coordination, harmonisation and codification(Mazzolani, 1988). In 1988 the ECCS Committeepublished its first proposal of codification, the Euro-pean Recommendations for Steel Structures in Seis-mic Zones (ECCS, 1988). This important issuerepresented in Europe the first international meeting

point between the well consolidated "steel culture"and the more recent "seismic culture" (Mazzolani,1991b). In the special climate characterised by theearly developments of the Eurocodes during theeighties, the ECCS-Committee TC13 devoted themain part of its activity as a consulting body of the

drafting panel of EC8. The pressing deadlines for theissue of EC8 forced the ECCS-Committee TC13 toconcentrate in a very short time the preparation of itsRecommendations, which were immediately incor-

porated as the "Steel section", since the first editionof the Eurocode 8 (1988), which is very similar toChapter 6 of the last edition (2004).

For these reasons, this first edition of ECCS Rec-ommendations has to be considered as a first honestattempt to clarify the right way to proceed, in viewthat the results of the current research activity should

help to fill some of the existing gaps (Mazzolani,1992).

After many years, due to the big developments inthe field of seismic resistant steel structures on in-ternational level, together with the lessons learnedfrom the last terrible earthquakes, the scenario iscompletely changed and new areas are not yet ex-

plored (Mazzolani, 1995b).An important issue of ECCS-Committee TC13

has been the ECCS Manual on Design of SteelStructures in Seismic Zones (Mazzolani and Piluso,1994). It intends to provide designers with the basic

principles which the ECCS provisions, and thereforethe EC8 ones, are based upon and, at the same time,to give some new results, which must be used in theimprovement of the present codification.

4 THE BASIC STRUCTURAL TYPOLOGIESThe main typologies of seismic-resistant dissipativestructures can be classified according to the type ofdissipative elements. Three different typologies can

be recognised (Mazzolani and Piluso, 1994):

Moment resisting frames (MRF);Concentrically braced frames (CBF);Eccentrically braced frames (EBF).

Moment resisting frames are rectilinear assem-blages of beams and columns, with the beams rigidly

a)

b)


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connected to the columns. Their primary source oflateral stiffness and strength are the bending rigidityand strength of the frame members. Dissipativezones form in a large number and they are concen-trated in discrete regions at the ends of the members,the so-called plastic hinges, which dissipate energythrough a quite stable bending cyclic behaviour. Inorder to maximise the energy dissipation capacity,

plastic hinges have to develop in beams and at thecolumn base. The corresponding failure mode iscalled global collapse mechanism (Figure 3).

Figure 3. Global-type collapse mechanism.

Moreover moment frames are also preferred fortheir architectural versatility: there are no bracingelements which block wall openings and the maxi-mum flexibility for space utilisation is provided.This advantage is accompanied by a poor lateralstiffness of the whole structure, so that the membersizes are larger than those required for strength, dueto the necessity to contain sway deflections within

the drift limits, imposed by the codes (Mazzolaniand Piluso, 1996, and Bruneau et al. 1998).Concentrically braced frames resist lateral loads

primarily by developing high axial forces in diago-nal members. Only in some cases, the bending ac-tions in moment resisting connections (when appro-

priate) can resist a small percentage of lateral loads.In general, the dissipative zones are represented bythe tensile diagonals, because of the assumptionusually made that the compression ones buckle. Theinelastic cyclic performance of concentric braces isaffected by an energy dissipation capacity degrada-

tion, because of the repeated buckling of diagonal bars. For this reasons the q reduction factor is as-sumed to be lower than the one corresponding toMRF. Besides, high elastic stiffness is achieved bythe presence of diagonal bracing members (Maz-zolani and Piluso, 1996, and Bruneau et al. 1998).The most typical configurations are shown in Figure4.

