RESPONSE CHARACTERISTICS OF HIGH STRENGTH STEEL AND ...

RESPONSE CHARACTERISTICS OF HIGH STRENGTH STEEL AND CONCRETE MEMBERS UNDER CYCLIC LOADING

A.S. Elnashai,1 A. I. Salama2 and R. C. Goodfellow3 Abstract In this paper, a wide-ranging experimental program on members constructed from high strength concrete and high yield steel is described. The main objective was to investigate the applicability of existing and established seismic design concepts to concrete of compressive strength of up to 130 MPa and steel with yield up to 1300 MPa. The investigation is focused on the material and member level; the structure-level implications are investigated by the authors in other publications. The testing programme comprises 90 column models with constant axial force and cyclic transverse displacement-controlled loading. A detailed account of the testing compaign and overall results are given in the paper. Three specimens are selected, representative of the range of concrete compressive strength used, are reported in detail. The main conclsions from the presented study are that existing seismic design deformation-based concepts are applicable, with exceptions. Perhaps the most important observation pertains to the role of stirrups in high strength members. Recommended target ductility as a function of axial force and member characteristics are given in the paper alongside comprehensive guidance on deformational seismic response parameters.

1 Professor, Willett Professor of Engineering, Director of Mid-America Earthquake Center, Department of Civil and Environmental Engineering, University of Illinois at Urbana-Champaign, Urbana, Illinois, USA. 2 Associate Professor, Faculty of Engineering of Shoubra, University Zagazig Banha Branch, Egypt. 3 Director, Walsh Associates, Dublin, Republic of Ireland.

Response Characteristics of high strength steel and concrete members under cyclic loading

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1. INTRODUCTION Increased urbanization, which has accelerated over the past few decades, has placed a large demand on the land use in major cities all over the world. This has lead to the necessity of constructing high-rise buildings on the limited and costly land space available. If land shortages and prices continue to rise at the current rate, the expansion in the vertical direction will grow, leading to ever increasing building heights. Coupled to this is the growth in population mobility resulting in large transportation infrastructure projects. Thus, the trend of high strength materials especially to reduce section sizes of vertical elements of reinforced concrete high-rise structures will increase. The advantages of section size and hence weight reduction may be further enhanced if high yield steel in used in conjunction with high strength concrete. Also, the increased demand for steel reinforcement forms another challenge for the construction industry. The utilization of higher strength material is a viable solution to construction in modern societies. Often high strength concrete has evolved as a consequence of the quest for high early age strengths particularly in fast-track construction, or the requirement for more durable concrete subjected to harsh exposure conditions. During construction there may be savings on concrete, reinforcement and shuttering materials. The increase in the stiffness provided by the high strength materials can result in a decrease in the axial shortening of a member and assist in the vibration control of small elastic displacements. The combined effect of the above-mentioned attributes renders high strength concrete economics most appealing. As population and economic growth continues, more new and existing cities are expanding in areas of medium to high seismic risk. Thus, guidance for the use of high performance materials in design and construction of large reinforced concrete structures subjected to earthquake loading is required in order that the advantages especially of section size and weight reduction can be realized.

The term ‘high strength concrete’ is not universally accepted for concrete developed solely for attaining higher compressive strengths. The more globally classification ‘high performance concrete’ encompasses a wide variation of new materials whether they were developed for increased durability, early age strengths, easy placing, or overall performance. However, within the scope of this paper, ‘high strength concrete’ will describe concrete whose compressive strength is greater than 50 MPa. A widely accepted strength categorization is detailed in Table 1 below.

Classification

Compressive Strength (MPa)

Normal < 50 High Strength 50 -100 Very High Strength 100 - 130 Ultra High Strength >130

Table 1 : Definition of concrete classification

Cement, concrete mix designs, mixing procedures, placing and curing have all improved over the years. The increase in concrete strength of a conventional concrete mix is primarily obtained by decreasing the water/cement ratio. However, increasing the cement content may prove uneconomical and decreasing the amount of free water results in a lower workability of the


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concrete mix. With the introduction of water reducing admixtures, satisfactory mix designs with water/cement ratios as low as 0.25 are achieved which yield compressive strengths up to 100 MPa. To achieve strengths over 100 MPa, microsilica as cement replacement is essential and when introduced into traditional mixes compressive strengths up to 130 MPa can be produced. A number of research programs have produced so call super-concretes with compressive stresses up to 800 MPa. These products are mainly developed for specialist applications and tend to consist of special constituent materials. Thus they can not be directly compared to concrete formed with primarily traditional cement and aggregates. It is further recognized that a great deal of work has been carried out on reinforced concrete members containing steel or polymer fiber reinforcement which give certain advantages to the ductility of reinforced concrete members. This however will not be covered by this text; instead it was decided to keep to the mainstream constituents of concrete and steel reinforcing bars.

At present the term high yield reinforcement already applies to the commonly used deformed type bars with yield strength in the range of 400-500 MPa. These were introduced in the 1960s and now most national codes stipulate their use as the main reinforcement within a concrete member. Consequently, these bars have almost universally replaced the use of plain carbon steel with nominal yield strength of 250 MPa. Reinforcing steel of grades higher than 500 MPa, such as pre-stressing tendons, have existed for some time but their use in conventional reinforced concrete applications is limited due the absence of a yield plateau and a low ductility. In the last decade reinforcing steels with yield strength of up to 1400 MPa have been produced in Japan and the USA. Other parts of the world do not seem to place the emphasis on the development of the yield strength, but concentrate on the ductility of the steel and its quality control in fabrication based on a 500 MPa yield limit. Some of the high yield steels so far developed by the Japanese (although the exact data is limited) have a ‘rounded’ stress-strain behavior, i.e. there is no clear yield point, and therefore the definition of this and there use in design needs special consideration as it will significantly affect the desired structural response. For the purposes of this paper high yield steel will denote reinforcement with yield strength greater than 500 MPa.

The behavior of high strength concrete members under static, monotonic loading is reasonably well established. However, much less is known about their response to cyclic loads imposed by earthquake excitations, in particular response due to low axial loads and hence low shear forces where confining forces have not been mobilized. While it is known that attention to the detailing will increase local ductility supply, little information is available regarding evaluation of this ductility supply with low axial loads. A full assessment of the flexural resistance of all members is necessary if the desired failure modes are to be assessed without resorting to undue conservatism. The variation in the resistance of the concrete components due to the confinement needs to be fully included. Whilst it is possible to make conservative assessment of the local ductility supply of individual members, an assessment of the maximum rotational supply of the members experienced under seismic cyclic loads are not as easily quantified. To advance design guidance in these areas an extensive experimental research program has been undertaken. To assess the response of the high strength concrete members, a series of cyclic tests under constant axial loads for increasing amplitude horizontal displacement was performed.


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2. PERFORMANCE OF CONSTITUENT MATERIALS

2.1 HIGH STRENGTH CONCRETE

For high strength concrete, all of its mix components are working closer to their critical limits. These components must all be considered explicitly in the design process as the overall performance relies on each of their optimum values. Consequently the crushing and/or splitting strength of the aggregates becomes the governing factor of the concrete strength. Added to this, strong quality control is required in all the areas of material selection, mix design, handling, placing and curing, especially for on-site conditions. (i) Cement High strength concrete may not only contain ordinary Portland cement, aggregate and water, but also admixtures to improve on the workability of the higher grade mixes as the water-cement ratios need to be reduced to achieve the strength. Portland cements vary widely in their fineness and chemical composition and hence when combined with other admixtures in high strength concrete mixes may give varying results. The greater fineness leads to an advance in the early age strength of the concrete, as an increase in the surface area gives a greater contact with the water and hence more rapid hydration. It is possible to achieve high compressive strengths using fly ash (PFA) or ground granulated blast furnace slag (GGBFS) as the supplementary cementing material. However to achieve strengths in excess of 100 MPa, use of silica fume as a cement replacement has found to be essential and is frequently used for concrete in the strength range of 60-100 MPa (Clayton, 1993)1.

(ii) Aggregates The properties of the aggregate are very important with regard to the concrete strength. Particle shape, size distribution, mechanical properties and chemical reactions with the paste are important factors. Angular aggregates provide better interlocking and bond properties than rounded gravel aggregates, and it is recommended that crushed rock aggregate should be employed within high strength concrete mixes. The particle size is limited to a maximum of 10mm as its strength may become the governing factor in limiting the concrete strength. Good grading of the aggregate is also very important in helping to reduce the water/cement ratio. (iii) Microsilica Microsilica is the ultra fine pozzolanic by-products of the silicon alloy industry, which have been described in many different ways. Most of the research on the effect of these materials on concrete has been concentrated on the high silica fumes, for which the term microsilica is widely accepted. The individual particles are glassy, amorphous micro spheres, with a mean particle size usually between 0.1 and 0.2 microns, i.e. 50 to 100 times finer than OPC or PFA, and as a consequence have a specific surface area that is about 100 times that of cement. When properly dispersed, thousands of reactive micro spheres surround each cement grain, thus densifying the matrix, hence filling the voids with strong hydration products, and therefore improving the bond with the aggregates. (iv) Compressive Strength The uni-axial compressive strength is the most common parameter used to characterize concrete. It depends on the geometry and size of the test specimen, curing conditions, age at testing, etc.


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Conversion factors have been proposed (Chin M.S., et al., 1997)2 which attempt to standardize the results from various types of specimens and methods of testing. There is also the trend that the compressive strength developed in the laboratory is superior to that achieved on site and design codes usually provide relations between them. CEB/FIP (1995)3 recommends that for the higher grade concrete only the cylinder strength at 28 days should be adopted as the characteristic strength. (v) Specimen Shape and Size CEB/FIP (1990)4 highlighted how the size and shape of test specimens varies greatly from country to country; this is summarized in Table 1.2. However it is widely accepted that the diameter of a cylinder, or the side of a cube should be at least three times greater than the maximum nominal size of aggregate. British Standard BS8110 (BSI, 1997) 5 specifies the use of 150 mm cubes, whereas Model Code 90 (CEB/FIP, 1993)6 and the ACI Standard (ACI, 1995)7 require that a cylinder specimen of 150 x 300 mm long must be used to measure the concrete compressive strength. (vi) Tensile Strength Jaccoud, Farra & Leclercq, 19968, derived these expressions from the collated results of approximately 400 different concrete mixes from experimental work. The mean values measured on a limited number of samples were correlated to the mean value of compressive strength. The tensile strength is defined as the axial tensile strength results from the splitting tensile tests transformed using a conversion factor of 0.81. Further research by the same authors lead them to propose a new relation as given below, which was again based on the previous work, carried out on the development of MC 90 for high strength concrete.

