ONG Chee Wee-PhD Thesis-2009-Centrifuge Model Study of Tunnel_Soil_Pile Interaction in Soft Clay.pdf

CENTRIFUGE MODEL STUDY OF TUNNEL-SOIL-PILE INTERACTION

IN SOFT CLAY

ONG CHEE WEE

NATIONAL UNIVERSITY OF SINGAPORE

2009

CENTRIFUGE MODEL STUDY OF TUNNEL-SOIL-PILE INTERACTION

IN SOFT CLAY

ONG CHEE WEE (B.Eng.(Hons),UPM, M.Sc.(Civil), NUS, P.Eng.)

A THESIS SUBMITTED

FOR THE DEGREE OF DOCTOR OF PHILOSOPHY

DEPARTMENT OF CIVIL ENGINEERING

NATIONAL UNIVERSITY OF SINGAPORE

2009

i

DEDICATION

To my dearest parents, my caring wife and my lovely twins……

ii

ACKNOWLEDGEMENTS

It was been sweat and tears the past four years to complete my PhD and coupled with countless trials and errors; the conclusion of this thesis has finally arrived! Foremost, I would like to extend my heartfelt gratitude and thanks to a special and wonderful person, none other than my main supervisor, Prof. Leung Chun Fai. To my co-supervisor, Prof. Yong Kwet Yew, who had throughout the research period, gave me constructive comments which have stimulated the success of this thesis. I would like to thank them for their willingness to share their vast experience and guidance. In addition to this, my appreciation goes out to Prof. Chow Yean Khow for his advice in the constant students group discussion held fortnightly. Gratitude also goes out to Prof Tan Thiam Soon and A/Prof. Harry Tan Siew Ann for their advices during my qualifying examination which has propelled me to move forward towards a clearer and better direction. Prof. Tan always encouraged his students to think out of the box. One day, when I was exhausted in innovating my new centrifuge model tunnel (see Figure 3.6), I have suggested the owner of our engineering canteen, Seton Lin to turn over my model tunnel and use it as cup holder (as shown in the Figure 1 below). With this ‘turned-over’ tunnel, you can save your queuing time next time when you buy a cup of coffee or tea from NUS canteen. The innovative centrifuge model tunnel is also part of the contributions from Mr. Wong C.Y, Dr. Okky Purwana, Dr. Shen Rui Fu, Mr. Martin Loh and Ian Kit.

Figure 1 The ‘turned-over’ model tunnel is now a cup holder in NUS Engineering canteen I also wish to express my sincere gratitude and thanks to Asst. Prof. Goh Siang Huat, who has been a source of endless ideas and inspiration, to Dr. Shen Ruifu, who has been advising me for my centrifuge model test and interpretation of my test results, and to Prof. K.K. Phoon, who has encouraged and supported me since I was reading my M.Sc. back in year 2002. I will never forget the wonderful time that I spent with him in organizing many conferences. Special thanks are also extended to the technical staffs of NUS Geotechnical Centrifuge Laboratory,Wong Chew Yuen, Mdm. Jamilah, Lye Heng, Shaja, John Choy, Loo Leong Huat and Adrian Tan. I would like to extend my sincere gratitude to the final year students that I had privilege of working with, i.e. Kai Yang (04/05), Eddie Hu (05/06), Ian Kit and Eliza (06/07) and Teng Hui (07/08) I also wish to credit the support of the following professionals, associates and friends for sharing their experiences and knowledge namely Prof. Andrew Palmer, Prof. I.H Wong, Dr. Dave White, Dr. Johny Cheuk, Mr. Nick Shirlaw, Mr. Ow Chun Nam, Dr.

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Jeyatharan Kumarasamy, Dr. Lin Kai Qiu, Mr. Cham Wee Meng, Mr. Lim Tuck Fang, Mr. Cheang Yew Kee, Mr. Poh Chee Kuan, Mr. Koo Chung Chong, Mr. Lee Hong Keow, Mr. Edmund Ng, Mr. Jonathan Ang, Mr. Phang Chu Mau, Mr. Chew Eik Khoon, Mr. Jimmy Chew and Mr. Lau Kim Hwa. I extend my appreciation to my many colleagues and friends who I have consulted during the course of the research, particularly Chin Hong, Ran Xia, Xie Yi, Yonggang, Xiying, San Chuin, Kheng Ghee, Kar Lu, Cheng Ti, Hung Leong, Czhia Yheaw, Haibo, Teik Lim, Okky, Dominic, William, Wang Lei, Chen Jian, Feijian, Sindhu, Ch’ng Yi, Poh Hai, Karma, Subhadeep, Khrishna, Jiangtao, Andy, Chong Hun & Ben. It was a memorable trip to Schofield Centre at Cambridge University in April/May 2007. Thank you for the kind arrangement of Prof. Robert Mair. My stay in Jesus College was wonderful and fulfilling. I would also like to thank Prof. Malcolm Bolton who has provided arrays of resource and pertinent pointers in improving my model and research. To Prof. Andrew Noel Schofield, Prof. Kenichi Soga, Dr. Stuart Haigh, Dr. I Thusyanthan, Dr. Johnson Chung, Sidney Lam, Fiona Kwok, Gui Chang Shin, Tricia Lee, Alec Marshall, Richard Laver, Hisham Mohamad, Senthil, Christopher and Dr Yueyang Zhao, all for giving me valuable advices and encouragement on my research. Thank you Geotechnical Society of Singapore for the sponsorship to Bangalore, India to attend the 6th Asian Young Geotechnical Engineers Conference (Dec 2008) in which my paper co-authored with Prof. C. F. Leung and Prof. K. Y. Yong have won the ‘Best Contribution Award’. A special acknowledgement is dedicated to Wendy and Angelia for their help and assistance in the compilation of this thesis. I would like to extend my gratitude to my parents, my wife, Shin Inn and my twins, Yee Heng and Yee Huan (born one month after I pursued my PhD), my sisters, brother-in-laws, grandparents, uncles, aunts and cousins for their never-ending love, support, tolerance and sacrifice in encouraging me to complete this research. A special mention to my brother-in-law, Moong Khai Chee for his support, guidance and technical advice ever since my undergraduate years. It is also with great pride despite the hectic schedule of my PhD research, I have also successfully passed the qualifying exams and registered myself as a Professional Engineer with Board of Engineers Malaysia in 2008; passed the Professional Engineer Fundamentals Engineering Examination, Singapore in 2007, and being registered as a Resident Engineer with Building and Construction Authority, Singapore also in 2007. To the remaining people whom I am unable to list down, please accept my sincere appreciation and thanks for the feedback, assistance, tolerance and help rendered. Last but not least, deepest appreciation and thanks to National University of Singapore for the award of this research scholarship throughout the four years period for without which this research program would not have been possible. Ong Chee Wee 28-02-2009

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TABLE OF

CONTENTS

Dedication i

Acknowledgements ii

Table of Contents iv

Summary x

List of Tables xii

List of Figures xiii

Nomenclature xxvi

CHAPTER 1 INTRODUCTION 1

1.1 Background 1

1.2 Tunnelling-Induced Soil Movements 2

1.3 Effects of Tunnelling on Piles 3

1.4 Objective and Scope of Study 4

1.5 Structure of Thesis 5

CHAPTER 2 LITERATURE REVIEW 9

2.1 Introduction 9

2.2 Techniques for Simulation of Tunnelling in Centrifuge

9

2.2.1 Simulation Technique 1 - High Density Polystyrene Foam 11

v

2.2.2 Simulation Technique 2 - Compressed Air 12

2.2.3 Simulation Technique 3 - Liquid – Oil / Water 14

2.2.4 Simulation Technique 4 - Mechanical Equipments 16

2.3 Tunnelling-Induced Soil Movements 17

2.3.1 Field Studies of Tunnelling-Induced Soil Movement 17

2.3.2 Centrifuge Model Tests of Tunnelling-Induced Soil Movement 22

2.3.3 Predictive Methods of Tunnelling-Induced Soil Movement 24

2.4 Tunnelling-Induced Pile Responses

29

2.4.1 Field Studies of Tunnelling-Induced Pile Responses 29

2.4.2 Centrifuge Model Tests of Tunnelling-Induced Pile Responses 33

2.4.3 Predictive Methods of Tunnelling-Induced Pile Responses 36

2.5 Summary 39

CHAPTER 3 EXPERIMENTAL SET-UP AND PROCEDURE 67

3.1 Introduction 67

3.2 Geotechnical Centrifuge Modeling 67

3.2.1 Principles of Geotechnical Centrifuge Modelling 67

3.2.2 NUS Geotechnical Centrifuge Facility 69

3.3 Experimental Set-Up 70

3.3.1 Model Tunnelling Technique 70

3.3.1.1 Advantages of Model Tunnel 72

3.3.1.2 Limitations of Model Tunnel 73

3.3.2 Instrumented Model Piles 74

3.3.3 Model Pile Cap 76

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3.3.4 Strong Box 77

3.3.5 Kaolin Clay 77

3.3.6 Toyoura Sand 78

3.3.7 Potentiometer 78

3.3.8 Pore Pressure Transducers (PPT) 79

3.3.9 Non-Contact Laser Transducers 80

3.4 Image Acquisition System 80

3.4.1 High Resolution Camera 80

3.4.2 Lighting System 81

3.4.3 On-Board and Command Computers 81

3.4.4 Post-Processing of Images 82

3.4.5 Assessment of Effectiveness of Image Processing System 83

3.5 Experimental Procedure

84

3.5.1 Preparation of The Soil Sample 84

3.5.2 Pre-Consolidation Process 85

3.5.3 Installation of Model Tunnel and PPTs At 1g 86

3.5.4 Preparation Works for PIV Analysis 87

3.5.5 Installation of Model Pile at 1g 87

3.5.6 Test Procedure 88

CHAPTER 4 BASIC TESTS ON VOLUME LOSS 102

4.1 Introduction 102

4.2 Test Program 102

4.3 Tunnelling-Induced Soil Movements (Tests 1, 2) 104

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4.3.1 Cumulative Soil Movements 104

4.3.2 Soil Surface Settlement Troughs 105

4.3.3 Subsurface Vertical Soil Movements 107

4.3.4 Subsurface Horizontal Soil Movements 110

4.3.5 Qualitative Assessment On Excess Pore Pressure Response 111

4.4 Typical Tunnel-Soil-Piles Interactions (Test 3) 112

4.4.1 Induced Axial Force and Settlement 113

4.4.2 Induced Bending Moment and Deflection 115

4.5 Test Series 1- Effects of Volume Loss (Tests 3, 4)

118



4.6 Concluding Remarks

122

4.6.1 Tunnelling-Induced Soil Movements 122

4.6.2 Tunnel-Soil-Piles Interaction 123

CHAPTER 5 EFFECTS OF TUNNELLING ON SINGLE PILES 151


5.2 Test Series 2- Effects of Pile Tip & Head Conditions (Tests 3, 9, 10, 13)

151

5.2.1 Effects of Pile Tip Conditions 151

5.2.2 Effects of Pile Head Conditions 154

5.3 Test Series 3 - Effects of Pile Length (Tests 3, 7, 8) 157

5.4 Effects Of Distance of Pile From Tunnel 162

5.4.1 Test Series 4 - Free-Head Floating Piles (Tests 3, 5, 16, 6) 162

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5.4.2 Test Series 5 - Free-Head End-Bearing Piles (Tests 10, 11, 12) 165