Eccentrically braced frames are hybrid lateralforce-resisting systems respect to the above men-tioned ones. In fact they combine the individual ad-

vantages of moment-resisting frames and concentri-cally braced frames, assuring a high elastic stiffness,together with stable inelastic response under cycliclateral loading and good ductility and energy dissi-

pation capacity. They are characterised by diagonals

eccentrically located in moment-resisting frames, producing a stiffening effect. In such a way the beams are divided in two or three parts, being theshortest one the dissipative element of the eccentri-cally braced frame, called link, which dissipatesthe earthquake input energy by means of the inelas-tic cyclic shearing and bending. The assumed valueof the q-factor in this case is almost the same than

the one corresponding to MRF, depending on the lo-cation of the diagonal members (Mazzolani andPiluso, 1994, and Bruneau et al. 1998). The commontypes of EBF are classified as in Figure 5.

Figure 4. Common configurations of concentrically bracedframes (CBF)

Figure 5. Typical configurations of eccentrically braced frames(EBF)

In order to provide lateral stiffness comparable tothat of CBF and to offer high energy dissipation ca-

pacity during large inelastic deformations, shortlinks, having the flanges connected to the columns

by means of complete penetration welded joints, areto be preferred. This is because experimental results

put in evidence that the links in shear possess agreater rotation capacity than the ones in bending(AISC, 1997).

A comparison between MRF and braced frames(CBF and EBF) can be qualitatively given on the

base of the requirements which a seismic resistantstructure has to satisfy: strength and stiffness againstmoderate ground motions with a small return period;

strength, ductility and energy dissipation capacityagainst severe earthquakes with a great return pe-riod. It can be synthesised as in Table 1 (Mazzolani,1995a).


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Table 1. Comparison between MRF and braced frames.

Strength Stiffness Ductility

MRF good poor good

CBF good good poor

EBF good good good

5 IMPROVED STRUCTURAL TYPOLOGIESA new family of structural typologies dissipatingenergy in bending could be conceived starting fromtwo bays CBF, which in practice is the most eco-nomic system (Mazzolani et al., 1995b). Simplyvarying the distance b between the central col-umns B and C of Figure 6, structures with differentvalues of stiffness and ductility can be obtained,ranging from CBF to MRF. As far as b increases,stiffness decreases, but ductility and energy dissipa-tion increase, due to the presence of a central weak

beams, where two plastic hinges form at the ends.In this way a rational improvement of the poorductility of a CBF can be achieved not so much byductilising the system itself, but conceiving a systemcomposed by two or more rigid CBF subsystems,connected by means of weak beams (Mazzolani etal., 1995b and 1995c).

The structural solutions ranging between the CBFand the MRF are called dual structures, becausehorizontal loading are resisted in part by moment re-sisting frames and in part by bracing systems actingin the same plane.

A similar evolution can be achieved starting froma classical one bay CBF system, with e=0, which istransformed in a MRF passing trough EBF systems,characterised by different dimensions of the links,from short to long (Figure 7).

Figure 6. New typologies of frames: from a two bays CBF to asingle bay MRF.

Figure 7. New typologies of frames: from a single bay CBF toa single bay MRF.

In conclusion, the generation of such structuralsystems offers the possibility to obtain a wide range

of structures, laying between a very rigid CBF up toa very ductile MRF, which can be characterised by agiven combination of stiffness and ductility, for thesame strength requirement.

6 ANALYSIS OF DAMAGEIt has been reported (Bertero et al., 1994) that, dur-ing the above-mentioned seismic events (see Section

1), many pathological failures occurred in steel buildings: structural designers expected that steelelements in building frames would be fully yield andfinally collapsed in a ductile manner, after absorbing

part of the energy input of strong ground motions;contrary, it was particularly astonishing to observetypical low-energy failures.

The damage can be characterised at differentstructural levels: material, cross-sections, membersand connections (Mazzolani, 1999b).

As it is well known, steel is a very ductile mate-

rial, but a big loss of ductility has been evidenced byfractures formed in the member far from the weldedconnections, crossing the entire sections, as it isshown in Figures 8 a,b. This happened in the col-umns of the Ashiyahama apartment building in Kobeduring the Hyogoken Nanbu earthquake in 1995,

but it must be pointed out that the material has beensubmitted to extremely severe loading conditions,due to very strong vertical quakes with very high ve-locity of propagation in presence of low temperature(it was 6 oclock a.m. in winter).