Concrete Grade

C30

C40

C50

C60

C70

C80

C90

C100

fck 30 40 50 60 70 80 90 100 fctm 2.8 3.2 3.6 4.0 4.4 4.7 5.0 5.3 fctk,min 1.9 2.2 2.5 2.7 3.0 3.2 3.4 3.6 fctk,max 3.7 4.3 4.8 5.3 5.8 6.2 6.6 7.0

Table 2 : Tensile strengths (MPa) for various concrete grades by Jaccoud, Farra & Leclercq

(vii) Modulus of Elasticity Jaccoud and Leclercq, 19959 derived a new expression for extension to MC90 to calculate the initial tangent modulus of high strength concrete. Results of 400 different concrete mixes were collected and although several parameters may have an important influence (different constituents, particularly aggregates etc.); the results are correlated only to the mean value of the compressive strength. Therefore this expression may be used in design but if the exact elastic modulus is important, then testing of the actual concrete should be undertaken, if results from tests on similar concrete are not known.

Concrete Grade

C30

C40

C50

C60

C70

C80

C90

C100

Ecm (GPa) 33 35 37 39 41 42 44 45

Table 3: Tangent moduli for various concrete grades Jaccoud, Farra & Leclercq


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(viii) Poisson’s Ratio Experimental data on Poisson’s ratio for high strength concrete is limited. However, a review paper on experience in the USA by Ahmed et al., 198210 and concluded that based on available information, Poisson’s ratio for high strength concrete in the elastic range seems comparable to the expected range of values for the lower strengths concrete, approximately 0.18-0.24 depending on the stress level. In the inelastic range, the relative increase in lateral strains is lower for high strength concrete due to less micro cracking. (x) Stress-Strain Behavior The shape of the stress-strain curve for concrete in uni-axial compression is closely related to that of a composite material. The cement paste and the natural rock aggregates are both brittle materials, however the concrete, composed of the two materials, displays some small ductile behavior which may be enhance with the provision of confinement. The difference in rigidity between cement paste and aggregates is far less in high strength than in normal strength concrete. Consequently, the internal stress distribution is more homogeneous, and as the tendency of early cracking is reduced, the ascending branch of the stress-strain curve is more linear. A less steady developed micro crack pattern also results in a more brittle failure, because the ability to redistribute stress is reduced. This manifests in a steep descending branch of the stress-strain curve. The search to define the full stress-strain relationship for normal strength concrete as well as high strength concrete has been on going for years in many research centers (e.g. Ahmad & Shah, 198211; Al Hussani et al., 199312; Almusallam & Alsayed, 199513; Carrasquillo et al., 198114; Mugurama et al., 198815, 199016, 199317) with varying methods of research and levels of conclusion reached.

0

20

40

60

80

100

120

140

0 0.001 0.002 0.003 0.004 0.005Strain

130 MPa

100 MPa

70 MPa

40 MPa

Figure 1: Curve showing stress-strain relationships for unconfined concrete according to model

code 90 and the recommendations of CEB-FIP Bulletin No.

Figure 1 displays the relationships for stress-strain in compression of various strengths of concrete in accordance with Model Code 1990 (CEB/FIP, 1993)18 proposals. The curves have been derived from the recommendations listed in CEB-FIP Bulletin No. 228 (CEB/FIP, 1995)19


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for the extension of MC 90 for concrete grades up to 100 MPa. For illustration purposes the curve for the 130 MPa has been developed simply by extension of these rules, which may not be strictly accurate. The ascending branch up to the maximum stress is more linear for a greater percentage of the maximum stress; the strains at maximum stress are not at a constant value of .0022 as usually adopted for normal strength concrete but increase with increasing strength. The descending branch of the curve is steeper for high strength concrete and there may also be a reduced ultimate strain for medium high strength concrete.

2.2 HIGH YIELD REINFORCEMENT

2.1 Types of Reinforcement With the trend towards higher strength steels and the requirement for good fabrication characteristics, reinforcing steels are now made to high quality standards. Whereas some reinforcement bars are supplied in the form of plain carbon steel with nominal yield strength of 250 MPa, extensive use is now made of higher strength steels with minimum yield strengths of 460 to 500 MPa. The new Euro codes under development will adopt minimum yield strength of 500 MPa for all reinforcing steel. The ACI 318-0220 defines two high grade steels: grade 60 (420MPa) and grade 75 (520Mpa). New types of steel reinforcement developed and produced as part of the Japanese ‘New RC Building Project’(Aoyama et al., 199021,1996; Morita and Shiohara, 199622) has raised the available tensile strength up to 1400 MPa.

2.2.2 Mechanical Properties (i) Yield and Ultimate Limits Morita & Shiohara (1996)23 reported on the development of high strength steel deformed bars for high strength reinforced concrete members. There are five categories of steel are listed in Table 4 and illustrated in Figure 2 and incorporate steels with or without a yield plateau.

Type of Steel Mechanical Properties USD685A

(1) USD685B (1)

USD980

USD785

USD1275

Tensile Yield Strength or 0.2% Offset Strength (MPa)

685-785

685-755

> 980

> 785

> 1275

Maximum Tensile Strength (MPa)

--- --- --- > 930 > 1420

Yield Ratio (Yield/Ult. Strength)

< 85% < 80% < 95% --- ---

Tensile Strain at Hardening (2)

> 1.4% > 1.4% --- --- ---

Fracture Elongation > 10% > 10% > 7% > 8% > 7% Limit of Application For longitudinal as well as

lateral reinforcement For lateral reinforcement only

Notes: (1) - USD685 has two categories A and B. (2) - definition of this property is as shown in Figure 2 (3) - d = nominal diameter of deformed bar. Table 4: Proposed properties of high strength reinforcement from Japanese 'New RC' project.


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Similarly Pipa & Carvalho, 199424 assessed the desired characteristics for reinforcing steel for earthquake applications for incorporation within Eurocode 825.

Figure 2: Stress-strain relationship of high strength reinforcement for the Japanese ‘New RC’

project reproduced from Morita & Shiohara,199626

The requirements of the current draft version of Eurocode 827 for reinforcing bars are derived from the design concepts of the code that aim at providing energy dissipation mechanisms through non linear hysteretic response. There are currently three design ductility classes (high, medium and low) for which the desired ductility of the reinforcing bars are defined. Proposed rules are based on past experience of hot-rolled sections, however recent production of reinforcing bars have resulted in heat treated steel of different properties which may effect the ductility supply.

Characteristic

Ductility Class in Eurocode 8

Tempcore Steel Statistics (from test results)

DCl DCM DCH B400 B500 su (1) > 5% > 6% > 9% 9.6 8.5

(ft/fy)

(2)

> 1.15 < 1.35

> 1.2 < 1.35

1.2 1.16

fy,act/fy,nom (mean value)

< 1.25 < 1.2 1.24 1.17

Notes: (1) Characteristic values at 10% fractile (as in ENV 10080). (2) In EC8 this is specified with reference to type A steel as in Eurocode 2 in which this ratio has to be verified in terms of the 10% fractile (ft/fy)k 0.1 > 1.08 and not in terms of the average.

Table 5: Requirements for reinforcing steel in critical regions from Eurocode 8.

Table 5 reproduced from a draft of Eurocode 8 and outlines the basic requirements for each proposed ductility class. From this table Pipa & Carvalho, 199428 made the following


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conclusions assuming that the statistical values made available were a fair representation of the current production of tempcore steels in Europe: (a) Uniform elongation su Tempcore steel of grade 400 satisfies the EC8 requirements for all ductility classes. Tempcore steel of grade 500 only satisfies the EC8 requirements for the two lowest ductility classes (DCL and DCM). As an aside it was noted that the grade 400 hot rolled steels, to the contrary of Tempcore steel, would not satisfy this requirement due to an excessive ft/fy ratio which was typically of the order of 1.4 to 1.6. (b) Ratio ft/fy Tempcore steel of the grade 400 satisfies the EC8 requirements for all ductility classes. Tempcore steel of grade 500, due to its small strain hardening, only satisfies the EC8 requirements for the two lowest ductility classes. (c) Ratio fy,act/fy,nom In this respect the situation is reversed as compared to the other two requirements above. The grade 400 steel satisfies the EC8 requirements for all ductility classes whereas the grade 500 steel does not satisfy DCH requirements. Both the grade 400 and 500 steels could be used for structures of ductility class L and M, however, for DCH the grade 500 steel is not appropriate but the 400 grade could be adopted provided that its present over strength fy,act/fy,nom is decreased. Such a decrease would also benefit the other requirements. The study by Morita and Shiohara,199629 concluded that high yield steel as lateral reinforcement may have an upper limit of attainable stress as yielding was not necessarily observed at maximum compressive strength of concrete in the straight portions of square hoops and interior ties. Hence the yield stress in the equation to calculate confined strength was recommended to be limited to 700 MPa. On the other hand for circular lateral reinforcement, yielding of the steel was always observed in tests, and the calculated confined strength using yield stress of the lateral reinforcement did not necessarily overshoot. Hence yield stress up to 1,130 MPa, which was the upper limit in the tests, may be utilized. The revised New Zealand Code NZS 3101,1995 places a maximum characteristic strength of 800 MPa on the lateral steel. (ii) Buckling The fundamental assumption in the design of ductile elements subject to compression and/or bending is the stability of the longitudinal bars. This is usually covered in design codes by stipulation of the maximum spacing of the cross-ties in the form of multiples of the diameter of the longitudinal reinforcement. These rules are based on the evaluation of the concrete strength under confinement and the analysis of the response of the longitudinal bars subjected to monotonic and cyclic stress (Monti and Nuti, 199230). As assumed by Monti and Nuti, 1992, the stability of the longitudinal bars in considered based on the centre to centre spacing of two consecutive transverse bars assuming the ends of the longitudinal bars are fixed. The transverse reinforcement of reinforced concrete members is usually designed with reference to the shear loads obtained through the elastic analysis. For inelastic analysis, the influence of the transverse reinforcement must be studied in relation to the inelastic shear and the flexural load carrying capacity in terms of ductility. The lateral confinement improves the peak compressive strength of the concrete core, and the ultimate compressive strain with compressive softening. This phenomenon is coupled with compressive stressing of the longitudinal bars and the design is aimed at reducing the lateral buckling of the bars. This formed the background to the work carried out by Albanesi and Bondi, 199431.