5.4.3 Test Series 6 - Fixed-Head End-Bearing Piles (Tests 13, 14A, 14B) 166

5.4.4 Comparison of Results from Test Series 4, 5 And 6 167

5.5 Effects of Time on Pile Responses in Soft Clay 171

5.6 Comparison of Soil and Single Pile Behaviours due to Inward and Outward Tunnel Deformations

172

5.6.1 Tunnel-Soil Interaction 172

5.6.1.1 Similarities (Tunnel-Soil Interaction) 173

5.6.1.2 Differences (Tunnel-Soil Interaction) 174

5.6.2 Tunnel-Pile Interaction 174

5.6.2.1 Similarities (Tunnel-Pile Interaction) 175

5.6.2.2 Differences (Tunnel-Pile Interaction) 175

5.7 Concluding Remarks 177

CHAPTER 6 EFFECTS OF TUNNELLING ON PILE GROUPS 208


6.2 Floating Pile Group 209



6.3 End-Bearing Pile Group 215

6.3.1 Capped-Head 215

6.3.2 Fixed-Head 218

6.4 Pile Group Size 220

6.4.1 Capped-Head 220

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6.4.2 Fixed-Head 225

6.5 Pile Cap Conditions 227

6.5.1 2-Pile Group 227

6.5.2 6-Pile Group 229


CHAPTER 7 CONCLUSIONS 281


7.1.1 Technique for Simulation of Tunnelling 282

7.1.2 Tunnel-Soil Interaction 283

7.1.3 Tunnel-Single Piles Interaction 283

7.1.4 Tunnel-Pile Groups Interaction 286

7.2 Recommendations for Future Studies 288

REFERENCES 290

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SUMMARY

Tunnels are often constructed close to existing pile foundation in dense urban areas. It

is challenging to carry out extensive instrumentation and monitoring in the field to

observe the pile responses due to tunneling activities. Hence, centrifuge modelling

emerges as an attractive alternative option to investigate the effects of tunnelling-

induced soil movement on adjacent piles. In the present study, a modeling technique

was developed to simulate the inward tunnel deformation due to over-excavation of

tunnel.

Phase 1 of the study was performed to investigate the free-field soil movements due to

tunneling. It is found that the surface settlement trough in clay generally follows the

Gaussian distribution curve in the short-term. The magnitude of maximum ground

surface settlement increases with time and tunnel volume loss. Though the settlement

magnitude is larger in the long-term, the settlement trough is wider and hence the

differential settlement at the ground surface is not as serious as compared to that in the

short-term. The data confirmed that the empirical equation proposed by Mair et al

(1993) is applicable in the prediction of the subsurface settlement troughs in clay in the

short-term. In the short-term, an “Immediate Shear Zone” with large soil movement

above the tunnel can be identified, while the zone outside the immediate shear zone is

identified as “Support Zone”. In the long-term, soil settlement is noted to be more

dominant than lateral soil movement.

Phase 2 of the study was conducted to study the tunnelling-induced single pile lateral

and axial responses in both short- and long-term. The effects of factors such as volume

loss, pile tip and head condition, pile length and pile-to-tunnel distance were examined.

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It was found that a floating pile is mainly governed by pile settlement and a socketed

pile is likely governed by its material stress when tunneling was carried out adjacent to

it. It was noted that tensile force and relatively large negative bending moments are

induced at the pile head due to total fixity resulting in a reduction in drag load, and

positive bending moment at the mid-pile shaft.

The centrifuge model study was subsequently extended to pile groups to evaluate the

effects of number of piles, pile cap and pile tip condition. It was found that the pile

group is generally beneficial as the average pile responses of a pile group due to

tunneling are smaller than the average of those of single piles at the same locations.

The pile-cap-pile interaction in a capped-head 6-pile group would moderate the

induced pile bending moments among the piles within a pile group. It was noted that

the induced pile bending moments in the middle row is smaller than that of rear row

which is contrary to the induced lateral soil movements at the respective locations. For

the piles in a totally fixed-head 6-pile group, the piles behaved like single piles

standing side by side in terms of axial force and bending moment, except that the

magnitude is affected by the total number of piles in the group.

A common trend was observed for the long-term over short-term ratio of pile

responses for both single pile and pile group. The results reveal that soil and pile

responses increase over time with long-term over short-term pile responses ratio

ranging from 1.32 to 2.4, regardless of pile size, pile head and toe conditions.

Keywords: tunnel; pile; interaction; axial force; settlement; bending moment;

deflection; clay; centrifuge modelling

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LIST OF TABLES

Table 3.1 Scaling relation of centrifuge modeling (After Leung et at, 1991)

Table 3.2 Physical properties of Malaysian kaolin clay (After Goh, 2003)

Table 3.3 Physical properties of Toyoura sand (After Teh et. al, 2005)

Table 4.1 Test program and parameters for the basic tests on volume loss

Table 5.1 Test program and prototype parameters in Phase 2 study

Table 6.1 Test program and prototype parameters for pile group tests

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LIST OF FIGURES

Figure 1.1 Pile responses induced by tunnel construction: (a) Tunnelling under pile foundation, (b) Tunnelling adjacent to pile foundation.

Figure 1.2 Pile foundations supported existing buildings normally designed to resist compression load only.

Figure 2.1 Simulation technique of tunnelling using high density polystyrene foam (After Sharma et al., 2001 and Feng, 2003)

Figure 2.2 Simplified tunnel lining deformation with time by simulation technique of tunnelling using high density polystyrene foam (After Ran, 2004)

Figure 2.3 Simulation technique of tunnelling - applying compressed air (After Grant & Taylor, 2000)

Figure 2.4 Simulation technique of tunnelling - model tunnel infilled with oil (After Loganathan et al., 2000)

Figure 2.5 Simulation technique of tunnelling - model tunnel infilled with water (After Jacobsz, 2002)

Figure 2.6 Sand pouring in process during model preparation with model tunnel infilled with water (After Jacobsz, 2002)

Figure 2.7 Strong-box with two openings to fix the model tunnel in place (After Jacobsz, 2002)

Figure 2.8 Simulation technique of tunnelling - mechanical equipment of miniature shield tunneling machine (After Nomoto et al., 1994)

Figure 2.9 Simulation technique of tunnelling - mechanical equipment of shield model machine (After Yasuhiro et al., 1998)

Figure 2.10 Simulation technique of tunnelling - mechanical model tunnel used to simulate the tunnel volume loss by decreasing the diameter of model tunnel under 1g (After Lee and Yoo, 2006)

Figure 2.11 Gaussian curve approximating transverse surface settlement trough for MRT project C852, Singapore (After Cham, 2007)

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Figure 2.12 Definition of parameters controlling tunnelling-induced settlement trough ( After Standing & Burland, 2006)

Figure 2.13 Gaussian curve approximating transverse surface settlement trough (After Peck, 1969)

Figure 2.14 Variation in surface settlement trough width parameter with tunnel depth for tunnels in clay (After Lake et al., 1992)

Figure 2.15 Variation of trough width parameter K with depth for subsurface settlement profiles above tunnels in clay (After Mair et al., 1993)

Figure 2.16 Initial settlement trough at Grimsby increased from 1.5Z to a final value in excess of 4Z in long-term (After O’Reilly et al,1991)

Figure 2.17 The maximum surface settlements at Grimsby increased significantly over the time (After O’Reilly et al, 1991)

Figure 2.18 The ratio of the maximum immediate settlement to maximum long-term settlement for Shanghai Metro Tunnel No.2 (After Zhang et al., 2004)

Figure 2.19 Normalized post-construction surface settlement troughs due to consolidation of soft clay (After Shirlaw, 1995)

Figure 2.20 Estimated trend of excess pore pressure in normally consolidated clay surrounding the tunnel (After Schmidt, 1989)

Figure 2.21 Changes in Pore pressure for Shanghai Metro (After Schmidt, 1989)

Figure 2.22 Change in pore pressure measured at Thunder Bay Sewer Tunnel (Adapted from data in Ng et al, 1986) (After Shirlaw et al, 1994)

Figure 2.23 Transverse movements in Toulouse subway line B were significantly increased over time, but stabilized after 15 days of tunnel excavation (After Emeriault et al, 2005)

Figure 2.24 Horizontal soil movement for the Singapore’s effluent outfall pipeline in tunnel (After Balasubramanian, 1987)

Figure 2.25 Comparisons of surface settlement troughs in sand (After Feng, 2003) and clay (After Ran, 2004)

Figure 2.26 Ground surface settlement trough over time from a typical test (After Ran, 2004)

Figure 2.27 Normalized Vertical and horizontal soil movement profile at different subsurface elevations with best-fit curves: (a) 10mm below ground level; (b) 30mm below ground level; (c) 70mm below ground level; (d) 100mm below ground level; (After Grant and

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Taylor, 2000) Figure 2.28 Comparisons of measured surface settlement and analytical

solutions (After Loganathan et al., 2000)

Figure 2.29 Plastic deformation mechanism for tunnels in clay (After Osman et al., 2006a)

Figure 2.30 Definition of GAP parameter (After Lee et al., 1992)

Figure 2.31 Oval-shaped soil displacement around tunnel boundary (After Loganathan and Poulos, 1998)

Figure 2.32 Boundary conditions of prescribed displacement (After Park, 2005)

Figure 2.33 Transverse movements in Toulouse subway line B were significantly increased over time, but stabilized after 15 days of tunnel excavation (After Emeriault et al, 2005)

Figure 2.34 A piled bridge pier foundation assessed during the CTRL project (After Jacobsz et al., 2005)

Figure 2.35 Typical section and instrumentation layout for pile-tunnel interaction study for MRT North East Line Contract 704 in Singapore (After Pang et al.,2005)

Figure 2.36 Post-tunneling measurement of the development of axial force in pile P1 at Pier 20 showing the time-dependent behaviour of drag-load MRT North East Line Contract 704 in Singapore (After Pang et al., 2005)

Figure 2.37 Responses of pile foundation in terms of (a) axial force and (b)bending moment for MRT North East Line Contract 704 in Singapore (After Pang et al., 2005)

Figure 2.38 Illustration of positions of existing instrumented piles relative to tunnels for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore. (After Cham, 2007)

Figure 2.39 Configuration of centrifuge tests (After Loganathan et al., 2000)

Figure 2.40 Tunneling-induced pile bending moments (After Loganathan et al., 2000)

Figure 2.41 Tunneling-induced pile axial loads (After Loganathan et al., 2000)

Figure 2.42 Variations of (a) induced pile bending moment profiles and (b) induced pile lateral deflection profiles with time in typical test (After Ran, 2004)

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Figure 2.43 (a) Induced pile axial force profile and (b) pile settlement profile at 2 days in typical test (After Ran, 2004)

Figure 2.44 Zone of influence around tunnel in which potential for large pile settlements exists (After Jacobsz et al., 2005)

Figure 2.45 Settlement, rotation and load distribution on triple pile group (After Jacobsz et al., 2005)

Figure 2.46 Layout of basic problem (After Chen et al., 1999)

Figure 2.47 Tunneling-induced pile responses and Greenfield soil movement (After Chen et al., 1999)

Figure 2.48 Numerical analysis of pile-group responses due to tunnelling (After Loganathan et al., 2001)

Figure 2.49 Typical 3D finite elements mesh to back-analyse a case history on the response of pile foundation subjected to shield tunnelling (After Pang et al., 2005b)

Figure 2.50 Prediction of responses of pile foundation in terms of axial force and bending moment using 3D finite element analysis (After Pang et al., 2005b)

Figure 3.1 Schematic diagram of NUS geotechnical centrifuge

Figure 3.2 Photograph of NUS geotechnical centrifuge with the model package mounted on the platform