Figure 8. Material brittle fracture during the 1995 earthquakein Kobe: a) building; b) bridge.

At the cross-section level, a loss of ductility canoccur due to local buckling phenomena. This hap-

pened in some bridge piles during the Kobe earth-quake, under form of panel buckling (Figure 9a) andof elephant foot buckled shape, typical for circularhollow sections (Figures 10a,b,c). In both cases thevery high vertical components seems to be the mainresponsible.

The local buckling in the column of the PinoSoarez building during the Mexico City earthquake

(1985) is mainly due to the bad execution of weld-

ings,

a) b)


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Figure 9. Cross-section failures.

Figure 10. The elephant foot pattern.

which form the square hollow section (see Figure9b).

Member ductility is conditioned by global buck-ling phenomena as it can be seen in Figure 11, wherethe overall buckling of bracing members producedlarge permanent deformations, leading to a changein the hysteresis loops and a consequent reduction inenergy absorption capacity.

Figure 11. Member failures.

But in general damage is mainly located in theconnections. Typical are the failure of column base

plates produced by the rupture of anchor bolts or thefracture of members at the bolted connections due tothe presence of holes which generate a section withreduced resistance (Figure 12a). The fracture ofwelded beam-to-column connections in moment re-sisting frames is undoubtedly the most widespreadtype of failure occurred in steel structures during

both Northridge and Hyogoken-Nanbu earthquakes(Figure 12b). However, it has been observed (Aki-yama and Yamada, 1995) that the fracture mecha-nisms were different between the two earthquakes,mainly due to the fact that different trends regarding

the detailing of beam-to-column connections are de-veloped in USA and Japan.

In fact according to the current U.S. practice forseismic resistant frames, both beams and columnsare H-shaped (Figure 13a). The beam web is field

Figure 12. Connection failures

bolted to a single plate shear tab which is shopwelded to the column. The beam flanges are fieldwelded to the column using complete penetrationwelds. Web copes are required to accommodate the

backup plate at the top flange and to permit makingthe bevel weld at the bottom flange.

The failure modes of this connections typologyduring Northridge earthquake mainly consisted inthe fracture of the welds connecting the beam

flanges to the column flanges and, in some cases, inthe propagation of this fracture within the column.

Contrary, the current Japanese practice uses H-shaped beams which are field welded to box-columns (Figure 13b). The stresses of the beamflanges are safely transmitted by means of continuity

plates welded through the box-columns. The trans-mission of the beam web stresses involves the out-of-plane bending resistance of the column wall

plates and, therefore, is not perfect. This causes highstress concentration at the ends of beam flanges,

which is further promoted by the web copes intro-duced for an easier welding (Mazzolani, 1998).

Figure 13. Beam- to column connections for MR steel frames,according to the current U.S. (a) and Japanese (b) practice.

7 LESSONS TO BE LEARNEDThe recent investigations have allowed to identifysome causes of damage in steel buildings.

On one hand, it can be ascribed to the use of fieldwelding so that poor workmanship is solely to blameand, therefore, it is necessary to tighten the site su-

pervision and to improve the welding details andprocedures; on the other hand, damage causes can beattributed to an excess of seismic loading and to de-fective design guidance leading to a rotation ductil-

ity supply in the connections lower than the earth-quake imposed demand (Elnashai, 1994).

Some other factors influencing fracture modalityare related to material properties, temperature, strainrate, joint geometry, plate thickness and so forth.

a) b)

a) b) c)

a) b)

a) b)


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More extensive considerations can be done (Maz-zolani, 1995c):- at the material level, damage can be imputed to

the concurrence of high velocity of load applica-tion and very low temperatures. These conditionsare not reproducible by means of the usual labo-ratory tests and until now they have been ignored

by the codes.

- local buckling of cross-sections in Kobe can be at-tributed to the important vertical component ofthe ground motions, which normally are underes-timate or sometimes neglected by seismic codes.