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3 EXPERIMENTAL SETUP

3.1 ADOPTED MIX DESIGNS FOR HIGH STRENGTH CONCRETE

The mix designs used were based on the extensive research program undertaken by the UK Building Research Establishment (Clayton, 199332). The main aim of the BRE was to produce practical high strength concrete mixes that could be easily employed in reinforced concrete design and construction.

Concrete Grade C70 C100 C130 OPC (kg/m3) 460 500 500 Sand (kg/m3) 659 665 639 10 mm Aggregate (kg/m3) 1104 1200 1200 Microsilica (kg/m3) - - 50 Superplasticiser (kg/m3) - 4 7.5 Water/Cementatious Ratio 0.31 0.22 0.22

Table 6: Details of experimental concrete mix designs.

British Standard 8110 (BSI, 1997)33 restricts the minimum water-cement ratio to 0.45 for normal strength concrete, this was reduced to 0.31 to achieve a characteristic strength of 70 MPa, whilst still achieving an acceptable workability without using a superplasticiser. To achieve a very high grade mix to a target strength of 100 MPa, a reduced water-cement ratio of 0.22 was used with 0.8% of superplasticiser to produce a satisfactory workability. For an ultra-high concrete mix, microsilica was added to the mix adopted for the very high concrete in order to produce a 130 MPa compressive strength. Additional superplasticiser was required to achieve a satisfactory workability. Based on these principals, a series of trial mixes were cast and tested. The mix designs selected for the testing programme are listed in Table 6. Four concrete cube specimens 100x100x100 mm were taken from the design mixes at the time of casting each of the beam-column models. In addition a series of three concrete cylinders 75 mm diameter and 150 mm high were extracted form a sample of each of the three concrete mixes. A number of cubes were tested at 7 and 28 days to monitor the compressive strength gain. As recommended by the proposed extensions to Model Code 1990 (CEB/FIP, 1995)34 the classification for compressive strength of concrete should be determined from the cylinder strength. Due to the number of beam-column tests (90 in total) and hence the number of material strength tests required, compressive cube tests formed the majority of the compressive strength tests for simplicity and to negate the need for preparation of the ends of cylinder specimens. However, for comparison purposes a number of cylinders where also taken from the model mixes and tested under uni-axial conditions. Due to the brittleness of high strength concrete, these tests were undertaken in a closed tri-axial load cell with no confining stress in order to establish the full stress-strain curve for the cylinder in addition to the maximum strength.


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0

20

40

60

80

100

120

-0.006 -0.004 -0.002 0 0.002 0.004 0.006

Lateral Strain Axial Strain

Axi

al S

tres

s (M

Pa)

Grade 130

Grade 100

Grade 70

Figure 3: Experimental stress-strain results for unconfined concrete cylinders

The uni-axial cylinder tests where compared to the equivalent compressive cube strengths to correlate the relationship between the maximum compressive strengths found from the respective test sample and method. The results are tabulated in Table 7.

Concrete Grade C70 C100 C130

Cube (100mm square)

71.4 100.6 125.0

Cylinder (75mm dia x 150mm)

60.7 80.6 101.8

Cylinder/Cube 0.85 0.80 0.81

Table 7: Compressive strength test results from cylinder and cube specimens (MPa)

The axial load was applied under displacement and strain control in order to attain the descending branch after ultimate. Many previous testing programs have reported that this descending branch is extremely difficult or even impossible to achieve due the brittleness of the material. The tests carried out at Imperial College attained a relationship for this descending branch but as can be seen from the results the number of points on the descending branches are limited and irregular, so the results should be viewed with some trepidation. Furthermore, the tests also monitored the lateral stress/strain properties of the concrete cylinders which highlighted the low levels of dilation common to high strength concrete. The lateral dilation was measured by attachment of surface electronic strain gauges to the sides of the specimen. The closed triaxial cell was used further used in a more conventional manner to obtain the stress-strain relationships of the high strength concrete cylinders subjected to 5, 10 and 15 MPa lateral confining stresses. Although the lateral pressure is applied as an active rather than a passive confining manner, the results provide useful measurement of the effects of the


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confinement on each concrete grade. This permits an accurate assessment of the loading history effects on the core concrete within reinforced concrete members. The results from the tests are plotted for each concrete grade in Figure 3 with the maximum stress and corresponding strain values collated in Table 8 under each level of confinement. Although the confined curves display a more parabolic stress-strain than the equivalent unconfined relationship.

Concrete

Grade

70

100

130

Confining Stress (MPa)

Max Stress (MPa)

Strain at Max Stress

Max Stress (MPa)


Max Stress (MPa)


5 87.8 0.0054 114.2 0.0048 142.9 0.0048

10 115.9 0.0085 140.6 0.0063 172.4 0.0063

15 133.1 0.0117 158.0 0.0077 189.2 0.0072

Table 8: Experimental maximum stress and corresponding strain results for confined

concrete cylinders for each grade of concrete subjected to 5, 10 and 15 MPa confinement

Grade C100 Concrete

020406080

100120140160180200

0 0.002 0.004 0.006 0.008 0.01 0.012 0.014 0.016Axial Strain

Axi

al S

tres

s (M

Pa)

15 MPa

10 MPa

5 MPa

Unconf

Grade C70 Concrete

020406080

100120140160180200

0 0.002 0.004 0.006 0.008 0.01 0.012 0.014 0.016Axial Strain

Axi

al S

tres

s (M

Pa)

15 MPa

10 MPa

5 MPa

Unconf

Grade C130 Concrete

020406080

100120140160180200

0 0.002 0.004 0.006 0.008 0.01 0.012 0.014 0.016Axial Strain

Axi

al S

tres

s (M

Pa)

15 MPa

10 MPa

5 MPa

Unconf

Figure 4 : Experimental uni-axial stress-strain curves for confined concrete cylinders for ach grade of concrete subjected to 5, 10 and 15 MPa confining stresses


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The combined results from the unconfined and confined tests confirm the brittleness of the high strength concrete and how this characteristic rises with an increase in the compressive strength of the concrete. However, the triaxial tests do indicate that if substantial high levels of the confinement can be induced; there is large potential in raising the strain capacity of high strength concrete.

3.2 TYPES OF HIGH STRENGTH STEEL UTILIZED

An extensive search of the steel reinforcement market revealed that although there has generally been no restriction on the upper level of yield strength, the use steel with an average yield of 400-500 MPa has become the most universally adopted steel in reinforced concrete design. Table 9 details the high yield steel reinforcement that was available to this research program along with their source and nominal material properties. The grade 500 steel produced by a British manufacturer conforms to the requirements of the proposed Eurocodes. The same manufacturer fabricated the grade 900 steel specially produced for a project in France. The other high grades of steel come from Japanese fabricators, developed for use in reinforced design and construction as a result of the ‘New Reinforced concrete Project’.

Application

Source

Nom. Steel Grade

Bar Dia. (mm)

Yield Stress fy (MPa)

Tensile Strength ft (MPa)

ft/fy

Elong. at failure u (%)

Cardiff Rod (UK)

500 16.0 570 670 1.18 8

Longitudinal

Kobelco (Japan)

685 19.0 721 921 1.28 10

Steel Kobelco (Japan)

785 16.0 859 1025 1.19 10

Cardiff Rod (France)

900 16.0 830 1100 1.33 4

Neturen (Japan)

1300 13.1 1408 1504 1.07 9

Cardiff Rod (UK)

500 6.0 570 670

Link Steel Kobelco(Japan)

785 10.0 957 1047

Neturen(Japan)

1300 9.0 1371 1470

Table 9: Nominal Steel Reinforcement Characteristics (values supplied by fabricators).

Monotonic tensile tests of a sample of each of the longitudinal reinforcing bars were carried out at Imperial College in order to confirm the exact properties of the longitudinal bars used in the beam-column tests. Direct tension tests were executed on 500 mm long specimens, with 100 mm of each end clamped into wedge profile jaws. A small area of the bar surface in the centre portion of each specimen was smoothed to receive two strain gauges diametrically opposite. The tensile tests were carried out under slow displacement control up to fracture in order to attain the full stress-strain curve of each steel grade as plotted in Figure 4.


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The axial force, strains from surface mounted gauges and displacement of the test machine jaws were automatically collated using electronic data acquisition. The axial stress is plotted against the average strain from the surface gauges for each grade of steel in Table 9. The test machine jaw to jaw displacement was used to supplement the strain values in instances where the surface gauges have failed or de-laminated especially at towards the latter reaches of the loading history and Table10 summarizes the values of stress and strain at yield and failure levels.