Figure 3.3 Sketch of a typical centrifuge model package (All dimensions in mm)

Figure 3.4 Photograph of a typical centrifuge model package

Figure 3.5 Longitudinal view of model tunnel set up

Figure 3.6 Cross-section of model tunnel

Figure 3.7 Instrumented model pile (All dimensions in mm)

Figure 3.8 Model pile caps

Figure 3.9 In-flight undrained shear strength of clay

Figure 3.10 Image acquisition system

Figure 3.11 On board set-up

Figure 3.12 Picture captured by JAI ©CV-A2 progressive scan camera for PIV analysis

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Figure 3.13 Image manipulation during PIV analysis. (After White et al., 2003)

Figure 3.14 Evaluation of displacement vector from correlation plane, Rn(s): (a)

correlation of Rn(s); (b) highest correlation peak (integer pixel); (c) sub-pixel interpolation using cubic fit over ± 1 pixel of integer correlation. (After White et al., 2003)

Figure 3.15 Experimental set-up for assessment of effectiveness of image processing system and comparison of performance of flocks and beads

Figure 3.16 Results of assessment of effectiveness of image processing system and comparison of performance of flocks and beads

Figure 3.17 Pore pressure dissipation and settlement during consolidation stage

Figure 3.18 Estimation of ultimate settlement by Asaoka’s method (1978)

Figure 3.19 Different colours of beads were randomly embedded on the surface of clay

Figure 3.20 The Perspex window is highly greased to ensure free movement of soil

Figure 3.21 Set-up of the entire model package in 1g (top view)

Figure 4.1 Schematic of viewing area in tunnel-soil interaction tests (all dimensions in mm)

Figure 4.2 Example of digital images taken during test for PIV analysis

Figure 4.3 (a) Vectors and contour plots of soil movements after 2 days (Test 1)

Figure 4.3 (b) Vectors and contour plots of soil movements after 180 days (Test 1)

Figure 4.3 (c) Vectors and contour plots of soil movements after 360 days (Test 1)

Figure 4.3 (d) Vectors and contour plots of soil movements after 720 days (Test 1)





Figure 4.5 Surface settlement troughs over time (Test 1)

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Figure 4.6 Surface settlement troughs over time (Test 2)

Figure 4.7 Maximum surface settlements over time (Tests 1 & 2)

Figure 4.8 Settlement troughs at surface, 4.3m and 9.3m depths (Test 1): (a) comparing with Mair et. al (1993) (b) comparing with Loganathan and Poulos (1998)

Figure 4.9 Settlement troughs at surface, 5m and 10.9m depths (Test 2): (a) comparing with Mair et. al (1993) (b) comparing with Loganathan and Poulos (1998)

Figure 4.10 Distribution of inflection point ‘i’ with depth in short- and long-term (Test 2)

Figure 4.11 Comparison of ratio of iLT/iST at different depths (Tests 1 & 2)

Figure 4.12 Horizontal soil movements at different distance from tunnel center-line at 2 and 720 days - Test 1


Figure 4.14 Pore pressure changes due to tunnelling (Test 1)

Figure 4.15 (a) Tunnelling-induced pile axial force (Test 3, 3% free-head floating long pile)

Figure 4.15 (b) Tunnelling-induced pile axial force (Pile BP1-G) for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore (After Cham, 2007)

Figure 4.16 Tunnelling-induced pile head settlement (Test 3) and observed free-field soil movement at pile location (Test 1, PIV)

Figure 4.17 Tunnelling-induced (a) maximum pile axial force (b) maximum pile head settlement and soil surface settlement (Test 1) (c) maximum pile bending moment (d) maximum pile head deflection and soil surface lateral movement (Test 1)

Figure 4.18 (a) Tunnelling-induced pile bending moment (Test 3, 3% free-head floating long pile)

Figure 4.18 (b) Tunnelling-induced pile bending moment (Pile BP2-E) for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore (After Cham, 2007)

Figure 4.19 Tunnelling-induced pile deflection (Test 3) and free-field lateral soil movement at pile location (Test 1)

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Figure 4.20 Variation of pile axial force with volume loss (Tests 3 and 4)

Figure 4.21 Variation of pile head settlement (Tests 3 and 4) and observed free-field soil movement at pile location (Tests 1 and 2) with volume loss

Figure 4.22 Variation of pile bending moment with volume loss (Tests 3 and 4)

Figure 4.23 Variation of pile deflection profiles (Tests 3 and 4) and observed free-field lateral soil movement at pile location (Tests 1 and 2) with volume loss

Figure 4.24 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with volume loss (Tests 3 and 4)

Figure 4.25 Long-term to short-term ratio of pile responses for different volume losses (Tests 3 and 4)

Figure 5.1 Pile base position investigated in the parametric studies (not to scale)

Figure 5.2 Variation of pile axial force with tip condition in (a) Short-term (b) Long-term (Tests 3, 9, 10 and 13)

Figure 5.3 Variation of pile bending moment with tip condition (a) Short-term (b) Long-term (Tests 3, 9, 10 and 13)

Figure 5.4 Variation of pile deflection with tip condition (a) Short-term (b) Long-term (Tests 3, 9, 10 and 13)

Figure 5.5 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with tip and head conditions (Tests 3, 9, 10 and 13)

Figure 5.6 Variation of pile axial force with pile length (Tests 3, 7 and 8)

Figure 5.7 Variation of pile head settlement and soil settlement profile (Test 1) with pile length (Tests 3, 7 and 8)

Figure 5.8 Variation of pile bending moment with pile length (Tests 3, 7 and 8)

Figure 5.9 Variation of pile head deflection and free- field lateral soil displacement (Test 1) with pile length (Tests 3, 7 and 8)

Figure 5.10 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with normalized pile length over tunnel depth (Tests 3, 7 and 8)

Figure 5.11 Short pile to long pile ratio of pile responses for different pile length over tunnel depth (Tests 3, 7 and 8)

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Figure 5.12 Variation of pile axial force with pile-to-tunnel distance for free-

head floating piles (Tests 3, 5 and 6)

Figure 5.13 Variation of pile head settlement for free-head floating piles (Tests 3, 5 and 6) and free-field soil settlement (Test 1) with pile-to-tunnel distance

Figure 5.14 Variation of pile bending moment for free-head floating piles with pile-to-tunnel distance (Tests 3, 5 and 6)

Figure 5.15 Variation of pile deflection for free-head floating piles (Tests 3, 5 and 6) and free-field lateral soil displacement profile (Test 1) with pile-to-tunnel distance

Figure 5.16 Variation of pile axial force for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)

Figure 5.17 Variation of pile bending moment for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)

Figure 5.18 Variation of pile deflection for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)

Figure 5.19 Variation of pile bending moment for fixed-head end bearing piles with pile-to-tunnel distance (Tests 13, 14A and 14B)

Figure 5.20 Variation of pile deflection for fixed-head end bearing piles with pile-to-tunnel distance (Tests 13, 14A and 14B)

Figure 5.21 Variation of (a) maximum pile axial force (b) maximum pile head settlement and soil surface settlement (Test 1) (c) pile bending moment (d) maximum pile head deflection for Test Series 4, 5 and 6 with pile-to-tunnel distance

Figure 5.22 Assessment of pile responses for different pile-to-tunnel distance (Tests 3, 5 and 6)

Figure 5.23 Lateral soil displacement profiles at different pile-to-tunnel distance (Tests 3, 5 and 6)

Figure 5.24 Long-term to short-term ratio of pile responses for all tests (Tests 3, 4, 5, 6, 7, 8, 9, 10, 11, 12, 13, 14A, 14B and 16)

Figure 5.25 Comparison of (a) ovalisation of tunnel lining by Ran (2004); and (b) over-cut of tunnel in the present study

Figure 5.26 Simplified tunnel lining ovalisation with time (not to scale) (Test 1) (after Ran, 2004)

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Figure 5.27 Development of subsurface soil movements at (a) 2 days and (b) 720 days after tunnel excavation (after Ran, 2004)

Figure 5.28 Variation of surface soil settlement troughs with tunnel deformation

Figure 5.29 Variation of maximum surface soil settlement at tunnel central line with tunnel deformation

Figure 5.30 Variation of vertical soil settlement at pile location with tunnel deformation

Figure 5.31 Variation of soil deflection at pile location with tunnel deformation

Figure 5.32 Variation of pile axial force with tunnel deformation

Figure 5.33 Variation of pile head settlement with tunnel deformation

Figure 5.34 Variation of pile bending moment with tunnel deformation

Figure 5.35 Variation of tunnelling-induced maximum pile bending moment over time for different tunnel deformation

Figure 5.36 Variation of pile deflection with tunnel deformation

Figure 5.37 Long-term to short-term ratio of pile responses over time for different tunnel deformation

Figure 6.1 Tunnelling-induced pile axial force (Test PG1)

Figure 6.2 Tunnelling-induced front pile (Test PG1) and corresponding single pile (Test 3) axial force

Figure 6.3 Tunnelling-induced rear pile (Test PG1) and corresponding single pile (Test 16) axial force

Figure 6.4 Tunnelling-induced pile head settlement (Tests PG1, 3, 16)

Figure 6.5 Tunnelling-induced pile bending moment (Test PG1)

Figure 6.6 Tunnelling-induced front pile (Test PG1) and corresponding single pile (Test 3) bending moment

Figure 6.7 Tunnelling-induced rear pile (Test PG1) and corresponding single pile (Test 16) bending moment

Figure 6.8 Tunnelling-induced pile deflection (Tests PG1)

Figure 6.9 Tunnelling-induced front pile (Test PG1) and corresponding single pile (Test 3) deflection

xxii

Figure 6.10 Tunnelling-induced rear pile (Test PG1) and corresponding single pile (Test 16) deflection

Figure 6.11 Tunnelling-induced pile head deflection (Test PG1, 3 & 16)

Figure 6.12 Single pile over pile group ratio for front pile (Test 3/ PG1)

Figure 6.13 Single pile over pile group ratio for rear pile (Test 16/ PG1)



Figure 6.16 Tunnelling-induced rear pile (Test PG2) and corresponding single pile (Test 11) axial force



Figure 6.19 Tunnelling-induced rear pile (Test PG2) and corresponding single pile (Test 11) bending moment

Figure 6.20 Tunnelling-induced pile deflection (Test PG2)

Figure 6.21 Tunnelling-induced front pile (Test PG2) and corresponding single pile (Test 10) deflection


Figure 6.23 Tunnelling-induced pile head deflection in the (a) short-term (b) long-term (Tests PG2, 10 & 11)







xxiii

Figure 6.30 Tunnelling-induced rear pile (Test PG3) and corresponding single

pile (Test 14A) bending moment


Figure 6.32 Single pile over pile group ratio for rear pile (Test 14A/ PG3)


Figure 6.34 Tunnelling-induced front pile in 2-pile group (Test PG2) and corresponding front pile in 6-pile group (Test PG4) axial force

Figure 6.35 Tunnelling-induced rear pile in 2-pile group (Test PG2) and corresponding middle pile in 6-pile group (Test PG4) axial force

Figure 6.36 Tunnelling-induced pile bending moment (a) in the short-term (b) in the long-term (Test PG4)

Figure 6.37 Tunnelling-induced front pile in 2-pile group (Test PG2) and corresponding front pile in 6-pile group (Test PG4) bending moment

Figure 6.38 Tunnelling-induced rear pile in 2-pile group (Test PG2) and corresponding middle pile in 6-pile group (Test PG4) bending moment

Figure 6.39 Tunnelling-induced pile deflection in the (a) short-term (b) long-term (Test PG4)