- a distinction must be done between far-fieldearthquakes which are characterised by the cyclicalternation of action and near-field earthquakewhere the impulsive feature of the ground motionis predominant, as it occurred in Kobe.It has also to be remembered that, regarding the

strength requirements, the American code UBC 91provides for special moments resisting frames a re-

duction factorRw equal to 12, which is equivalent toa value of the European q-factor equal to 8. On thecontrary, Eurocode 8 provides a q-factor value equalto 6 and the Japanese code a structural coefficientDsequal to 0.25 which corresponds to q = 4. Therefore,concerning the damage experienced by steel build-ings during the Northridge earthquake, it could beattributed to an excess of local ductility demand dueto the high earthquake intensity compared with theadopted design level.

Furthermore, in U.S. practice the beam flange is

welded to the column flange by provisionally usingbolted connection in the beam web during the erec-tion. In such a connection, the bending moment atthe beam web is hardly transmitted to the column,resulting in the stress concentration in flanges at theend of beams. This normally occurs in a beam-to-column connection because, while the stress trans-mission between the beam flange and the column iscompletely made through the diaphragm plate, the

bending moment in the web of the beam cannot betransmitted completely to the column, since thestress transmission is made through the out-of-plane

bending of the column flange. The diaphragm plateis usually thicker and wider than the beam flangeand, therefore, the fracture develops on the side ofthe beam (Mazzolani, 1998).

These effects were increased by the fact that theweak-beam type is preferred both in Japan and USAas a yield-mechanism. In addition the use of com-

pact beam sections, with small width to thickness ra-tio, in order to avoid local buckling, gives rise tostress concentration in the beam flanges and brittlemode of failure in the connection. Besides, defects

in material of the heavy column must be an incentivefor the propagation of brittle cracks on the side ofthe column.

Both in USA and in Japan extensive programs ofexperimental tests has been carried out on beam-to-

column sub-assemblages, in bi-dimensional and tri-dimensional frames.

It has been evidenced (Bertero et al., 1994) thatthe types of failure occurred in welded beam-to-column connections during the earthquakes have

been already observed in laboratory experiments. Numerical analyses of the seismic response of

steel framed buildings damaged during Northridge

earthquake have pointed out that there were severalground motions, recorded during this earthquake,able to significantly lead the structure into the ine-lastic range. In many cases, the plastic rotation de-mand at the beam ends exceeded 0.02 rad; therefore,on the base of the available experimental data andcodes provisions (UBC 91 and AIJ 90), it is clearthat the cracking occurring in the connections cannot

be considered unusual.In the experimental tests carried out in USA, a

plastic rotation supply equal to 0.02 rad has beenused as a benchmark to judge the seismic perform-

ance of beam-to-column connections, because it was believed to be sufficient to withstand severe earth-quakes. As the recent experience demonstrated thatthis limit value can be exceeded, it is clear that theattention should be focuses on the design value ofthe q-factor which could be reduced in order to limit

plastic rotation demands occurring during severeearthquakes or, as an alternative, on the improve-ment of the seismic performances of dissipativezones.

The SAC Steel Project has been developed to de-

rive new design procedures accounting for the les-sons learned from the Northridge earthquake. In par-ticular, the structural design philosophy, as well asthe characteristics of welding and of the structuraldetails, the velocity of load application and the in-fluence of the earthquake vertical component are themain issues under investigation within the SAC Pro-

ject (STESSA 2000).Also in Europe, from the observation of damage

in steel structure connections after the earthquakesof Northridge and Kobe it seemed that there is anurgent need to investigate new topics such as the in-fluence of the strain rate on the cyclic behaviour of

beam-to-column joints and more in general to re-view the whole background of modern seismic codesin order to grasp the design rules which failed (Maz-zolani, 1998).

The following question can raise: Can the re-sults of the American and Japanese on field ex-

perience be applied to European practice ? .It has to be considered that the steel grade, the

chemical composition and the mechanical character-istics of the steel can be different. Also the welding

technique can be different. In addition, differentstrength requirements and different levels of seismicinput lead to different plastic rotation demands.