0

2 0 0

4 0 0

6 0 0

8 0 0

1 0 0 0

1 2 0 0

1 4 0 0

0 % 1 % 2 % 3 % 4 % 5 % 6 % 7 % 8 % 9 %

S tr ain

Str

ess

(MP

a)

Gr a de 1 3 0 0

Gr a de 9 0 0

Gr a de 7 8 5

Gr a de 6 8 5

Gr a de 5 0 0

Figure 4: Monotonic Tests Results for Longitudinal Reinforcement Each grade of reinforcement was tested under direct tensile loads to failure. The elongation measured at fracture of the reinforcement showed reasonable values of approximately between 6 and 8%. Thus all bars satisfy the Eurocode 8 requirements of elongation for medium ductility class (DCM) but fall just short of the limits for high ductility class (DCH). The modes of failure were varied with the majority of the bars displaying a traditional ‘cup and core’ failure plane. The grade 500 steel broke with jagged sharp edges of the failure zone whereas the grade 900 steel resulted in a clean horizontal fracture. The stress ratio values for the steel (ultimate to yield stress ) were experimentally derived as between 1.13 and 1.27. Grade 500 steel possessed the 1.13 value which is just below the minimum requirement of 1.15 for DCM of Eurocode 8 and well below the lower limit of 1.2 for DCH. The values of minimum and maximum stress ratios for the remaining reinforcement types satisfy both the ductility classes.

Grade fy (MPa) fu (MPa) fu/fy ��(%) 500 540 613 1.14 7.3 685 729 926 1.27 7.2 785 815 984 1.21 7.5 900 851 1113 1.31 6.6 1300 1204 1315 1.09 8.0

Table 10: Experimental Characteristic Properties of Longitudinal

Reinforcement


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fy (Mpa) fu (Mpa) fu/fy

Grade Nom Exp

Compar Nom Exp Compar Nom Exp Compar

500 570 540 0.95 670 613 0.92 1.18 1.14 0.97 685 721 729 1.01 921 926 1.01 1.28 1.27 0.99 785 859 815 0.95 1025 984 0.96 1.19 1.21 1.02 900 830 851 1.03 1100 1113 1.01 1.33 1.31 0.97 1300 1408 1204 0.86 1504 1315 0.88 1.07 1.09 1.02

Table 11: Comparison of Actual-to-Nominal Characteristic Properties of Longitudinal

Reinforcement

Actual-to-nominal values of yield and ultimate strengths and stress ratio are compared in Table 11. For every grade of steel apart from grade 1300, the experimental values of yield and ultimate strengths are within -/+ 5% of the nominal values stipulated by the fabricator. This concludes the fact that chosen samples of reinforcement used within this project were manufactured to a very high quality standard. Under dynamic loading especially earthquakes all structural members may be subjected to a large number of stress reversals. The flexural response of a column within a reinforced concrete frame will be dictated by the effect of the cyclic loading on the confined concrete core and the longitudinal reinforcement. Thus it is important to try and quantify the effect of cyclic loads on the reinforcing bars.

Cyclic tests were carried out on samples of each grade of the longitudinal bars at increasing levels of elastic and subsequent plastic strain amplitude. A small specimen length of 200 mm was adopted in an attempt to prevent premature buckling. However, the specimen had to be long enough to allow access for a surface mounted electronic clip gauge. The bars were tested within a purpose built apparatus connected to an Instron controlled cyclic loading jack. The ends were attached to specially fabricate high-grade steel sleeves which provide restraint in both tension and compression. Firstly the ends of the specimen received a cut thread which are screwed into the end of the sleeves. Secondly high grade steel pointed screws along the length of the sleeves connect to sockets drilled with in the ends of the test sample to hold it in place. The tests were executed under strain control via the specimen extension measured within the zone of the clip gauge. The strain amplitude was gradually increased until fracture or buckling of the reinforcing bar.

The experiments met with differing degrees of success. Generally for all of the tests it was virtually impossible to achieve large levels of plastic strain cycles. However the curves do illustrate that cyclic degradation of the steel occurs even under small strain amplitudes. Therefore the effect of cyclic loading on the capacity of a section is significant. Generally the specimens failed due to compressive buckling. Plastic cycling of the grade 1300 bars was not possible due to the smaller diameter of the bars buckling prematurely.


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Grade 500

-600

-400

-200

0

200

400

600

-0.01 -0.0075 -0.005 -0.0025 0 0.0025 0.005 0.0075 0.01

Grade 685

-600

-400

-200

0

200

400

600

-0.01 -0.0075 -0.005 -0.0025 0 0.0025 0.005 0.0075 0.01

Grade 900

-1000

-800

-600

-400

-200

0

200

400

600

800

1000

-0.01 -0.0075 -0.005 -0.0025 0 0.0025 0.005 0.0075 0.01

Grade 1300

-1200

-1000

-800

-600

-400

-200

0

200

400

600

800

1000

1200

-0.01 -0.0075 -0.005 -0.0025 0 0.0025 0.005 0.0075 0.01

Figure 5 : Experimental Results of Cyclic Tests of Longitudinal Reinforcement

4. EXECUTION OF THE MEMBER TESTS

4.1 Test Procedure And Programme

The beam-column test members represent half of a storey high column in a theoretical shear frame, i.e. modelling the column between the point of maximum moment (assumed to be at floor slab level), and the point of contra-flexure (assumed to be at mid-height). A member with a cross-section 250 mm square and 1250 mm high was adopted for the experiments. The cross-section was chosen as the practical minimum dimensions that a vertical member could have. Likewise a storey height of 2500 mm would be absolute minimum for most modern day buildings. As supported by the numerous testing programmes carried out by Park35 and his co-workers (Park,1995), and confirmed by extensive experimental research at Imperial College, vast knowledge of the behaviour of members and sub-assembalges is gained from tests carried out under displacement controlled static cyclic loading. Seisimic forces are dynamic by nature and an increase in the loading rate will generally produce an increase in the strength and stiffness of concrete and steel. Therefore the reulsts from the static cyclic tests yield conservative estimates of the real strength of a reinforced concrete member.


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Figure 6: (top) Beam-Column member test set-up, (bottom left) Strain Gauges linked from back of model, (bottom right) testing member at 30 mm displacement


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4.2 BEAM-COLUMN SPECIMEN

Each beam-column member was designed and detailed in accordance with the draft rules of EC8 for a medium ductility class. The members contain eight longitudinal bars, one in each corner and an intermediate bar on each face. Confining reinforcement arrangement is via a pair of links, one around the outside of all eight longitudinal bars, and the other around the four intermediate bars only. This configuration of the link steel has been widely proven to give effective confinement of the concrete core. The spacing of the links along the height of the model are dictated by the design rules of EC8, some are placed tighter than these rules and others further apart. The typical configuration of the models is illustrated in Fig.6. The models, ninety in total, were constructed from combinations of the three grades of high strength concrete; (70, 100 and 130 MPa), the five grades of high yield longitudinal reinforcement and the three grades of link steel at various centers; (500, 685, 785, 900 and 1300 MPa). This involved the procurement, detailing and fixing of approximately 4 tonnes of reinforcing steel; the processing, batching mixing and casting of 18 tonnes of concrete; the preparation and placing of 1500 strain gauges; the casting, demoulding and testing of 500 concrete test cubes and cylinders; and the processing of approximately 40 Mb of output from the experimental tests. Each grade of the longitudinal reinforcement with exception of the grade 1300 bars was cropped to correct length and received a cut thread to one end. The grade 1300 bars were supplied to the correct model length and had a preformed thread provided by the fabricators. Providing threaded ends to the reinforcement was the only practical way of attaining anchorage of the bars to the model base plate. Ten model base plates each fabricated from 75 mm high grade steel. Internal electronic strain gauges are placed on the surface of a selection of the longitudinal and lateral reinforcement. Assembly of the reinforcing cages was undertaken after the relative bars had fully gauged. During this operation particular care was taken not to damage any of the gauge assembly especially when introducing steel wire to connect the bars together. Once the reinforcement cages were assembled, the models were placed on their sides in timber formwork. The longitudinal reinforcing bars were passed through and bolted to the back of the 75 mm thick steel baseplate with high grade steel washers and nuts, thus ensuring anchorage of the bars. The timber mould and the steel baseplate were fully aligned then clamped together and the complete assembly placed on a vibrating table. At this point the electronic gauges were re-tested to ensure that no damage had occurred during construction.

4.3 EXPERIMENTAL SET-UP

4.3.1 Test Rig After the minimum curing period the beam-column members were transferred to an internal reaction steel framed test rig as illustrated in Fig.6. The test-rig composed of an internal reaction frame, assembled from custom-made structural sections connected by tensioned bolts, has a design capacity of 400 tones. Prior to installation the beam-column model was fitted with a removable steel collar clamp, which ensures that the model fails at a position away from the junction of the concrete and the steel baseplate. The model/collar assembly is placed inverted via the top of the test frame for simplicity in installing and removing each of the models without having to disturb the loading


19

jacks. The model base plate and collar are clamped to the top plate of the test rig via high-stress steel threaded bars. The top plate itself, of dimensions 1200 x 1200 x 130mm, is in turn stressed to the test frame by 44 no. 24 mm diameter bolts. A constant axial load of 10% of the gross axial capacity of the section is applied to the test member, plus a displacement controlled horizontal loading history. The horizontal displacements are mainly increasing amplitude cyclic; however, a small number of monotonic tests were also executed. Two double-acting hydraulic actuators are used to apply the axial loads and lateral displacements. A vertical actuator, located directly below the centerline of the specimen, is employed to impose axial loads. Connection to the specimen is by means of a low-friction custom fabricated steel ball and socket bearing which allows free rotation of the cantilever tip to occur. An identical arrangement is provided at the base of the actuator, enabling the actuator to rotate freely as lateral displacements are imposed on the specimen. The horizontally-acting actuator, which is used to impose the control displacements at the tip of the cantilever specimen, is connected to model via a prefabricated steel box by means of a 500kN capacity Instron low-friction swivel bearing. An identical bearing is attached to the opposite of the actuator, enabling the actuator to deflect freely as lateral displacements are imposed on the specimen. This hinge bearing is attached by 8 no. 24mm diameter bolts to a 120mm plate. This plate is in turn connected to the test frame by 24 no. 24mm diameter grade 8.8 prestressed bolts.