Figure 6.40 Tunnelling-induced pile deflection (Tests PG2 and PG4)

Figure 6.41 Tunnelling-induced pile bending moment (Tests PG2 and PG4)

Figure 6.42 2-pile over 6-pile group ratio for front pile (Test PG2/PG4)

Figure 6.43 2-pile over 6-pile group ratio for middle pile (Test PG2/PG4)



Figure 6.46 Tunnelling-induced rear pile in 2-pile group (Test PG3) and corresponding middle pile in 6-pile group (Test PG5) axial force

Figure 6.47 Tunnelling-induced pile bending moment in the (a) short-term (b) long-term (Test PG5)


xxiv

Figure 6.49 Tunnelling-induced rear pile in 2-pile group (Test PG3) and

corresponding middle pile in 6-pile group (Test PG5) bending moment



Figure 6.52 Tunnelling-induced front pile in capped-head 2-pile group (Test PG2) and corresponding front pile in fixed-head 2-pile group (Test PG3) axial force

Figure 6.53 Tunnelling-induced rear pile in capped-head 2-pile group (Test PG2) and corresponding rear pile in fixed-head 2-pile group (Test PG3) axial force

Figure 6.54 Tunnelling-induced front pile in capped-head 2-pile group (Test PG2) and corresponding front pile in fixed-head 2-pile group (Test PG3)bending moment

Figure 6.55 Tunnelling-induced rear pile in capped-head 2-pile group (Test PG2) and corresponding rear pile in fixed-head 2-pile group (Test PG3) bending moment

Figure 6.56 Capped-head pile over fixed-head pile ratio (front pile, Test PG2/PG3)

Figure 6.57 Capped-head pile over fixed-head pile ratio (rear pile, Test PG2/PG3)


Figure 6.59 Tunnelling-induced middle pile in capped-head 6-pile group (Test PG4) and corresponding middle pile in fixed-head 6-pile group (Test PG5) axial force

Figure 6.60 Tunnelling-induced rear pile in capped-head 6-pile group (Test PG4) and corresponding rear pile in fixed-head 6-pile group (Test PG5) axial force

Figure 6.61 Tunnelling-induced front pile in capped-head 6-pile group (Test PG4) and corresponding front pile in fixed-head 6-pile group (Test PG5) bending moment

Figure 6.62 Tunnelling-induced middle pile in capped-head 6-pile group (Test PG4) and corresponding middle pile in fixed-head 6-pile group (Test PG5) bending moment

xxv

Figure 6.63 Tunnelling-induced rear pile in capped-head 6-pile group (Test PG4) and corresponding rear pile in fixed-head 6-pile group (Test PG5) bending moment

Figure 6.64 Variation of maximum bending moment for front, middle and rear pile in the short-term (Tests PG4 and PG5)

Figure 6.65 Variation of maximum bending moment for front, middle and rear pile in the long-term (Tests PG4 and PG5)


Figure 6.67 Capped-head pile over fixed-head pile ratio (middle pile, Test PG4/PG5)


Figure 6.69 Long-term over short-term ratio (Tests PG1, 2, 3, 4 & 5)

xxvi

NOMENCLATURE

C Tunnel cover

Cu Undrained shear strength of clay

Cv Coefficient of consolidation

D Tunnel diameter

EA Pile axial rigidity

EI Pile flexural rigidity

g Gravitational acceleration (9.8 m/s2)

Gs Specific gravity

H Tunnel depth

i Point of inflection

k Coefficient of permeability

K Trough width parameter

L Pile length

S Soil settlement induced by tunneling

Smax Maximum ground surface settlement

U3D Parameter accounting for three dimensional heading effects (Gap Parameter Method)

V Tunnel Volume loss

X Horizontal distance between pile and tunnel centre

Z Depth below ground surface

Zo Depth to tunnel axis level

φ’ Soil effective friction angle

φcrit Critical state friction angle

γ’ Effective unit weight of soil

γw Unit weight of water

minρ Minimum dry density

maxρ Maximum dry density

xxvii

εx,z Ground loss with horizontal and vertical distance

ε0 Ground loss ratio

M Slope of critical state line in p’-q space

λ Slope of isotropic compression line in p’-v space

κ Slope of swelling line in p’-v space

Abbreviations

LL Liquid Limit

LT Long-term

NP Neutral Plane

NSF Negative Skin Friction

NUS National University of Singapore

PI Plasticity Index

PIV Particle Image Velocimetry

PPT Pore Pressure Transducer

PL Plastic Limit

ST Short-term

VL Volume loss

Chapter 1 Introduction

1

CHAPTER ONE

INTRODUCTION

1.1 BACKGROUND

In urban areas, tunnels are often constructed close to existing buildings due to space

constraints. As cities develop rapidly, the need for underground transportation and

utility tunnels have greatly increased. Unfortunately, many structures exist long before

the tunnels are planned. As such, it is increasingly complex and challenging to build

tunnels underground as the tunnels almost inevitably run close to or underneath some

piled foundations supporting existing buildings.

Tunnel excavations generally cause ground settlement and deformation nearby,

especially in soft clay. Ground movements due to tunnelling would induce additional

axial (settlement and axial force) and lateral (deflection and bending moment)

responses on adjacent pile foundations (Figure 1.1). As pile foundations supporting

existing buildings are generally designed to resist compression load only (Figure 1.2),

the foundations may not be safe to resist the induced loads due to tunnelling. At

present, there are very few reliable design methods (Chen at al., 1999; Loganathan et

al., 2001; Pang et al., et al. 2005b) to assess induced pile responses due to tunnelling.

Although many studies have been conducted to investigate tunnel-soil-pile

interactions in the short-term, impacts in the long-term are still not well understood. A


2

published field study covering an 11-year period of post-tunnelling monitoring for the

Haycroft Relief Sewer in Grimsby, UK, (O’Reilly et al., 1991) in very soft clay

recorded that the soil settlement in the long-term had increased significantly. The

results of centrifuge model tests (Ran, 2004) conducted at the National University of

Singapore (NUS) also revealed that for tunnelling in clay, the ground continues to

deform long after the completion of tunnelling, thus inducing further settlement,

deflection, axial and lateral loads on adjacent piles in the long-term.

In view of the complexity of field instrumentation and monitoring, physical

modelling can be an attractive mean to study the tunnel-soil-pile interaction problem in

both short-term and long-term. One effective way is to conduct centrifuge model tests

employing artificial gravitational field to replicate the prototype stress level as

experienced by the ground in the field. Under a well-controlled environment,

centrifuge tests can provide flexibility and repeatability to study tunnel-soil-pile

interaction problems that could not be achieved in field tests (Mair et al., 1984;

Loganathan et al., 2000; Ran, 2004; Jacobz et al., 2004).

1.2 TUNNELLING-INDUCED SOIL MOVEMENTS

Many research studies have been carried out to investigate tunnelling-induced soil

movements. Peck (1969), Schmidt (1969), O’Reilly and New (1982), Lake et al.

(1992), Mair et al. (1993) and others developed empirical formulae from field studies

to predict the soil movements induced by tunnelling. In addition, several centrifuge

model studies including Loganathan (2000) and Ran (2004) were conducted to

examine soil movements due to tunnelling.


3

Besides empirical formulae and centrifuge model studies, analytical solutions

have been developed by researchers including Sagaseta (1987), Verrujit and Booker

(1996), Loganathan and Poulos (1998), Park (2005) and Osman et al. (2006a) to

predict the ground displacements for various shapes of tunnel deformation. However,

such analytical solutions cannot account for all aspects, in particular the time effects.

Many researchers reported field measurement results for soil movements induced by

tunnelling. However, most of these studies did not report field measurements during

the post-tunnelling period despite some reports on significant long-term settlements

after tunnelling in soft clay; see for example, Haycroft Relief Sewer in Grimsby, UK,

(O’Reilly et al., 1991) and Shanghai Metro Tunnel No.2, China, (Zhang et al., 2004).

From the above review, it is evident that the mechanism and calculation of

tunnelling-induced soil movements in the short-term are reasonably well studied.

However, the long-term tunnel-soil interaction in soft clay clearly needs further

investigation.

1.3 EFFECTS OF TUNNELLING ON PILES

It is generally not viable to monitor the responses of existing piles due to tunnelling.

This is because the piles usually exist long before the tunnels are constructed and it is

almost impossible to install strain gauges in the existing piles to monitor the pile

responses. However, some valuable field measurements in limited cases have been

made available where the existing piles were instrumented prior to tunnel construction,

see Selemetas et al. (2005), Jacobsz et al. (2005), Pang et al. (2005) and Cham (2007).


4

Although a good number of centrifuge model studies (Loganathan et al., 2000; Feng,

2003; Ran, 2004 and Jacobsz et al., 2005) and numerical analysis and analytical

solutions (Chen et al., 1999; Loganathan et al., 2001 and Pang et al.; 2005) have been

attempted to investigate the effects of tunnelling on piles, the results of literature

review presented in Chapter 2 reveal that there are still major gaps for a better

understanding of pile responses due to tunnelling. Two examples of major gap include

limited field studies on long-term tunnel-soil-pile interaction, and lack of predictive

methods available to evaluate the effects of pile group or soil-structure interaction. The

development of a qualitative and quantitative framework to assess potential impact of

tunnel construction on existing piled foundations is needed to improve the current

knowledge on the prediction of pile responses due to tunnelling.

1.4 OBJECTIVE AND SCOPE OF STUDY

The main aim of the present study is to investigate tunnel-soil-pile interaction in soft

clay. More specifically, the objectives of this research are:-

1. To develop a simulation technique of tunnel excavation associated with

inward tunnel deformation pattern using centrifuge modelling technique. This

pattern is chosen because it is often observed in practice. However, the wish-

in-place model tunnel is a simplification and idealization of a cavity

contraction, but with an oval-shaped GAP and well controlled volume loss to

simulate the soil movements induced by tunnelling when the tunnel

excavation has passed a particular section. The minimum volume loss that


5

can be simulated by the present model tunnel is only 3% in order to maintain

the accuracy of the experiment results.”

2. To study tunnelling-induced soil movement in free-field experimentally with

subsequent Particle Image Velocimetry (PIV) analysis. The purpose of this

investigation is to examine the mechanism of tunnel-soil interaction. Long-

term post-tunnelling soil movement was also studied.

3. To study the induced pile responses due to tunnelling experimentally. The

vertical (axial force and settlement) and lateral (bending moment and

deflection) pile responses were examined. In addition, parametric studies

were performed to evaluate the effects of various tunnel volume losses, pile

tip conditions, pile lengths, pile distances to tunnel and pile groups. Long-

term post-tunnelling pile performance was also studied.

1.5 STRUCTURE OF THESIS

This thesis consists of seven chapters and the contents of subsequent chapters are

briefly described as follows:

a) Chapter 2 reviews the existing literature on to tunnel-soil-pile interaction. The

review is divided into three parts. Firstly, various methods of simulating tunnel

excavation in centrifuge tests are reviewed. Subsequently, ground movements

caused by different methods of tunnel excavation are examined. Finally, existing

research studies concerning pile responses due to tunnelling are highlighted.

b) Chapter 3 describes the present centrifuge model set-up and experimental


6

procedures. A detailed description of the novel technique for simulation of

tunnelling during centrifuge flight is also presented.

c) Chapter 4 first presents the experimental results on tunnelling induced soil

movements analysed by PIV. The test results on the basic tests of volume loss for

tunnelling on single piles are then presented in detail.

d) Chapter 5 presents the results of parametric studies on the effects of tunnelling on

single piles

e) Chapter 6 presents the results on effects of tunnelling on pile groups

f) Chapter 7 summarizes the main findings of the present study and proposes future

works.