In this context, a group of 8 European Countries(Italy, Romania, Greece, Portugal, France, Belgium,


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Bulgaria, Slovenia) developed a joint research project (INCO-COPERNICUS), co-called RECOS(Reliability of Connections of Steel Frame Build-ings in Seismic Areas), sponsored by the EuropeanCommission (Mazzolani, 1999c). The program ofthe joint project has been established in order to

provide an answer to the above questions, by ac-complishing the following objectives: a) Analysis

and synthesis of research results, including CodeProvisions, in relation with the evidence of North-ridge and Kobe earthquakes. Particular attention has

been devoted also to those research results whichhave not accounted for in preparing the provisions ofthe modern seismic codes. b) Assessment of new cri-teria for selecting the behaviour factor for differentstructural typologies and definition of the corre-sponding range of validity. In particular, in thisfield, the need to provide simplified methods forevaluating the q-factor is felt more and more urgent.The aim is to provide the designer with an operative

tool, which allows him to be aware of the inelastic performances of the designed structure. c) Identifi-cation and evaluation of the structural characteristicsof connections influencing the seismic response ofsteel buildings. Therefore, the research has been de-voted to the strength and stiffness evaluation ofmoment resistant connections and to the predictionof their degradation under cyclic loads, by consider-ing also the effect of strain rate and temperature.Low cycle fatigue has been also investigated. d)Definition of criteria for designing and detailing

beam-to-column connections for seismic resistantstructures, considering also the seismicity of the site(far-source and near-source effects) (Mazzolani,1999d, 2000).

8 DESIGN AND APPLICATIONSA slow but increasing tendency to use steelworkagainst earthquake has been observed in the lastdecades. Since 1980 this phenomenon was clearlyevident in Italy mainly in the construction of build-ings, which play a fundamental role during the emer-

gency, like hospitals, fire stations, electrical facili-ties, town halls, and so on.

Also in other southern European earthquakeprone Countries (Bulgaria, Greece, Romania, Portu-gal) this process started and some steel and compos-ite buildings have been erected for important appli-cations.

The Naples Management Centre (CDN) is an ex-ample of this new trend (Figure 14). It is the largesturban development initiative currently underway inItaly and one of the largest in Europe. It consists on

the creation of a new city quarter, which providesresidential and business functions in such a way tomeet the more and more growing demand of the

Naples metropolitan area (Mazzolani, 1995d). Afterthe Campania-Basilicata earthquake of 1980 the city

Figure 14. General view of the Naples Management Centre(CDN).

of Naples was included in the third sismicity zone,the lowest one, which corresponds to a seismic in-tensity coefficient equal to 0.04, according to the na-tional seismic code. At that time for buildings al-ready designed, but not yet under construction or attheir early beginning it was necessary to provide theup-grading of the structural scheme, by keeping thesame typology and introducing suitable strengthen-ing changes in order to fulfil the new seismic re-quirements. It was in the case of many buildings ofthe Management Centre. They range from 25 m to120 m high and most of them belong to the relevantcategory of primary importance for civil protection.Many different solutions have been conceived forthe seismic-resistant structures, both classical andinnovative. The most used classical solution is themixed one, in which reinforced concrete cores pro-vide to stabilize the steel skeleton against the effectof horizontal loads (Figure 15). Many of the follow-ing applications shown here belong to this typology.

Figure 15. Some mixed solutions for buildings in the CDN ofNaples.

The Law Court building is composed by threetowers, which are equal in plan, but have differentheight varying from 78 to 117 m (Figure 16). Theyare connected by an horizontal building in the front

part, containing the Court-rooms. For each tower thestructural system is composed by reinforced con-crete curved walls and cores, which provide strengthand stability under horizontal loads (wind, earth-quakes), and a steel skeleton resisting vertical loadsonly. The steel columns with square box-section arelocated along the perimeter. The main floor beamshave a radial direction and they are pinned to thesteel columns and to the reinforced concrete walls.