Both the horizontal and vertical actuators are Enerpac actuators connected to an Instron servo-controller. The load capacities and stroke capacities of the vertical and horizontal actuators differ, being 1000kN with +/-100mm and 250kN with +/125mm respectively.

Out-of-plane movement of the specimen during testing is unavoidable especially after major spalling of the concrete cover. In order to limit the extent of this movement, a pair of cam-rollers were fixed to 150 x 150 x 6mm rectangular hollow sections the side of the test frame.

4.3.2 Hydraulic Power and Servo-Control Systems The hydraulic power used by the actuators to impose displacements and loads on the test specimen is provided by two power packs with a combined flow capacity of 460 liters per minute connected to a common ring main complete with remote start up facility. The hydraulic flow is fed through a 230 liters per minute hydraulic sub-station to the actuators via six hydraulic hoses (pressure, return and drain for both actuators). A start up pressure of 100 psi and a full operational pressure of 3000 psi was available to the actuators. The force applied by the actuator is developed by the hydraulic pressure acting on the cross-sectional area of the piston. Control of the actuator is achieved by means of a servo-controller, which compares the command signal voltage with the displacement transducer voltage output signal. The difference between these is converted into a current which is fed though a servo-valve, causing the actuator to move by directing oil to either end of the actuator piston. This process is performed in a closed loop until the output displacement signal has reached the command signal, as determined from the displacement feedback from a Linear Variable Displacement Transducer (LVDT). If the actuator is being operated under force control, as is the case for vertical actuator described above, the feedback signal comes from a load cell, not the LVDT. In either case, optimum


20

controller loop gain settings are used to ensure best performance of the equipment and to avoid an oscillatory approach to the command setting. The LVDT is permanently fixed within the actuator, but the load cell must be independently selected and fitted to the end of the piston by means of male/male couplings. To match the capacities of the actuators, Instron 2518 Series Fatigue Resistant Load Cells with dynamic ratings of 500kN and 250kN were employed for this purpose.

The proprietary test software employed for all the tests was MAX, which is specifically design by Instron for use with their loading controllers. The software allows not only preset automatic loading curve but also import of any random defined curve in terms of displacement against time. This latter capability was availed of during this research program. Cyclic curves on increasing amplitude were determined within the software program EXCEL and exported to the computer connected to the Instron controller. A 32 channel Solartron data logger was incorporated into the test system connected to computer separate from the loading control system. Four of the channels were used in the to record the measured imposed displacement and resulting horizontal restoring force and similarly to monitor the variation in vertical load and displacements. The remaining channels were employed to record the strains from the sixteen internal electronic gauges placed on both the longitudinal and transverse reinforcement within the critical region of the member were also monitored, plus displacements from four transducers positioned along the length of the specimen.

5. EXPERIMENTAL RESULTS Ninety models were tested under cyclic loading, three samples representing of three different material categories are being presented within the current study. These had test numbers of 8, 50 and 82 with. The following material characteristics were used for the models:

Test Model 8: Concrete grade 70 MPa, Longitudinal Steel grade 785 MPa and steel links 500 MPa@ 50 mm centre line to centre line (c/c).

Test Model 50: Concrete grade 100 MPa, Longitudinal Steel grade 900 MPa and steel links 500 MPa@ 50 mm c/c.

Test Model 82: Concrete grade 130 MPa, Longitudinal Steel grade 900 MPa and steel links 785 MPa@ 70 mm c/c.

All beam-column members were tested under increasing amplitudes of cyclic load up to failure. The majority of models were subjected to a horizontal loading history of two cycles per amplitude of; +/- 10, +/- 15, +/- 20, +/- 30, +/- 40, +/- 60, +/- 80, +/- 100 mm. Not all models reached the later amplitudes as generally they had degraded to a level where their horizontal load capacity was greatly reduced, in such cases the testing was terminated early. By referring solely to the cyclic response graphs, comparisons between models for the full range of the horizontal load-displacement responses can be difficult. However in the main they illustrate that the models display reasonable symmetrical hysteretic behaviour. The main feature of the graphs is that the level of confinement does not appear to have any major influence on the member behaviour particularly after ultimate capacity of the members were reached. This is due to the fact that the full confining effects of stirrups was not mobilized during the tests. Hence although the available confinement rises for each series of tests, the effective confinement would appear to be almost constant..


21

Test Model 8 showed no cracks at +/- 10 mm maximum displacement amplitude (MDA), four tensile cracks developed at 15 mm (MDA) covering half the section depth. These progressed into five cracks at 20 mm (MDA) and covered two third of the depth. At 30mm (MDA), minor crushing of concrete cover to both faces with slight spalling of outer layers. Spalling continued at 40 mm (MDA), this continued into large amount of cover spalling at 60 mm (MDA). At 80mm (MDA) severe degradation took place along with drop of strength by 25% while no severe degradation within the central core took place. The tested model showed large amount of energy dissipation from cyclic hysteresis loops and failed in a most ductile failure mode. Test Model 50 showed some tensile cracks evenly spaced at +/- 10 mm maximum displacement amplitude (MDA). Very slight crushing of corner concrete took place at 20 mm (MDA). At 40 mm (MDA), tensile cracks to three quarters of the height took place along with failure of cover concrete to half the depth without large spalling. this continued into large amount of cover spalling at 60 mm (MDA) along with severe degradation. Further degradation took place at 80mm (MDA) without severe deterioration to the central core. The tested model showed very large amount of energy dissipation from the hysteresis loops. Longitudinal steel proved to have yielded and recorded a higher ultimate strength when compared with TM 8. Test Model 82 showed no cracks at +/- 10 mm maximum displacement amplitude (MDA), four tensile cracks developed at 15 mm (MDA) covering half the section depth. Initial very slight crushing of cover took place on both faces at 20 mm (MDA). At 30mm (MDA), sudden explosive failure took place of cover concrete but with little spalling. Spalling continued at 45 mm (MDA), this continued into large amount of cover spalling at 60 mm (MDA) along with degradation. At 75mm (MDA) some small degradation of core between stirrups took place. Failure of one longitudinal bar in tension on first cycle terminated the test. The tested model showed reasonable amount of energy dissipation from cyclic hysterisis loops. The highest ultimate strength was achieved. Relatively larger link spacing allowed damage to central core.

5.1 DAMAGE PATTERNS

The following forms an overview of the visual observations of the beam-column member tests. The models all developed evenly spaced horizontal tensile cracks with the early stages of loading, and each reached the maximum load capacity with a compression failure of the cover concrete. Spread of damage along the length of the member tended to be larger with increased axial capacity of the longitudinal steel except for the grade 1300 steel. Failure of the cover concrete generally occurred at horizontal deflections of between 25 to 35 mm of the tip of the model and so represents an inter-storey drift of 2 to 3%. The amount and brittleness of the concrete cover failure increased with both an increase in the compressive strength of the concrete and in the axial stresses applied to the member.

There was no noticeable failure of any of the links and in a number of cases there was no yielding of the highest grade steel. For members with confining steel at 35 mm spacing, no buckling of the longitudinal steel bars occurred, even after the concrete cover had spalled and at high displacements. There was also no degradation of the majority of core concrete, indicating that the link steel provided a good level of confinement as well as restraint to the longitudinal bars. In a few cases, the longitudinal steel in members with the wider spaced confining steel (50 and 70 mm) showed signs of having buckled to varying degrees. This again always occurred at high horizontal drift values (5-6%) when the cover concrete had completely spalled.


22

Figure 7: Hysterisis Loops & Crack Pattern for Model 51

Figure 6: Hysterisis Loops and Crack Pattern for Model 8

Figure 8: Hysterisis Loops & Crack Pattern for Model 82


23

5.1 DAMAGE PATTERNS

The following forms an overview of the visual observations of the beam-column member tests. The models all developed evenly spaced horizontal tensile cracks with the early stages of loading, and each reached the maximum load capacity with a compression failure of the cover concrete. Spread of damage along the length of the member tended to be larger with increased axial capacity of the longitudinal steel except for the grade 1300 steel.

Failure of the cover concrete generally occurred at horizontal deflections of between 25 to 35 mm of the tip of the model and so represents an inter-storey drift of 2 to 3%. The amount and brittleness of the concrete cover failure increased with both an increase in the compressive strength of the concrete and in the axial stresses applied to the member.

There was no noticeable failure of any of the links and in a number of cases there was no yielding of the highest grade steel. For members with confining steel at 35 mm spacing, no buckling of the longitudinal steel bars occurred, even after the concrete cover had spalled and at high displacements. There was also no degradation of the majority of core concrete, indicating that the link steel provided a good level of confinement as well as restraint to the longitudinal bars. In a few cases, the longitudinal steel in members with the wider spaced confining steel (50 and 70 mm) showed signs of having buckled to varying degrees. This again always occurred at high horizontal drift values (5-6%) when the cover concrete had completely spalled.

5.2 HORIZONTAL LOAD-DISPLACEMENT RESPONSE

All beam column members except for models 1, 49, 74 and 85 were tested under increasing amplitudes of cyclic load up to failure. The majority of models were subjected to a horizontal loading history of two cycles per amplitude of; +/-10, +/-15, +/-20, +/-30, +/-40, +/-60, +/-80 and +/-100 mm. Not all models reached the latter amplitudes as generally they had degraded to a level where their horizontal load capacity was greatly reduced, in such cases the testing was terminated early. The summary sheets for each of the beam-column tests contained within Appendix A, give full details of the load history and response of each model. However all the tests specimens did successfully reach a displacement amplitude of 60 mm which equates to a drift of approximately 5% of the member height. In practice if this were an inter-storey drift value, extensive damage to any structure and its fixtures would have occurred. For this reason the graphs plotted in Figures 7, 8 and 9 show only the load–displacement responses up to and including +/- 60 mm.

The main feature of the graphs is that the level of confinement does not appear to have any major influence on the member behavior particularly after ultimate capacity of the members were reached. The hysteretic behavior of the models displays good symmetry for both positive and negative for a vast majority of the models. This helps to confirm the accuracy of the experimental set-up, testing procedure and processing of the output.