7

Figure 1.1 Pile responses induced by tunnel construction: (a) Tunnelling under pile foundation, (b) Tunnelling adjacent to pile foundation.

1

5

10

15

20

25

30

Depth (m)

Induced Pile Settlement

Induced BM Tunnel

Contraction

Settlement Trough

Induced Axial Loads35

Existing Building

(NTS)


8

Figure 1.2 Pile foundations supported existing buildings normally designed to resist compression load only.

1

5

10

15

20

25

30

Depth (m)

Existing Building

New tunnel

Steel reinforcement

Existing Pile

(NTS)

Chapter 2 Literature Review

9

CHAPTER TWO

LITERATURE REVIEW

2.1 INTRODUCTION

Tunnelling would cause soil movements that would in turn induce axial and lateral

loads on adjacent pile foundation. Hence, it is of great interest to understand the effects

of tunnelling induced soil movements on existing piles. Many studies on free-field soil

movements due to tunnelling have been reported. However, few studies concerned the

pile responses subject to soil movements caused by tunnelling, especially for long-term

pile behaviours. The methods, effects and difficulties faced in investigating the

behaviour of piles and soil due to tunnelling will be reviewed in detail in this chapter.

The topics investigated include: (a) model simulation technique of tunnelling; (b)

tunnelling-induced soil movements and (c) tunnelling-induced pile responses.

2.2 TECHNIQUES FOR SIMULATION OF TUNNELLING IN

CENTRIFUGE

Prior to the study of pile behaviours due to tunnelling, the correct simulation of tunnel

excavation plays an important role. The soil movement pattern is directly caused by

the model tunnel and affects the pile responses.


10

Tunnel construction is three-dimension in nature and time dependent. In

general, the settlement due to shield tunnelling can be derived from the face loss

during tunnel advancement, shield loss due to over-cut, tail void closure and

consolidation (Shirlaw et al., 2003). With the increasing availability of large-scale

computing resources, 3D finite elements simulations of tunnels are reported in the

literature in recent years (Komiya et al., 1999; Augarde and Burd, 2001; Lim, 2003;

Lin et al., 2002; Melis et al., 2002; Ng et al., 2004; Lee and Ng, 2005; Ng and Lee,

2005; Lee et al., 2006; Phoon et al., 2006). However, realistic three-dimensional finite

element analysis has still not reached a point of development where it can be routinely

used in engineering design. Two common problems associated with three-dimensional

analyses mean very large memory requirement and long computing times (Lee et al.,

2006; Phoon et al. 2006).

In order to compare the performance of 2D and 3D analyses, Moller (2006)

performed a back analysis for the case of Second Heinenoord Shield Tunnel. It is

reported that the surface settlements of 2D analysis compare well with the 3D analysis,

as the face pressure does not much contribute to the development of the surface

settlement. Moreover, from the field monitoring of Mass Rapid Transit North East

Line Contract C704 studied by Pang (2006), it is apparent that the maximum axial

force developed in the pile when the Earth Pressure Balance machine (EPBM) was

directly adjacent to the pile. Besides, the induced pile bending moment in the

longitudinal direction is either equal or smaller than the induced pile bending moment

in transverse direction. This study clearly demonstrated that the transverse section is

more critical, and hence the simulation of plane strain model of a long tunnel is

comparable to 3D modelling in this case.


11

With the availability of centrifuge modelling technique, a prototype problem

can be simulated in a scaled laboratory model hence overcoming the limitation of 1-g

model. Most of the centrifuge tests on tunnel-soil interaction that have been carried out

(Bezuijen and Schrier, 1994; Hergarden et al., 1996; Loganathan, 1999; Ran et al.,

2003; Jacobsz et al., 2002; Feng et al., 2002; Lee and Chiang, 2004) thus far

considered only a wish-in-place plane strain tunnel in the simulation. This is probably

because the development of 3D model tunnel in centrifuge is very difficult and has its

limitation of boundary conditions and constraints. Therefore, If thein a situation

whenre the tunnel excavation has passed a particular section is considered, the vectors

of the ground movement developed will be more or less in the plane perpendicular to

the tunnel axis. Consequently it is reasonable to assume that a plane strain model of

long tunnel section would be a good representation of tunnelling-induced soil

movements; this is usually referred to as a two-dimensional simulation (Taylor, 1998).

Nevertheless, if the tunnel face passes directly under a foundation, the 3D soil response

around tunnel face would be important. However, this scenario has not been examined

in this research.

The review of various methods of simulating tunnel excavation in the

centrifuge is briefly discussed here in order to gather useful information and guidance

for the development of model tunnelling technique in the present study.

2.2.1 Simulation Technique 1 - High Density Polystyrene Foam

Sharma et al. (2001) presented a method to simulate tunnel excavation in the

centrifuge by dissolving polystyrene foam quickly using an organic solvent. Figure 2.1

shows the arrangement for the inflow of solvent into the tunnel core. The polystyrene


12

foam core was placed tightly inside the model tunnel lining, which was made by

wrapping a half hard brass foil around the foam core and soldering the lap joint with

the help of tin solder and an electronic gun. The flow of this liquid into the polystyrene

foam (model tunnel) is controlled by using solenoid manifold and solvent reservoir.

The stiffness of the filled tunnel can approximately be made to be equivalent to that of

the parent soil. The lining is left in place when the foam core has been dissolved. This

technique is an improvement over other methods of modelling tunnel excavation, such

as reduction of air pressure supporting the tunnel lining or gradually draining heavy

liquid from within the lining. A limitation of this modelling technique is that this

approach may not correctly simulate the actual tunnel excavation situations in the field

as it can only simulate excavation cases with tunnel spring line moving outwards.

The technique proposed by Sharma et. al. (2001) for the simulation of tunnel

excavation in the centrifuge has been adopted by Feng (2003) and Ran (2004) at

National University of Singapore (NUS). Feng (2003) carried out a centrifuge model

study to investigate tunnel-soil-pile interaction in sand. Ran (2004) extended Feng’s

study on tunnel-soil-pile interaction to clay.

Figure 2.2 shows the shape of the deformed tunnel lining over time obtained

from a typical test. It can be seen that the lining deforms into an oval shape with tunnel

spring lining protruding slightly outwards. In most cases in practice, the over-

excavation of tunnel and the gap between the shield tunnel machine and lining would

cause the tunnel spring line to move towards the tunnel resulting in inward soil

movements at the tunnel spring line.


13

2.2.2 Simulation Technique 2 - Compressed Air

One of the earlier model tunnelling technique has been developed at the University of

Cambridge (Potts, 1976; Mair et al., 1984) in which air pressure is applied to control

tunnel support conditions. This technique has subsequently been used by Chambon et

al. (1991), König et al. (1991), Lee et al. (1999), Grant and Taylor (2000) and Bilotta

and Taylor (2005). Figure 2.3 shows the schematic diagram of a typical plane strain

centrifuge model tunnel using compressed air (Grant and Taylor, 2000). In this

modelling technique, the soil mass, which has to be removed during the excavation, is

represented in the model by a rubber membrane pressurized with air pressure. The

shape of the rubber membrane is identical with the contour of excavation. The

equilibrium is kept between the internal air pressure and earth pressure during model

preparation and increasing g-level. The excavation is simulated by decreasing the

internal air pressure.

To keep equilibrium between internal air pressure and the earth pressure, it

would be necessary to apply internal support according to the theoretical earth pressure

at rest. This support cannot be achieved by the air pressure technique, because air

pressure remains constant along the contour of excavation whereas earth pressure

varies with the change of orientation from vertical to horizontal and due to increase in

vertical stress with depth.


14

2.2.3 Simulation Technique 3 - Liquid – Oil / Water

The limitation of simulation of tunnelling by air pressure can be eliminated by

applying fluid pressure inside the rubber membrane. This was realized in several

studies related to excavation of shafts and trenches (Lade et al., 1981; Mair et al.,

1984). A zinc chloride solution has been used to achieve a fluid with a density similar

to that of the surrounding soil. This technique is well suited to study the relationship

between the displacements of soil into the excavation and surface settlements or to

evaluate the failure conditions. However, the technique needs to be evaluated carefully

if the stress strain conditions (depending on the tunnelling technique, soil stiffness,

stiffness of lining as well as load concentration at the end of the tunnel lining) in the

soil close to the excavation are most relevant for the investigated effect. (König, 1998).

Figure 2.4 shows a cross-section of the model tunnel infilled with oil developed

by Loganathan et al. (2000). The stability of tunnelling procedure was represented by

the equivalent surface ground loss values. A technique of decreasing tunnel diameter

during centrifuge flight was adopted to model the required ground loss. The inner core

of the tunnel was made of a long aluminium tube, and a very thin rubber membrane

was placed on top of the inner core cylinder. The rubber membrane was attached at

both ends of the inner core cylinder by end caps which prevented any leakage of oil

from between the inner core and the membrane. A small hole was drilled through the

end cap and inner core to allow the passage of oil. The cylindrical face of the assembly

was then covered by a 0.5-mm thick smooth-surfaced overlapping PVC spring to

enhance the tunnel lining stiffness, and to ensure a uniform change in the tunnel

diameter during the test. A syringe pump was fabricated to control the volume of oil in


15

the tunnel assembly. The advantage of this model is that various volume losses can be

simulated in one test by extracting oil slowly.

The recent model tunnel reported by Jacobsz (2002) was infilled with water to

study the tunnelling effects on single piles in sand. The model tunnel consisted of a

brass mandrel surrounded by a 1-mm thick latex membrane as shown in Figures 2.5

and 2.6. During centrifuge tests, the approximately 4-mm thick annulus between the

mandrel and the membrane was filled with water that could be extracted to accurately

impose volume losses from 0% to about 20% on the surrounding soil. The outer

diameter of the model tunnel was 60 mm, representing a 4.5-m diameter tunnel at

prototype scale. A pressure transducer was incorporated into the tunnel control system

to monitor the pressure in the annulus of water between the latex membrane and the

brass mandrel.

The tunnel was connected via a solenoid valve to a standpipe in which a

constant water level was maintained to automatically balance the tunnel pressure with

the overburden pressure during the acceleration of the centrifuge. After the desired

acceleration had been achieved and the model piles installed, the solenoid valve was

closed and the volume loss was imposed by extracting water slowly from the annulus

around the tunnel.

Nevertheless, according to König (1998), there are indeed short-comings of the

above model tunnels which control the volume loss by changing the volume of gas or

fluid to simulate the volume loss. This is because the density of gas or fluid used is

different from that of soil. Thus, the initial stress condition of the model tunnel in this


16

simulation technique is one of the major concerns.

Jacobsz (2002) reported that due to the self-weight of the water in the tunnel, a

hydrostatic pressure gradient would have existed in the tunnel, with the pressure near

the crown 22 kPa lower and the pressure at the invert 22 kPa higher than measured.

Besides, a small amount of water trapped near the tunnel ends, which would have been

difficult to extract completely, would also have increased the registered tunnel pressure.

In addition, the stiffness of the tunnel membrane and the fact that the contraction of the

model tunnel occurred non-uniformly, would have affected the measured pressures.

Most factors above would have resulted in the pressure being over-registered.

Moreover, the model tunnel is fixed at both ends of the strong box, as shown in

Figure 2.7, which would induce boundary condition and is not realistic as the middle

part of the tunnel will suffer much higher stress and displacement as compared to the

both ends of the tunnel when subject to high-g acceleration.