The Electrical Department is composed by twotwin towers 120 m high (Figure 17a). Their plan hasa lozenge shape with dimensions 58 x 14. The struc-

ture of each tower is composed by two reinforcedconcrete cores, containing stairs and elevators, con-nected at the top by a box-section girder which the29 stories are suspended to. The suspended structureis made of steel ties and steel-concrete composite


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Figure 16. The Law Court building of Naples: a) layout of thethree towers; b) general view.

floor beams. The horizontal connection betweencores and composite floor structure is provided bymeans of elasto-plastic dissipative devices (Figure

17b), which allow for a significant reduction ofseismic effects, mainly at the base of the cores,where bending moment and shear are reduced of30%. Numerical and physical models have beenused in order to validate this structural solution. Themodel in scale 1:20 has been tested on a shaking ta-

ble with six controlled degree of freedom. The experimental results confirmed the advantages of theuse of such dissipative devices in a suspended struc-ture (Belli and Guarracino, 1997).

The new Fire Station in Naples is composed by

eight buildings in total (Figure 18). They have beendesigned by using steel and composite structureswith both traditional and innovative solutions forseismic protection (designer F.M. Mazzolani, 1986).Building A, containing garage, headquarters, lodg-ings and canteen for fire brigade, is the most impor-tant at least for two reasons: first, because this build-ing, initiated in 1981 and completed in 1985, wasthe first example in Italy of a base isolated structure;

Figure 17. The Electrical Department of Naples: a) general

view; b) dissipative device.

second, because it received the award of the Euro- pean Convention for Constructional Steelworks in1987. The structural scheme is based on a composite

structure, in which the concrete cores, containingstairs, elevators and poles, are spaced about 18 x 18m. They provide the resistance against horizontalforces. The steel skeleton is suspended to a top grid

by means of vertical ties (Figure 19a). The top gridis connected to the upper part of concrete cores bymeans of special devices, which isolate the steelskeleton from the vertical and horizontal ground mo-

tions transmitted to the towers. The bearing devices(Figure 19b), made of a combination of neopreneand teflon, play the double role to allow for freemovements under serviceability conditions and to

provide damping and energy absorption during anearthquake (Mazzolani and Serino, 1997a).

Figure 18. The new Fire Station of Naples: general view.

Figure 19. The new Fire Station of Naples (Building A): a) thesuspended scheme; b) the bearing devices.

Building B, containing garage, pole, storehouse,

guesthouse, training, has a similar plan distributionwith the same structural mesh as the previous build-ing, but the structural solution is different (Figure20a). The only vertical elements are four couples ofsteel towers which have the double function to resistseismic actions and to support the floor slabs bymeans of neoprene bearings. The connection be-tween the main floor beams and the adjacent steeltowers is made by means of shock-block devicesmade of oleodynamic cylinders (Figure 20b). Theyact as provisional restraints, which allow free move-

ments of the structure under service loads (liveloads, wind, temperature, ...), but become rigid un-der seismic actions. They produce the effect tochange the structural scheme in a more favourableway under low intensity earthquakes and to reducethe degree of damageability of the structure undersevere earthquakes (Mazzolani and Serino, 1997b).

Figure 20. The new Fire Station of Naples (Building B): a)general view; b) the oleodynamic cylinder.

a)

b)

a) b)

a) b)

a) b)


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Building E, containing head-quarters, offices andlodgings, contrary to the first two, has a classical so-lution. The steel structure is braced by means of re-ticular concentric bracings (CBF), which stabilisethe structure under horizontal loads (Figure 21). Thefloor structure is composed by primary and secon-dary beams supported by columns, the beam-to-column joints being simply pinned.

Figure 21. The new Fire Station of Naples (Building E).

Different typologies are used in Italy in highersismicity areas, like in Messina (Sicily), where ec-centric bracing frames (EBF) are used for the struc-

ture of the seven storeys building built in 1996 forthe Library of the new Department of Literature andPhilosophy of the University (Figure 22). The shapeof the building is an octagonal prism. The eccentri-cally braced system of K shaped bracings has beencalibrated in order to achieve the simultaneous en-gagement into plastic range of the links located atthe different stories. All connections in the momentresisting frames are rigid and of full-strength type(DAmore and D'Amore, 1997). A seismic resistantsteel structure with eccentric bracings (EBF) has

been designed for the first time in Romania by Dra-gos Georgescu (1996). The Mara Hotel (Figure 23a),located in the picturesque mountain town of Sinaia,has a structure composed by a gravitational systemfor vertical loads and an EBF system resisting seis-mic actions and providing the required stiffness.