It can be confirmed that the level of confinement has no noticeable effect on improving neither the ultimate capacity nor the slope of the descending branch of the envelope curves. The ultimate capacity of a member is dependent mainly of the strengths of the concrete and longitudinal steel, and is little affected by the amount of confining steel. The post-ultimate drop in the capacity of the specimens is governed more by the spacing than the yield strength of the confining reinforcement. This and the point above can de attributed to the fact that as the level


24

of axial load is low (i.e. 10% of axial capacity), the full possible passive confinement characteristics of the confining steel is not realized. The most common trends are that a general increase in the longitudinal steel grade, with the exception of grade 1300, produces a steeper ascending branch and shallower descending branch for equivalent concrete grades. Alternatively, an increase in the concrete strength leads to both steeper ascending and descending branches. The effects that these characteristics have on the plastic hinge length and ductility of the members will be discussed in detail in the following relevant sections.

It may be concluded that there is reasonable linear correlation between an increase in the ultimate strength of the member and the compressive strength. This indicates reasonable repeatability of the tests with the same concrete and longitudinal steel values. It also confirms that the level of confining steel present within the model does not have any great effect on the ultimate strength gain.

5.3 LONGITUDINAL STEEL STRAINS

Firstly by studying the strain values, it is clear that the vast majority of the gauges on the grade 1300 do not register yield values throughout the duration of the tests. In a few isolated cases the yield strain was attained but this generally occurred at large tip displacements of the model. This latter effect is more likely to be due to the anchorage of the bars to the back plate rather than interaction with the steel core.

The steel yield values were used to derive the yield limit states of the models, in addition to the derivation of the section curvature along the height of the model at the various limit states as discussed later in the appropriate sections.

For the members without grade 1300 steel, yield strain of the reinforcement was achieved in the vast number of cases before or close to the maximum capacity of the section confirming that the models have failed in under-reinforced or balanced section flexural mode. In the cases where the strain gauges have not recorded a yield strain in the critical section of the model, the yield limit has been estimated by comparing them with models of equivalent load capacity. The reasons for these low readings of strain may be attributed to the localized bond slip and/or delamination of the strain gauge, especially when failure of the cover occurs.

5.4 CONFINING STRAINS

Strain gauges are attached to the outer reinforcement link just above the critical section of each of the members. When the initial axial loads were applied to the models the measured strains on the gauges varied from 0.00003 for an axial load of 500 kN to 0.0001 for the maximum axial load of 950 kN (for models 1 to 90 only). These stain values confirm that at such low levels of axial capacity there is little or no mobilisation of the confining stresses.

By inspection of the tabulated values for maximum confining strains throughout the tests, it is noted that although the values of strain increase there is little occurrence of yield in the any of the links. Where some yielding has taken place (e.g. in selected models 31, 50, 67, 68, 80) this is always after extensive spalling of the concrete cover and the values of strain are still close to


25

the yield level. However, these strains may not solely be due to deformation of the core and can be caused by buckling of the longitudinal steel and/or bending of the link across the critical section.

5.5 YIELD AND ULTIMATE LIMIT STATES

To derive design values for both force and displacement based design of reinforced concrete members, levels have to be set on the yield and ultimate limit states for the beam-column sections. The derivation of these limits are not strictly formulated and a universal method must be set in order to be able comparison between members with different steel and concrete properties A common approach for normal strength reinforced concrete members is via the Bertero and Mahin36 approach, which is based on fitting an equivalent straight line to the rising branch of the load-deformation curve. This line is drawn from the origin through a point on the ascending branch of the rising curve to intersect a horizontal line at the ultimate load level, so that there are equal areas above and below the rising curve. The value of displacement at the intersection between the two straight lines defines the yield limit. This is due to the fact that the steel has yielded at a low displacement of the member and to use that as the yield point would be unrealistic. Failure is taken not at the ultimate capacity of the member, as it still has some residual load carrying capacity, but at the level of 85% ultimate load capacity on the descending branch. At this point the member is assumed to have degraded to an extent where it is no longer capable of supporting design load levels Adopting the Bertero and Mahin37 approach, each ascending branch of experimental beam column envelope curves was studied separately in order to find their yield values. From this equal area method the yield limit definition was calculated to be consistently between 70 % and 80 % of the ultimate load capacity of the member. However in a large number of cases the reinforcement had not actually reached its yield strain at this derive limit. Therefore it was deemed more appropriate to set the yield limit state at the yield strain of the longitudinal steel. The ultimate limit state is calculated at the 15% drop in the maximum strength capacity of the beam column section but is limited to a maximum horizontal drift of 3% which equates to an inter-storey drift of 80 mm.

5.6 STIFFNESS

(i) Initial Stiffness As expected for each steel category, the initial stiffness increases with an increase in the concrete compressive strength. This is due to the increase in the concrete initial modulus with strength. The stiffness increases by up to 10% when grade 100 concret is replaced by grade 70 concrete and by a further 10% maximum when grade 130 concrete is used. As the elastic modulus for each of the longitudinal reinforcing bars are similar, it would be expected that members with the grade 685 bars possessing the larger diameter would have greater initial stiffness than the other models. This is true when the grade 685 models are compared to those with grade 500 and 900 steel, however this does not extend to the grade 785 members. The variation in maximum to minimum stiffness for each steel within the same grade of concrete is 9%, 17% and 10% for 70, 100 and 130 MPa concrete respectively. As the


26

reinforcement cages are all constructed to the same tolerances, and the steel will have little effect on the initial stiffness of the member compared to the net area of concrete. This is further confirmed if you compare the experimentally adjusted values for grade 685 steel i.e. grade 685 a to the other grades. Therefore differences in the stiffness values can be attributed mainly to variations in the concrete mix and/or experimental errors. The following calculated experimental stiffness values at maximum capacity of the section, and at the yield and ultimate limit states are average values for each combination of steel and concrete normalised to the initial stiffness values derived above. This is to allow for direct comparison between the models that have varying material and sectional properties.

The yield stiffness values range between 18 and 41% of the initial stiffness. The stiffness values obviously increase with concrete strength. This increase is on average approximately 5% for an increase in concrete strength from grade 70 to grade 100, and a further 5% for grade 130 concrete from 100. The lowest grade of reinforcement has the highest normalised stiffness at yield. This is most likely due to the fact of the definition of the yield limit state, i.e. the delay in the yield of the higher grade bars at a greater displacement results in lower stiffness. The stiffness coefficients for the higher grades of reinforcement are approximately equal within each grade of concrete strength. The variation is between 2 and 6 % depending on the concrete. The slightly higher results for the grade 685 steel may be due to the larger diameter, although they only decrease slightly when adjusted for the lower diameter. However there is a limitation on the validity of the adjusted values in that the analytical model will always assume perfect bond between the steel and concrete. (iii) Stiffness at Maximum Capacity

The stiffness values range between 18 and 29% of the initial stiffness with again the values increasing with a rise in the concrete strength. This rise has a maximum value of approximately 5% for each jump in the concrete grade. Comparing values at yield and maximum limits, for steel grades higher than 500, the stiffness changes very little, up to a maximum of 6%. This again highlights the fact at the higher grades of steel produce almost balanced reinforced section conditions. This however delays the formation of the plastic hinge. The fall in the stiffness for models with the grade 500 steel is 7 % 10% and 14% with grades 70, 100 and 130 concrete respectively. (iv) Stiffness at 15% Drop in Maximum Capacity

The average stiffness values at 15% drop in the maximum capacity normalised to the yield stiffness are given in Table 5.8. The stiffness values range between 7 and 18% of the initial stiffness. There is again a steady increase in the stiffness for higher values of concrete compressive strength to a maximum of 5% for each increase in the concrete grade. The higher grades of steel display lower coefficients than the grade 500 steel. By reference to the envelope curves of horizontal load versus tip displacement, this is due to the larger displacement of the members with higher grade steel at this limit state. This could be a function of both the superior stress ratio of the higher grade steels and the delay in the attainment of yield strain.


27

For the higher grade steels, the concrete strength rather than the yield strength of the steel has the major effect on the stiffness coefficient. (v)Stiffness at 3% Drift

The yield stiffness values range between 13 and 20% of the initial stiffness. Within each level of steel grade, the stiffness is virtually constant for each value of concrete compressive strength. The rise is a maximum of 2% for each increase in concrete grade. By comparing the relevant values of Tables 5.7 and 5.8 for each combination of steel and concrete, the stiffness values at both ultimate limit states for the grade 500 steel with the grade 70 and 100 MPa concrete are approximately equal. For grade 500 steel with 130 MPa concrete, the drift 3%drift coefficient is lower and hence the 15% drop in maximum capacity will be the governing ultimate limit state. This is due to the higher value of applied axial load and hence greater P- effect. In contrast to the above, for the remaining grades of steel with each value of concrete compressive strength, the 3% drift value will be the ultimate limit state which is achieved first. As the axial load level within each combination is equivalent, this effect may be as a product of both the increased yield strength of the steel and the superior steel stress ratio.

5.7 PLASTIC HINGE LENGTH

The plastic hinge length defined as the portion of the member where the section yield capacity has been exceeded may be calculated in two different ways via external loads or internal stains. The results of the calculation of both methods are normalised to the length of the member.

For the external forces, the yield moment is calculated at the attainment of yield strain in the longitudinal bars. The maximum moment capacity of the section is calculated from the resolution of the horizontal and vertical jack loads for each value of displacement. The maximum moment is then plotted over the height of the model and the point where the yield moment is achieved is deemed to be the extent of the plastic hinge. The main drawback to this method is in cases where a balanced reinforced section occurs, by calculation the plastic hinge length would be zero.