2.2.4 Simulation Technique 4 - Mechanical Equipments

Relatively complicated mechanical equipments have been developed to simulate the

process of shield tunnelling in centrifuge. Figure 2.8 shows an in-flight miniature

shield tunnelling machine developed by Nomoto et al. (1994). This machine allowed

the in-flight excavation of soil and also an in-flight installation of the lining. This time

the whole shield tunnelling process was successfully reproduced in a centrifuge force

field. Owing to the small size of the centrifuge and the restrictions in terms of

dimensions and weight of the model, the attempt to evaluate earth pressure acting on


17

the tunnel lining quantitatively was not yet fulfilled. The tail void was also too large

compared with the prototype machine (Nomoto et al., 1994).

A plane-strain shield model machine was developed by Yasuhiro et al. (1998)

to perform centrifuge tests in both sand and clay to study the tunnelling-induced

ground movement. The shield model machine consists of steel rings and a wedge

shaped shaft, which are able to simulate the tail void and backfill grouting in flight by

contraction and expansion of shield model rings controlled by a motor, see Figure 2.9.

Recently, Ghahremannejad at el. (2006) simulated volume loss by varying the

diameter of the model tunnel along the tunnel. The desired volume loss was achieved

when the smaller aluminum tubes were pushed through the sand box. Lee and Yoo

(2006) adopted another technique simulating the volume loss by decreasing the

diameter of model tunnel as shown in Figure 2.10. However, the recent model tunnels

developed by Ghahremannejad at el. (2006) and Lee and Yoo (2006) can only simulate

tunnel volume loss under 1g condition.

2.3 TUNNELLING-INDUCED SOIL MOVEMENTS

2.3.1 Field Studies of Tunnelling-Induced Soil Movement

Peck (1969), Schmidt (1969) and subsequently many other researchers have shown

that the transverse settlement trough after tunnel excavation in short-term (Cham, 2007)

can be well-described by a Gaussian distribution curve as shown in Figure 2.11. The

method needs an estimate of volume loss (V) and the trough width parameter (K) to


18

obtain the maximum ground surface settlement (Smax) and subsequently the surface

settlement profile, as defined in Figure 2.12. The ground settlements are generally

negligible beyond an offset of 3i from the tunnel centre line for Peck’s (1969)

proposed curve (Figure 2.13). The surface settlement trough,S, and volume loss,V, are

approximated as follows:

⎟⎟⎠

⎞⎜⎜⎝

⎛−= 2

2

max 2exp

ixSS (2.1)

max2 iSV π= (2.2)

where x is the offset to the tunnel vertical line and i is point of inflection.

The trough width parameter K is relatively easy to quantify as it is largely

independent of construction method and operation experience (Fujita, 1981; O’Reilly

and New, 1982). Numerous field data have been collected and hence many estimates

of trough width parameters have been proposed. However, a comprehensive summary

done by Lake et al. (1992) on tunnelling data from many countries on clay shown in

Figure 2.14 has established the general variations of i as follows:

Simple approximate relationship oKzi =

Where K = 0.4 to 0.6 for clays (soft and stiff), and

K = 0.25 to 0.45 for sands and gravels.

zo= depth to tunnel axis

Lake et al.’s (1992) field studies supported the various proposals that K can be

assumed solely as 0.5 for tunnelling in clay (O’Reilly and New, 1982; Mair et al.

1993).


19

The subsurface settlement profiles can also be reasonably approximated by a

Gaussian distribution curve in the same way as surface settlements. However, based on

the field data collected as shown in Figure 2.15, Mair et al. (1993) proposed that at a

depth z below the ground surface and above a tunnel depth Z0, the trough width

parameter for tunnels constructed in clay, i, can be expressed as:

)( zzKi o −= (2.3)

⎟⎠⎞⎜

⎝⎛ −

⎟⎠⎞⎜

⎝⎛ −+

=

o

o

zz

zz

K1

1325.0175.0 (2.4)

Trough width parameter is shown to increase with depth and significantly

underestimated if a constant value is assumed. Similar patterns of increase in K was

observed in studies by Moh et al. (1996) and Dyer et al. (1996) irrespective of the soil

conditions encountered. Centrifuge studies by Grant and Taylor (2000) confirmed that

the proposed variation of K with depth for clay by Mair et al. (1993) provided a good

fit to the data obtained from tests within a certain range between the ground surface

and tunnel axis level.

Ran et al. (2003) observed that the Gaussian Curve can only describe the short-

term surface settlement well. A comprehensive review of field data of post-

construction settlements above tunnels in soft clay has been carried out by Shirlaw

(1993), Shirlaw et al. (1994) and Shirlaw (1995). Shirlaw (1993) recorded examples of

long-term settlements where the long-term component had the effect of increasing the


20

short-term settlement by up to a factor of 10. The more typical settlement increase in

the long-term is in the order of 30% to 100%.

A case history study covering an 11-year period on the Haycroft Relief Sewer

at Grimsby in very soft clay soil had recorded that the initial settlement trough

increased from 1.5Z to a final value in excess of 4Z, as shown in Figure 2.16. The

maximum surface settlements over time is shown in Figure 2.17 reveals that the

consolidation of the disturbed settlements around the tunnel in long-term cannot be

neglected. Similar time dependent responses can also be found in Shanghai Metro

Tunnel No.2 (Zhang et al., 2004). The ratio of the maximum immediate settlement to

maximum long-term settlement shown in Figure 2.18 clearly demonstrates that the

long-term settlements were significant, and hence cannot be neglected.

Thus, generally it has been clearly shown that post-construction settlements can

be significant, particularly for tunnels in soft and compressible clay. Some case

histories shown in Figure 2.19 have much wider settlement troughs in the long-term.

Similar widening of settlement trough has been reported by a number of authors (e.g.

Glossop, 1978; Howland, 1980; O’Reilly et al., 1991 and Bowers et al., 1996).

Mair and Taylor (1997) concluded that the long-term settlement troughs are

similar to classical Gaussian curve associated with short-term settlement when positive

excess pore pressure are generated during tunnelling; whereas wider long-term

settlements are related to tunnel lining acting as drain and the development of steady

state seepage towards the tunnel. Hence, the major factors influencing the development

of long-term settlements are (i) initial magnitude and distrbution of pore water pressure;


21

(ii) compressibility and permeability of the soil; and (iii) pore pressure boundary,

particularly the permeability of the tunnel lining relative to the permeability of the soil.

For normally consolidated clays, significant zones of positive excess pore

pressure can be generated even for a tunnel where unloading occurs, as shown by

Schmidt (1989). Negative excess pore pressures were noted included near the tunnel,

while positive excess pore pressures due to shearing may result a short distance away.

The pattern of pore pressure is shown in Figure 2.20, whereby two similar case

histories were observed for Shanghai Metro (Schmidt, 1989) in Figure 2.21 and

Thunder Bay Sewer Tunnel (Shirlaw et al., 1994) in Figure 2.22. The extent of the

positive excess pore pressure in Thunder Bay Sewer Tunnel spanned over six tunnel

diameters from the tunnel centre line.

For lateral soil movement due to tunnelling in long-term, very limited field data

are available. Emeriault et al, 2005 showed that in Toulouse subway line B, the

transverse movements were significantly increased over time, but stabilized after 15

days of tunnel excavation, as shown in Figure 2.23. Besides, Balasubramanian (1987)

reported both vertical and horizontal soil movement over almost a year time for the

Singapore’s largest effluent outfall pipeline in tunnel through various types of soils,

mostly marine clay. The data shown in Figure 2.24 demonstrated the time dependent

behavior of horizontal soil movements.


22

2.3.2 Centrifuge Model Tests of Tunnelling-Induced Soil Movement

With the advances of new image processing technology, centrifuge modelling emerges

to be an attractive technique to investigate the effects of tunnelling-induced soil

movement on adjacent piles. The results enable an in-depth understanding into the

mechanism of ground responses associated with tunnel construction in terms of surface

and subsurface soil movements, as well as soil stresses.

Feng (2003) and Ran (2004) adopted the technique proposed by Sharma et al.

(2001) for the simulation of ovalisation of tunnel using high density polystyrene foam.

Figure 2.25 shows the surface settlement troughs under approximate similar volume

loss in clay and sand, as reported by Ran (2004). The two measured settlement troughs

follow Gaussian curve well. However, it is evident that the settlement troughs are

markedly different as the sand settlement trough is much narrower than that of clay.

This indicates different settlement propagation mechanisms in clay and sand. In clay,

the deformation of the soil propagates ‘gradually’ upwards and outwards from the

tunnel cavity to the ground surface. However, the deformation in sand propagates

sharply and almost vertically from the tunnel to the ground surface. Therefore, the

different mechanisms suggest that the ground deformation in sand may cause more

severe damages to the ground surface or structures above and nearby the tunnel. For

clay, the soil movements cause differential settlement spreading a wider range. This

may explain why sinkholes on the ground surface associated with tunnelling are

mainly spotted in competent soils like sand in the field; while in clay, such drastic

settlements are less common.


23

Figure 2.26 shows the surface settlement trough over time in clay (Ran, 2004).

It is noted that the measured short-term surface settlement trough follows the Gaussian

distribution curve fairly well. After the completion of tunnel excavation, the soil

continues to settle with time and the rate of increase in settlement decreases with time.

The incremental soil settlements become negligible after 720 days of tunnel excavation.

However, Gaussian curve is found to be inappropriate to depict the measured long-

term surface settlement troughs. The measured final trough has a somewhat wider

parabolic shape than that of Gaussian curve. Furthermore, Gaussian distribution curve

largely underestimates the measured settlement at the far end of the ground surface,

showing that the spread of the surface settlement trough increases over time. Grant and

Taylor (2000) carried out a series of centrifuge tests to investigate tunnelling-induced

ground movements in clay. As discussed in Section 2.2.2, tunnelling was simulated by

reducing the compressed air pressure in a model tunnel lined with a latex membrane

and this caused a uniform radial contraction of tunnel. Figure 2.27 shows the profile of

normalized vertical and horizontal ground movements at different subsurface

elevations. In the near surface region, the horizontal movements are not well described

by assuming an average vector focus (Based on Grant and Taylor (2000), the average

position of the vector focus can be used to give average distribution of horizontal

movement) but the agreement is very good at all other elevations. The studies also

suggested that a constant trough width regardless of volume loss is common to all of

the tests and the centrifuge studies confirmed that the proposed variation of K with

depth for clays by Mair et al. (1993) provides a good fit to their centrifuge test data.

However, the long-term behaviour of the soil movements is not studied, probably due

to the soil condition of moderately stiff clay with less significant in long-term

settlement.


24

Simulation of tunnelling using liquid pressure was proposed by Loganathan

et al. (2000). The tunnel deformation pattern is uniform radial contraction. Figure 2.28

shows that the ground settlement troughs measured in the centrifuge tests match well

with his analytical prediction (Loganathan et al., 1998) which is based on oval-shaped

deformation. However, the long-term effect has not been studied by Loganathan et al.

(2000).

2.3.3 Predictive Methods of Tunnelling-Induced Soil Movement

Although attractive as a predictive tool, analytical methods are mathematically limited

in the efforts required to derive solutions accounting for material non-linearity,

complex geometries and time effects. Thus, relatively few analytical solutions are

available.

Sagaseta (1987) presented an analytical solution to predict tunnelling induced

ground movements for a weightless incompressible soil by simulating ground loss

around a tunnel in the form of a point sink. The tunnel is first assumed to be located

within an elastic infinite medium where it collapses uniformly. The soil is modelled as

a linear-elastic material and the solution is based on fluid mechanics concepts.