Figure 22. The library of the new Department of Literature and

Philosophy of the University of Messina (Italy): a) general

view; b) during erection.

The EBF are of K-braced (Figure 23b) and D-braced (Figure 23c) types, according to architecturalreasons. The links are short and the link-to-column

connection is made of a rigid segment shop-weldedto the column and field-bolted to the link. Also thelink-to-brace connection is field bolted (Figure 23d).The dog-bone solution has been adopted at the

base of the columns, in order to allow for the forma-

tion of a plastic hinge there and consequently guar-antee a global failure mechanism (Figure 23e).

Figure 23. The Mara Hotel in Sinaia (Romania): a) generalview; b) the eccentric K-bracing; c) the eccentric D-bracing; d)link-to-brace connection; e) 'dog-bone' connection at the col-umn base.

EBF systems have been used by the same Authorin the office building of the ROMTRANS Companyin Constanta and in the retrofitting of an office

building in Bucharest (1996).Again EBF are used more recently in the Banc

Post Building (designer D. Dubina, proof engineerV. Gioncu) in Timisoara (Figure 24a). The steelskeleton of five levels is a space moment resistingframe composed by cruciform columns and inte-grated by K-bracings (Figure 24b). The bolted end-

plate beam-to-column connections (Figure 24c) have

been preliminarily tested under cyclic loading at thelaboratory of the University of Timisoara (Dubina etal, 2000). Mixed solutions are very popular also inPortugal. In the campus of the University of Lisbontwo new ten storeys buildings, shaped as mush-rooms, have been recently erected (1994-1999); theyare the Departments of Electronic and Chemistry(Figure25a).Both buildings consist in a central rein-forced concrete core for stairs and elevators, whichsupports large cantilever beams every three levels.

Figure 24. The Banc Post Building in Timisoara (Romania): a)general view; b) the eccentric bracing; c) haunched beam-to-column connection.

a) b)

a) b)

c) d) e)

a)

b)c)


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An apartment and office building of 30 floors is

now under construction in the EXPO area of Lisbon

(Figure 25b). Also in this building the bracing func-

tion is fulfilled by a reinforced concrete core and the

floor structures are composite.

Figure 25. The Departments of Electronic and Chemistry (a)and a multistorey building in the EXPO area (b) in Lisbon(Portugal).

Steel structures are used also in Greece, where thereinforced concrete tradition is very strong. One ofthe first steel and composite building is the Facultyof Physics and Mathematics at the Aristotle Univer-sity in Tessaloniki (1997). The structural scheme is aCBF with inverted V bracings, which are visible insome part of the faade giving an architectural fea-ture to the building (Figure 26). Until now EBFsstructures have been used in Greece just for power

plant structures (Figure 27a) and for upgrading adamaged building (Figure 27b).

Figure 26. The Faculty of Physics and Mathematics at the Aris-

totle University in Tessaloniki (Greece): a) general view; b)

details of bracings during erection.

Figure 27. The use of EBFs in Greece: a) power plant struc-ture; b) upgrading of a damaged building.

EBFs have been recently used in the Dolapderecampus second building at the Istanbul Bilgi Uni-versity, which received the ECCS award in 2005

(Figure 28).Special EBFs have been recently applied as seis-

mic up-grading system for r.c. structures designedfor vertical loads only (Figure 29). The full scale ex-

periments have been performed in Bagnoli (Naples),

Figure 28. The use of EBF in Turkey: Istanbul Bilgi Univer-sity.

giving rise to an increase of load carrying capacityin horizontal direction of about 5 to 6 times (Maz-zolani et al, 2004).

Figure 29. The use of EBF in Italy: seismic upgrading of r.c.building .

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