By considering internal strains, the gauge values for the longitudinal steel within the critical section of the member are used to derive curvature values. Again yield curvature is taken as the section curvature at the initial yield of the longitudinal bars. At higher displacements of the model, the curvature is calculated over its height via the strain gauge readings on the longitudinal bars. The plastic hinge length is deemed to be the section of the member where the yield curvature is exceeded. However due to the problems of recording strains especially when the yield value of the steel have been exceeded this method may not be very accurate.

5.8 DISPLACEMENT DUCTILITY

The values of displacement ductility range from 1.2 to 2.6. At the 3% drift limit there is little variation in the displacement ductility within each grade of reinforcement, a maximum of 10%.


28

The full ductility display reasonable scatter confirm that for the ultimate limit state of 15 % drop in the maximum capacity, the ductility of the models decrease with an increase in the concrete compressive strength. This is the normal trend expected when considering the effects of concrete strength alone. Contrary to the previous observation, when the ultimate limit condition is set at 3% drift, for each level of reinforcement, the ductility fluctuates with an increase in the compressive strength. Comparing the values of ductility for both ultimate limit states, as the concrete and steel strength increase, the 3% drift limit is generally satisfied before the 15% drop is met. This is due to the delay in reaching the yield strength of the steel for the higher steel grades and to the presence of relatively low axial loads. Consequently the steel will be strain hardening after the cover has spalled and the P- effect of the axial load will be minimal. The steel material properties have a major influence on the displacement ductility. As illustrated in Figure C.8, the ductility decrease with an increase in both the longitudinal steel yield and steel stress ratio for each category of concrete strength studied. Whereas an increase in the yield strength of the member should cause a decrease in the ductility due to the delay in attaining the yield in the reinforcement, an increase in the steel stress ratio should raise the ductility of a member. However this depends on the full level of strain hardening being achieved within the model that will be compounded by the initial delay in the yield of the steel member.

5.9 CURVATURE DUCTILITY

The values of curvature ductility range from 1.0 to 4.3, although the higher values only occur in isolated cases. The average values for each combination of steel and concrete are in the region of 1.12 and 2.68, which are more representative of the calculated values. The values of curvature ductility calculated are approximately equal to those for displacement ductility, this immediately questions the validity of the former as they should be much greater than latter especially for large displacements coupled with short plastic hinge lengths as discussed earlier. The full ductility displays large scatter in the results especially for the ultimate limit state of 15 % drop in the maximum capacity. For an increase in the concrete strength, the ductility of the models with grade 500 and 900 steel is approximately constant, whereas for the other steel grades the ductility rises slightly (0.5 to 1.0).with the concrete strength. Comparing the values of ductility for both ultimate limit states, as the concrete and steel strength increase, the 3% drift limit is generally satisfied before the 15% drop is met. This is due to the delay in reaching the yield strength of the steel for the higher steel grades and to the presence of relatively low axial loads. Consequently the steel will be strain hardening after the cover has spalled and the P- effect of the axial load will be minimal. Depending on which ultimate limit state is selected, there are conflicting trends on the influence of the steel yield strength the ductility. However as stated above, the 3% drift limit governs the member behavior, considering this alone an increase in the longitudinal steel yield reduces the available ductility of the member for each grade of concrete. However for 130 MPa concrete the ductility is approximately constant.


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An increase in the steel stress ratio should raise the ductility of a member but this is not confirmed by the results. However this depends on the full level of strain hardening being achieved within the model. In conclusion on the presentation of the results for the stiffness, plastic hinge lengths and ductility calculations, there are a number of contradictions due to the differences in the definitions of the ultimate-limit-state and the constituent material properties of the models. As a consequence the experimental results need to be mirrored and extended further by parametric study, in order to isolate and assess the effects they have on the member behavior.

5.10 AXIAL LOAD

It is clear that the spread on cracking along the members and the amount of spalling of the cover concrete increases with the applied axial loads. Further it was observed from the tests that the spalling was more sudden and explosive when higher loads applied. Models 91 and 92 are compared with model 31, and separately model 93 is compared to model 35. This is due to the fact that although all the models have the same longitudinal steel and concrete, the level of confining steel is different. Models 31, 91 and 92 contain grade 500 links at 35mm centres whereas, models 35 and 93 have grade 1300 links at 50mm centres. The full base shear versus horizontal tip displacement up to and including the 60mm horizontal displacement cycles are plotted in Figure 7. Study of the envelope curves illustrates clearly the effects of the axial load on the overall behavior of the beam-column members. As the axial load increases, the initial stiffness increases along with the base shear capacity of the section. Both these effects are attributed to the requirement of a larger horizontal force to overcome the presence of the increased axial load. Once the maximum base shear is reached, the descending branch of the load displacement curves are steeper for the higher axial loads as the greater P-D effects leads to a faster degradation of the member strength. This has obvious effects on reducing the attainable plastic hinge length and the ductility supply of the member. The following is a summary of the experimental observations of the effect of the axial load on the member stiffness

There is a general increase in the initial and yield stiffness values as well as that at maximum

capacity of the members with an increase in the axial load applied. This is due to the fact that the presence of a greater axial load requires initially larger horizontal loads in order to overcome the effective pre-compression on the member section.

The initial stiffness rises by 12% for the increase in axial load from 10 to 15% and a further 18% from 15 to 20% axial load.

The rise in the stiffness value at yield and maximum capacity are not so easily defined from the test results but overall the yield stiffness increases on average by 8% when the axial load changes from 10 to 20%. Likewise the stiffness at maximum capacity increases by an average of 10%.

At the ultimate limit states of 15% post-max drop and 3% drift there are few definitive trends in the actual stiffness values, and generally the results show little variation. This is attributed to the P- effects of the axial loads in addition to the horizontally applied loads which will decrease as the axial load increases.


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It is concluded form the above that as the axial load increases so does the initial stiffness but the relative normalised stiffness values at each limit state considered decreases. The results for the calculation of the plastic hinge length from both methods clearly show that with an increase in the axial load there is a definite decrease in the attainable plastic hinge length. This is due to the fact that higher axial load will restrict the spread of yield of the reinforcement along the length of the member. The plastic hinge length is reduced by an average of of 5 to 7% when the axial load increases from 10 to 15% but may be reduced by up to 20% when the load increases from 10 to 20%. This latter increase will have a large effect on localising the plastic hinge within a member, hence restricting its ductility supply. Partially ignoring the low value of displacement ductility of the beam-column member subjected to 15% axial load, Table 5.18 indicated that the values of displacement ductility are approximately constant for both ultimate limit states. There is a decrease in displacement ductility with and increase in the axial load applied of up to 11% when the load increases from 10 to 20%. However, there is no measured difference in the ductility supply between the 15 and 20% applied load cases suggesting that his may represent a upper limit on the potential displacement ductility capacity of the member. This along with relationships for stiffness, plastic hinge length and curvature ductility are investigated further in the parametric study described in Chapter 6 of this thesis.

6. CONCLUSION The flexural ductility of high strength reinforced concrete members has been investigated experimentally on the material and member levels. An extensive experimental program involving primarily cyclic testing of beam-column members under constant axial load was undertaken. In addition testing of the constituent materials, i.e. concrete under uniaxial and confined conditions, and reinforcing steel under monotonic and cyclic loading was carried out. The following is a summary of the main aspects of the research work and the conclusions of this paper.

It is possible to produce concrete of compressive cube strength up to 130 MPa under laboratory conditions with techniques that can largely be reproduced on site.

The primary drawback in the employment of high strength concrete within ductile structures is the inherent brittleness of the concrete as the compressive strength rises. The enhancement of concrete ductility by use of large levels of confining reinforcement was demonstrated with limits proposed on the effective yield strength of the steel.

For unconfined concrete, after attainment of the maximum strength and corresponding strain, the descending branch of the stress-strain curve is virtually impossible to achieve due to the sudden and brittle failure of the concrete. This brittleness increases with concrete strength and only a small portion of the descending branch was achievable for the higher concrete strengths.

Triaxial testing of concrete cylinders showed that with sufficient confining levels, it is possible to produce large ductile responses from high strength concrete. However, these


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confining stresses were active hydraulic rather than passive pressures provided by reinforcing steels.

Five grades of steel having yield strengths between 540 and 1200 MPa were procured for the beam-column test members. Four of these steels possessed profiles similar to the conventional hot-rolled deformed type reinforcement, whilst the remaining grade 1300 steel was produced with a continuous spiral groove along its length.

Each grade of reinforcement was tested under direct tensile loads to failure. The elongation

measured at fracture of the reinforcement showed reasonable values of approximately between 6 and 8%.

The derived stress-strain relationship for the grade 785 steel displays a ‘round house’

behavior to which a definition of yield was required. The established method of using a tangent drawn through a point at 0.2% strain was deemed to be unrepresentative for this steel type and a new definition proposed. The remaining steel grades all had definitive yield points although the grade 1300 steel did not have a yield plateau. The stress ratio values for the steel (ultimate to yield stress ) were experimentally derived as between 1.13 and 1.27.

An attempt was made to experimentally derive the cyclic curves of samples of the

longitudinal steel bars. A purpose made testing apparatus was developed with the main emphasis placed on the end anchorage of the specimens. The samples were tested under strain control via gauges.

An approximation of the average curvature ductility within the plastic hinge zone of the

member derived from the previously calculated plastic hinge lengths and displacement yielded values of 4.0 to 11.2 that are more representative. Generally, ductility is adversely affected by the concrete strength and steel yield but the trends are not consistent. The combined effects of the steel yield and the stress ratio values govern member behavior and must be considered in isolation to derive their individual influences.

A general increase in the longitudinal steel grade produces a steeper ascending branch and

shallower descending branch for equivalent concrete grades. Alternatively, an increase in the concrete strength leads to both steeper ascending and descending branches.