Solutions derived by Sagaseta (1987) are subsequently extended by Verrujit

and Booker (1996) to account for compressible materials and the effects of ovalisation

of the excavated tunnel boundary. However, the method gives a wider surface

settlement profile and larger horizontal movements than observed in practice.

Loganathan and Poulos (1998) modified Verruijt and Booker’s solution to give


25

narrower settlement troughs and to account the construction effects empirically.

Osman et al. (2006a) developed a kinematic plastic solution for ground

movements around a shallow, unlined, tunnel embedded within an undrained clay layer.

In this solution, the pattern of deformation around the tunnel is idealised by a simple

plastic deformation mechanism (see Figure 2.29). Within the boundaries of the

deformation zone, the soil deforms compatibly following a Gaussian distribution.

Outside this zone, the soil is assumed to be rigid. This mechanism does not incorporate

slip surfaces. Osman et al. (2006a) demonstrated that the upper bound theorem applied

to distributed shearing mechanism offered a reasonable assessment for collapse, and

also matches the displacement field observed in centrifuge tests of tunnel failure in

clay (Mair, 1979). It was also demonstrated that the shape of the surface settlement

profile remained the same as the magnitude of settlement increased towards failure

(Osman et al. , 2006a).

In addition, Osman et al. (2006b) also demonstrated that an upper bound style

of calculation is also capable of predicting ground displacements at any stage prior to

failure, representing the soil as a strain-hardening plastic material. A simplified closed-

form solution is provided for the prediction of maximum surface ground settlement for

the particular case of deep tunnelling. This solution is obtained by integrating the

equilibrium equations along the tunnel centre-line from the tunnel circumference up to

the ground surface and by invoking radial symmetry. A simple power curve was used

to model the stress–strain relations. These analytical solutions for maximum surface

settlements have also been validated against the centrifuge test data, and gave close

correspondence for deep tunnels but under predicted tunnel support pressure by about


26

20% for shallow tunnels (Depth to tunnel crown/ depth of tunnel axis, C/D <3).

Analytical solutions providing the most convenient way in predicting

tunnelling induced ground movements. However, arguably accurate empirical method

has certain limitations in accounting for the effect of ground conditions, construction

methods, post-construction ground responses and precise tunnelling model.

In view of the above, a well-calibrated ‘GAP’ method was proposed by Rowe

et al. (1983). The ‘GAP’ parameter is defined as the magnitude of the equivalent two-

dimensional (2D) void formed around the tunnel due to the combined effects of the

three-dimensional (3D) elastoplastic ground deformation at the tunnel face, over-

excavation of soil around the periphery of the tunnel shield, and the physical gap that

is related to the tunelling machine, sheild, and linming geometry.

The GAP parameter can be estimated using a theorectical method developed by

Lee et al. (1992), once the details of the machine support system and the soil

parameters are given. Comprehensive guidelines have been provided to calculate the

gap parameter (Lee et al. 1992), as

GAP = Gp + U3D + w (2.5)

where Gp represents the difference between the cutter head and outer lining diameter

while U3D and w account for 3D heading effects and workmanship quality,

respectively.


27

In practice, as pointed out by Rowe et al. (1983), the radial ground deformation

is not uniform since the equivalent 2D gap (tail void) around the tunnel is non-circular

(e.g. typically oval-shaped) as shown in Figure 2.30. The possible reasons for the

formation of an oval-shaped gap around the tunnel are: (1) tunnel operators advance

the shield at a slightly upward pitch relative to the actual design grade to avoid the

diving tendency of the shield; (2) the tunnel lining settles on the ground when the tail

piece is removed; and (3) 3D elasto-plastic movement of the soil occurs at the tunnel

face.

Loganathan and Poulos (1998) extended the study from Lee et al. (1992) by

incorporating the shape of tunnel deformation to predict tunnelling-induced undrained

ground movements around a tunnel in soft ground. The traditional definition of the

ground loss parameter is redefined as ‘equivalent ground loss parameter ε’ with respect

to the gap ‘g’ parameters and incorporated in the analytical solutions. The nonlinear

ground movement due to the formation of an oval-shaped gap is then modelled by

adopting an exponential function to the equivalent undrained ground loss with

appropriate boundary conditions. Hence, the analytical solution models the effect of

non-uniform soil convergence around a deforming tunnel as shown in Figure 2.31.

The analytical solution of ground loss with horizontal and vertical distance єx,z

from the tunnel centre is given as:

⎭⎬⎫

⎩⎨⎧

⎥⎦

⎤⎢⎣

⎡+

+= 2

2

2

2

0,69.0

)(38.1exp

Hz

RHx

zx εε (2.6)


28

where ε0 is the ground loss ratio, H is the tunnel depth, z is the depth below ground

surface and x is the lateral distance from tunnel centre-line.

Although the method has been successfully used to back analyze some case

histories in clay, calculated results have to be treated with caution as certain important

conditions necessary in the derivation of analytical solutions are violated and the

volume loss is not conserved for undrained cases when empirical assumptions are

introduced (Cheng, 2003).

The accuracy of the oval-shaped radial displacement of tunnelling has been

further verified by the analytical solutions presented by Park (2005) through the case

studies. The oval-shaped ground deformation pattern (Loganathan and Poulos, 1998;

Park, 2004) is imposed as the boundary condition of the displacement at the opening to

consider real non-uniform ground deformation pattern. The gap parameter (Lee et al.,

1992) is used to describe the displacement at the opening.

Four simple boundary conditions, one for uniform radial deformation pattern

(B.C.-1) and three for oval-shaped deformation pattern (B.C.s-2–4), are considered at

the tunnel opening as shown in Figure 2.32. The boundary condition B.C.-2 (oval-

shaped) is chosen for further study to give a conservative estimation for lateral

displacement. Five case studies have been used to check the applicability of the

proposed analytical solutions. The surface and maximum subsurface settlements and

the lateral displacement predicted using the proposed analytical solutions are

comparable to those from the methods of Verruijt and Booker (1996) and Loganathan

and Poulos (1998), and in reasonable agreement with field observations for tunnels in


29

uniform clay. Hence, it can be concluded that the oval-shaped radial displacement is a

better tunnelling model that provides more realistic predictions of ground movement in

uniform clay.

2.4 TUNNELLING-INDUCED PILE RESPONSES

2.4.1 Field Studies of Tunnelling-Induced Pile Responses

Selemetas et al. (2005) presented the results of a full-scale trial investigating the

effects of tunnelling on piles. The field study took place in Channel Tunnel Rail Link

(CTRL) Contract 250, in Dagenham, Essex, UK. The Contract involved the

construction of 5.2 km of twin 8-m diameter bored tunnels using two Lovat Earth

Pressure Balance (EPB) machines with tail-skin grouting. The study involved the

installation, loading and monitoring of four instrumented piles along the route of the

twin tunnels . A comparison was made between the resulting pile head settlements due

to tunnelling with the surrounding ground surface movements. The results identified

three zones of influence in which the pile settlements were correlated to the ground

surface settlements (see Figure 2.33):

1. Piles in Zone A settled 2-4mm more than the ground surface,

2. Piles in Zone B settled by the same amount as the ground surface,

3. Piles in Zone C settled less than the ground surface.

4. The well-established Gaussian curve describing the magnitude of ground surface

settlement due to tunnelling could be used as a reference frame for the assessment

of pile settlement due to tunnelling.


30

The study confirmed that pile head and ground surface settlement can be

correlated as presented in other studies by Kaalberg et al. (2005) and Jacobsz et al.

(2001).

Jacobsz et al. (2005) reported the case studies for the construction of the

tunnels for the CTRL project in London on the effects of tunnelling on piled

foundation. Three piled bridge pier foundation are described, one with end-bearing

piles and the other two friction piles. In the case of end-bearing piles, the settlement of

the superstructure was judged to be the same as the soil (Terrace Gravels) at pile toe

level. These were estimated and the bridge structure deemed safe for the level of

movement anticipated. Figure 2.34 shows a section of the friction pile case studies

where the pile toes were very close to the tunnels. The Terrace Gravels were grouted

as a mitigation measure to increase shaft capacity at that elevation and to create a

pseudo-slab beneath the pile caps. Total surface settlements of 8 mm to 10 mm were

observed (volume loss of 0.6%) with no detrimental effects on the bridge. In the third

case, the strains along the length of the pile, both vertically and laterally (to obtain

bending strain) were estimated from the ground movements with depth assuming full

friction at the soil-pile interface. The results indicated that the piles would not be over-

stressed and that assuming that the pile movement is the same as that for the Free-field

surface settlement is conservative. No mitigation measure was implemented and no

damage was sustained for the bridge. It is recommended that the pile capacities should

be re-evaluated as there is potential redistribution of loads in the piles.

Pang et al. (2005a) presented data from part of the MRT North East Line

Contract 704 in Singapore, where forward-thinking enabled instrumentation to be


31

installed in bridge pier piles so that the influence of future planned tunnels, running

parallel to the bridge could be assessed. The data from 2 piles of a four-pile group

supporting a bridge pier are presented. The piles are 62 m long and 1.2 m in diameter

with four sets of strain gauges installed orthogonally, in pairs, to enable average axial

loads and bending moments in transverse and longitudinal directions to be determined,

as shown in Figure 2.35. The data presented related to a 6.3-m diameter EPBM tunnel,

constructed in residual soils, at 1.6m from the nearest piles at a depth of 21 m (to its

axis). The surface settlement profile due to tunnelling follows a Gaussian form with a

maximum value of about 18 mm. Correlating the developing settlements with TBM

position has enabled the volume losses related to the different phases of the tunnel

process to be identified. It is reported that the range of volume loss was between 0.32

and 1.45% only.

Information from the strain gauges within the piles reveals that the piles

experience down-drag, registered as increasing axial force, with greater force

developing in the pile nearer as might be expected. Calculations indicate that down-

drag loads were between 9 and 43% of the structural capacity of piles with peak value

occurring when the face of the TBM was in line with the piles. Post-tunneling

measurement of the development of axial force in pile P1 at Pier 20 showing the time-

dependent behavior of drag-load as indicated in Figure 2.36. Clear trends in bending

moment distributions along the length of pile are also shown, with maximum values,

although small, occurring in the close vicinity of the tunnel (see Figure 2.37). Also

evident is shielding of the outer pile by the inner pile between it and the tunnel (see

Figure 2.37). Some interesting relation between volume loss and axial force and

bending moment are also presented, showing increase in both quantities with volume


32

loss. It is also concluded that a volume loss up to 1.5% does not seem to have a

significant effect on the piles.

Cham (2007) analysed the responses of instrumented piles and ‘Free-field’

ground instrumentations to tunnelling processes such as tail void grouting, application

of face pressure, tunnel advancement and thrust force in a field study involving twin

tunnelling in MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore. The tunnel

diameters are 6m and tunnelling was carried out with 2 Earth Pressure Balance

Machines (EPBM) at a depth of 23m below the ground surface, through residual soil of

Bukit Timah Granite, Kallang Formation and Old Alluvium. Tunnelling of the second

tunnel commenced 4 months after the first and at the site of field study, the twin

tunnels have a lateral clearance of 2.5 m. Figure 2.38 illustrates the positions of

existing instrumented piles relative to tunnels.