Acknowledgment


32

References

1 Clayton N., 1993. ‘High Grade Concrete - Strength Results from Specimens up to C130

Grade and Age 1 Year.’ Building Research Establishment, Note N139/93 2 Chin M.S., Mansur M.A. and Wee T.H., 1997. ‘Effects of Shape, Size, and Casting direction

of Specimens on the Stress-Strain Curves of High Strength Concrete.’ ACI Materials Journal, 94-M24: 209-219

3 CEB-FIP Working Group On HSC/HPC, 1995. ‘High Performance Concrete -

Recommended Extensions to the Model Code 90, Research Needs.’ CEB Bulletin d’Information, No. 228

4 CEB-FIP Working Group On HSC/HPC, 1990. ‘High Strength Concrete - State-of-the-Art

Report.’ CEB-FIP Bulletin d’Information No. 197 5 BS8110, 1997. ‘Structural Use of Concrete: Code of practice for design and construction.’

British Standards Institution, London. 6 CEB-FIP, 1993. ‘Model Code 1990.’ Thomas Telford, London. 7 ACI Committee 318, 1995. ‘Building Code Requirements for Reinforced Concrete’

American Concrete Institute Detroit. 8 Jaccoud J-P., Farra B. and Leclercq A., 1996. ‘Improvement of Existing Codes for Their

Application to Crack Control of HSC/HPC Structures.’ Fourth International Symposium on Utilisation of High Strength/High Performance Concrete, Paris.

9 Jaccoud J-P. and Leclercq A., 1995. ‘Some Aspects Concerning Extension of Present Design

Rules to HPC-Structures.’ Fourth Weimar Workshop on ‘High Performance Concrete: Material Properties and Design.’ Weimar, Germany

10 Ahmad S.H. and Shah S.P., 1982. ‘Stress-Strain Curves for Concrete Confined by Spiral

Reinforcement.’ ACI Journal, 79-46: 484-490. 11 Ahmad S.H. and Shah S.P., 1982. ‘Stress-Strain Curves for Concrete Confined by Spiral

Reinforcement.’ ACI Journal, 79-46: 484-490. 12 Al-Hussaini A., Regan P.E., Xue H-Y. and Ramdane K-E., 1993. ‘The Behaviour of High

Strength Concrete Columns under Axial Load.’ Third International Conference on Utilisation of High Strength Concrete. Lillehammer Norway, 83-90.

13 Almusallam T.H. and Alsayed S.H., 1995. ‘Stress-Strain Relationship of Normal High-

Strength and Lightweight Concrete. Magazine of Concrete Research, 47-170: 39-44. 14 Carrasquillo R.L., Nilson A.H. and Slate F.O., 1981. ‘Properties of High Strength Concrete

Subjected to Short-Term Loads.’ ACI Journal, 78-14: 171-178 15 Muguruma H., Watanabe F. and Nishiyama M., 1988. ‘Curvature Ductility Design of

Reinforced and Prestressed concrete Members.’ Ninth World Conference on Earthquake Engineering, l4: 407-412.


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16 Muguruma H. and Watanabe F., 1990. ‘Ductility Improvements of High-Strength Concrete

Columns with Lateral Confinement.’ High Strength Concrete - Second International Symposium, 121-4: 47-60.

17 Muguruma H., Nishiyama M. and Watanabe F., 1993. ‘Stress-Strain Curve Model for

Concrete with a Wide Range of Compressive Strength.’ Third International Conference on Utilisation of High Strength Concrete. Lillehammer Norway, 314-321.

18 CEB-FIP, 1993. ‘Model Code 1990.’ Thomas Telford, London. 19 CEB-FIP Working Group On HSC/HPC, 1995. ‘High Performance Concrete -


20 ACI Committee 318, 1995. ‘Building Code Requirements for Reinforced Concrete’

American Concrete Institute Detroit. 21 Aoyama H., Murota T., Hiraishi H. and Bessho S., 1990. ‘Outline of Japanese National

Project on Advanced Reinforced Concrete Buildings with High-Strength and High-Quality Materials.’ High Strength Concrete - Second International Symposium, ACI SP-121-2: 21-31.

22 Morita S. and Shiohara H., 1996. ‘Development of High Strength Mild Steel Deformed Bars

for High Performance Reinforced Concrete Structural Members.’ Eleventh World Conference on Earthquake Engineering, 1742.


for High Performance Reinforced Concrete Structural Members.’ Eleventh World Conference on Earthquake Engineering, 1742

24 Pipa M. and Carvalho E.C., 1994. ‘Reinforcing Steel Characteristics for Earthquake

Resistant Structures.’ Tenth European Conference on Earthquake Engineering, 2887-2892. 25 Eurocode 8, 1995. ‘Earthquake Resistant Design of Structures ENV 1998’, European

Committee for Standardisation 26 Morita S. and Shiohara H., 1996. ‘Development of High Strength Mild Steel Deformed Bars

for High Performance Reinforced Concrete Structural Members.’ Eleventh World Conference on Earthquake Engineering, 1742.

27 Eurocode 8, 1995. ‘Earthquake Resistant Design of Structures ENV 1998’, European

Committee for Standardisation 28 Pipa M., Carvalho E.C. and Otes A., 1994. ‘Experimental Behaviour of Reinforced Concrete

Beams with Grade 500 Steel.’ Tenth European Conference on Earthquake Engineering, 2405-2410.


for High Performance Reinforced Concrete Structural Members.’ Eleventh World on Earthquake Engineering, 1742


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30 Monti G. and Nuti C., 1992. ‘Nonlinear Cyclic Behaviour of Reinforcing Bars Including

Buckling.’ ASCE Journal of Structural Engineering, 118-12: 3268-3284. 31 Albanesi S. and Biondi S., 1994. ‘On Longitudinal Reinforcement Buckling in Reinforced

Concrete Members Subjected to Strong Inelastic Loads.’ Tenth European Conference on Earthquake Engineering, 1613-1618.

32 Clayton N., 1993. ‘High Grade Concrete - Strength Results from Specimens up to C130

Grade and Age 1 Year.’ Building Research Establishment, Note N139/93 33 BS8110, 1997. ‘Structural Use of Concrete: Code of practice for design and construction.’

British Standards Institution, London. 34 CEB-FIP Working Group On HSC/HPC, 1995. ‘High Performance Concrete -


35 Park R., ‘Application of the New Zealand Concrete Design Code NZS 3101 to High

Strength Concrete.’ 36 Bertero V.V., 1996. ‘State-of-the-Art Report on: Design Criteria.’ Eleventh World

Conference on Earthquake Engineering, 2005. 37 Bertero V.V., 1996. ‘State-of-the-Art Report on: Design Criteria.’ Eleventh World

Conference on Earthquake Engineering, 2005.

Table of Contents

1. Introduction ................................................................................................................................ 2 2. Literature Review ..................................................................... Error! Bookmark not defined.

2.1 High Strength Concrete .................................................................................................. 4 2.2 High Yield Reinforcement ............................................................................................. 7

2.1 Types of Reinforcement ............................................................................................. 7 2.2.2 Mechanical Properties ............................................................................................ 7

3 Experimental Setup .............................................................................................................. 10 3.1 Adopted Mix designs for High Strength Concrete ............................................................. 10 3.2 Types of High Strength Steel Utilized ............................................................................... 13

4. Execution of the Member Tests ............................................................................................... 16 4.1 Test Procedure And Programme .................................................................................. 16 4.2 Beam-Column Specimen .............................................................................................. 18 4.3 Experimental Set-up ..................................................................................................... 18

4.3.1 Test Rig ................................................................................................................ 18 4.3.2 Hydraulic Power System ............................................................................................. 19 4.3.3 Electro-Hydraulic Servo-Control ............................... Error! Bookmark not defined.


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4.3.4 Test Software ........................................................ Error! Bookmark not defined. 4.3.5 Data Acquisition ................................................... Error! Bookmark not defined.

5. Experimental Results ................................................................................................................ 20 5.1 Damage Patterns ........................................................................................................... 23 5.2 Horizontal Load-Displacement Response .................................................................... 23 5.3 Longitudinal Steel Strains ................................................................................................. 24 5.4 Confining Strains .......................................................................................................... 24 5.5 Yield and Ultimate Limit States ........................................................................................ 25 5.6 Stiffness ............................................................................................................................. 25 5.7 Plastic Hinge Length ......................................................................................................... 27 5.8 Displacement Ductility ...................................................................................................... 27 5.9 Curvature Ductility ............................................................................................................ 28 5.10 Axial Load ........................................................................................................................ 29

6. Conclusion ............................................................................................................................ 30 List of Figures .............................................................................................................................. 35

List of Tables Table 1 : Definition of concrete classification ............................................................................... 2 Table 2 : Tensile strengths (MPa) for various concrete grades by Jaccoud, Farra & Leclercq ... 5 Table 3: Tangent moduli for various concrete grades Jaccoud, Farra & Leclercq ..................... 5 Table 4: Proposed properties of high strength reinforcement from Japanese 'New RC' project. . 7 Table 5: Requirements for reinforcing steel in critical regions from Eurocode 8. ....................... 8 Table 6: Details of experimental concrete mix designs. ............................................................. 10 Table 7: Compressive strength test results from cylinder and cube specimens (MPa) ............... 11 Table 8: Experimental maximum stress and corresponding strain results for confined concrete

cylinders for each grade of concrete subjected to 5, 10 and 15 MPa confinement ............. 12 Table 9: Nominal Steel Reinforcement Characteristics (values supplied by fabricators). ......... 13 Table 10: Experimental Characteristic Properties of Longitudinal Reinforcement ................... 14 Table 11: Comparison of Actual-to-Nominal Characteristic Properties of Longitudinal

Reinforcement ....................................................................................................................... 15

List of Figures

Figure 1: Curve showing stress-strain relationships for unconfined concrete according to model

code 90 and the recommendations of CEB-FIP Bulletin No. ................................................ 6 Figure 2: Stress-strain relationship of high strength reinforcement for the Japanese ‘New RC’

project reproduced from Morita & Shiohara,1996 ................................................................ 8


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Figure 3: Experimental stress-strain results for unconfined concrete cylinders ........................ 11 Figure 4: Monotonic Tests Results for Longitudinal Reinforcement ........................................... 14 Figure 5 : Experimental Results of Cyclic Tests of Longitudinal Reinforcement ........................ 16

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