The instrumented piles are reinforced-concrete bored piles with diameter of

either 600mm or 800mm, and of penetration depth ranging from 27m to 33m. At the

closest points, the piles lie within 1m of the tunnel lining. It was observed that down-

drag forces developed when the TBMs were directly adjacent to the piles. The

maximum induced axial force due to twin tunnel advancement ranged from 7% to 72%

of the pile structural capacity. In addition, piles experienced additional bending

moments due to tunnelling and the induced bending moment was observed to increase

with increasing volume loss.

The field studies of tunnel-soil-pile interaction help to improve the current

knowledge on for the prediction of pile responses due to tunnelling. Prior planning and


33

arrangements are made to instrument the pile as reported by Pang et al. (2005) are

strongly recommended to study the influence of future planned tunnels to existing piles.

2.4.2 Centrifuge Model Tests of Tunnelling-Induced Pile Responses

Nearly all existing field cases are limited in various aspects of measurements, which

may be due to deficiencies in instrumentation planning or difficulties in collecting field

data. Thus, centrifuge modelling is an alternative method to study the tunnel-soil-pile

interaction problem.

Loganathan et al. (2000) presented the model tunnel in-filled with oil to

simulate the uniform radial contraction in centrifuge to the pile responses due to tunnel

excavation in overconsolidated clay. The scope of the study focused on friction piles

(single pile and a 2x2 pile group). The effects of pile tip elevation relative to tunnel

axis level and volume loss on the displacements and performance of piles were

investigated to study the interaction problem. The relative position of the piles in

various tests is shown in Figure 2.39. Three tests were performed with the tunnel

located above, at and below the pie tip level. The induced bending moment and axial

force profiles at a volume loss of 1% are presented in Figures 2.40 and 2.41,

respectively. It is observed that both the induced maximum bending moment and axial

force occurred approximately at the tunnel spring line level in long pile cases where

the pile tips were below the tunnel spring line. The maximum bending moments

occurred just above the pile tips and the pile axial force increased from the pile head to

the pile tip in a short pile case with pile tip at or above the tunnel spring line level. The

comparison of the three tests showed that for single piles, the maximum bending


34

moment was the largest when the pile tip was located at the tunnel spring elevation,

whilst the maximum axial force was the largest when the pile tip was above the tunnel

spring line. It was concluded that the maximum measured bending moments vary

almost linearly with ground loss values below 5%. As such, it was postulated that an

elastic analysis may be performed to predict tunnelling-induced pile behaviour if the

ground loss value was less than 5%.

Feng (2003) performed a series of centrifuge tests to investigate the pile

responses associated with a lined tunnel in dry sand at NUS and Ran (2004) extended

the studies by Feng (2003) to clay. The simulation method of tunnel excavation

proposed by Sharma et al. (2001) using polystyrene foam was adopted and the tunnel

deformation is oval shape. Ran (2004) reported two major series of tests involving the

effects of pile-to-tunnel distance and pile length, as well as on the effects of volume

loss. Figures 2.42 and 2.43 present the typical pile axial and lateral responses over time.

From the studies, it is demonstrated that the pile responses in clay are time-dependent.

Moreover, it is evident that the induced pile responses in clay are smaller than those in

sand (Feng, 2003). The limitation in the studies is that the model tunnel simulating the

ovalisation of tunnel, and hence the induced pile bending moment demonstrates an

opposite behaviour as most practical cases.

A recent study on the influence of tunnelling on piled foundations was

completed at the University of Cambridge by Jacobsz et al. (2005). The centrifuge

model study (Jacobsz, 2002; Jacobsz et al., 2004) focused on tunnelling near driven

piles in dense sand. Both single piles and pile groups were considered and the model

was constructed at a scale of 1:75. The model tunnel comprised a 50-mm brass pipe


35

surrounded by a 1-mm thick latex rubber membrane. The 4-mm thick annulus between

the pipe and membrane was filled with water which could be discharged accurately to

impose volume losses from 0% to approximately 20% on the surrounding sand. It was

intended to use the model tunnel to impose relatively realistic plane-strain tunnelling-

related ground movements on the surrounding ground, rather than model the

progressive advance of a tunnel face and uniform radial displacement around tunnel is

simulated. A number of instrumented model piles were located at various offsets and

depths during the tests.

A zone of influence around a tunnel was established (see Figure 2.44) in which

significant base load reduction, accompanied by large pile settlements, were noted

should a certain volume loss be exceeded. Piles with their bases outside the zone of

influence did not suffer large settlements even at volume loss up to 10%. The stresses

exerted by piles on the surrounding ground result in a subsurface settlement profile

different from the Free-field situation. Pile settlements can however be approximated

by the Free-field surface settlement should the pile shaft capacity not be exceeded due

to volume loss. The piles in the centrifuge study possessed significant reserve

(immobilized) shaft capacity. Should piles not have the reserve shaft capacity, e.g.

where piles are end-bearing in sand with the shafts surrounded by soft clay, volume

loss may cause more rapid settlement. Pile groups behave in a similar fashion to

volume loss than individual piles (see Figure 2.45). Load transfer from one pile to

another within a group only occurs once the shaft capacity of a given pile has been

mobilized causing its settlement to become significant. For pile groups, these usually

occurred at large volumes which are undesirable in practice.


36

In practice, the over-excavation of the tunnel and the gap between the shield

tunnel machine and lining would cause the tunnel spring line to move inwards into the

tunnel resulting in inward soil movements towards tunnel at the tunnel spring line. Lee

et al. (1992), Loganathn and Poulos (1998) and Park (2005) had successfully proposed

analytical solutions based on inward tunnel deformation or oval-shaped deformation.

Hence, it is of great interest to improve the model tunnelling technique to study the

pile responses due to inward tunnel deformation. It should be noted that based on St

Venant’s Principle, the exact deformation shape should be immaterial if the piles are

located sufficiently far away. However, in the present study, the deformation shape of

the tunnel is of great importance as the piles are close to the tunnel and the soil

movements would significantly affect the pile responses.

2.4.3 Predictive Methods of Tunnelling-Induced Pile Responses

Chen et al. (1999) presented a simple approach to assess tunnelling induced pile

responses where a two-stage uncoupled method was introduced. In the method, free-

field tunnelling induced ground movements at the pile location is first approximated

based on the quasi-analytical method proposed by Loganathn and Poulos (1998) . The

movements were then applied to the soil elements surrounding the pile using separate

numerical programs (PALLAS and PIES) to assess the lateral and vertical pile

responses. The approach started with a basic problem as illustrated in Figure 2.46

where an existing single pile is situated adjacent to a tunnel under construction. The

induced pile responses together with the free-field soil movement of the basic problem

are shown in Figure 2.47. Subsequent parametric studies provided valuable insight into

the various factors affecting pile performance, in particular the variation of maximum


37

induced bending moment and axial force with pile-to-tunnel distance and relative

position of pile tip to tunnel axis level. In general, the maximum bending moment and

axial force values decrease to insignificant magnitudes (less than 10% of value at

X=1D) beyond a respective distance of 2D and 5D from the tunnel centre line. At a

given horizontal offset from the tunnel centre line, the pile bending moment is

generally the greatest when its tip is below the tunnel axis level, decreasing as the pile

tip moves upwards. However, the pile lateral deflection profiles are almost identical in

shape and magnitude to imposed free-field soil displacements. This is probably due to

the low flexible stiffness of the pile and the homogeneous clay profile with constant Cu

and Young’s modulus with depth used in the analysis.

Free-field displacements are movements of the soil that occur at a distance

from the pile such that the displacements are not affected by the presence of the pile. A

free-field soil displacement method, in which a pile was represented by beam elements

and the soil was idealized using the modulus of subgrade reaction, was proposed by

Chow and Yong (1996). The magnitude of soil movement profile serves as input to the

method. With this idealization, non-homogeneous soil can be easily represented. This

approach requires the knowledge of pile bending stiffness, distribution of lateral soil

stiffness and the correct limiting soil pressure acting on the pile with depth.

Comparisons with available well-documented case histories suggest that the method

gives reasonable prediction of behaviour of pile subject to lateral soil movements.

Loganathan et al. (2001) incorporated the analytical solutions for tunnelling-

induced ground movement (Loganathan and Poulos, 1998) into the computer

programme GEPAN (Xu and Poulos, 2000) to compute the response of a 2 x 2 pile


38

group as shown in Figure 2.48. Generally, the ‘front’ pile has slightly higher responses

than the ‘rear’ pile. The lateral deformation and bending moment profiles for single

piles and piles in group are almost similar, except for a small difference in bending

moment at the pile cap location due to the fixity condition. However, for single

isolated piles, the settlement is slightly higher that the piles in the group. In addition,

the axial down-drag force estimated for a single pile is about 20% higher than the

down-drag force induced in a pile within the pile group.

As the pile responses due to tunnelling is essentially a 3D problem, Pang et al.

(2005b) carried out three-dimensional finite element analysis to back-analyse a case

history on the response of pile foundation subjected to shield tunnelling as described in

Section 2.4.1. The analysis back-analysed the behaviour of an instrumented pile group

where the full shield tunnelling process, including the application of face pressure,

shield tunnel advancement, over-cutting, tail void closure and installation of lining,

was simulated (see Figure 2.49). The trend of pile group deformation was noted to

deflect towards the tunnel transverse direction. In the longitudinal direction, prior to

the tunnel arriving at the pile group, the piles deflected towards the tunnel with

maximum movement at the pile head level and not at the tunnel spring line.

Subsequently, as the tunnel advanced past the pile group, the piles were gradually

pushed forward in the same direction of tunnel advancement. In addition, the front pile

was found to be subjected to higher response of up to 2.5 times compared to the rear

pile. Good agreement between the results from analysis and measured data was

obtained which validated the FE analysis shown in Figure 2.50.


39

2.5 SUMMARY

A review of the literature to-date on effects of tunnelling on adjacent pile is reported in

this chapter. An overview of the published field studies, centrifuge model test and

predictive methods for tunnelling induced soil movements and pile responses are

presented. In addition, the short-comings from existing literature are discussed and

summarised as follows.

• Considerable research studies have been carried out to simulate the process of

tunnels excavation in centrifuge. It should be noted that all these model

tunnels could not exactly replicate the prototype tunnelling process in the

field. The techniques for simulation of tunnelling in centrifuge developed

thus far have certain limitations such as maintaining initial stress-strain

behaviour of soil prior to tunnelling, precise tunnel deformation pattern,

boundary effects of model tunnel, and the complication in implementing the

excavation process during centrifuge flight.

• Very few centrifuge studies had been carried out regarding the soil

movements and pile behaviours associated with inward tunnel deformation,

which is common in practice.

• Opposite induced-pile bending moment was observed in centrifuge tests

carried out by Ran (2004) who had simulated the ovalisation of tunnel lining.

Hence, an improved simulation of tunnelling technique should be developed.

• Limited field studies on the long-term tunnelling-induced ground movements

and pile responses. Centrifuge model tests emerge to be an alternative to

study such long-term behaviour, particularly in soft clay.


40

• Predictive methods available do not take into account the effects of pile group

or soil-structure interaction. Moreover, design charts proposed by Chen et al.

(1999) are confined to free head piles and linear elastic soil model.

• Certain important conditions necessary in the derivation of analytical

solutions may be violated (e.g. volume loss is not conserved for undrained

cases developed by Loganathan and Poulos (1998)) when empirical

assumptions are introduced. Hence, extra care has to be exercised when

employing the analytical solution.

• Although analytical solutions provide the most convenient way in predicting

tunnelling induced ground movements, these methods have certain

limitations in accounting for the effect of different ground conditions,

construction methods, post-construction ground responses and precise

tunnelling model.

The following chapters aim to present the detailed results of the present

centrifuge model study to investigate the observations and mechanisms of tunnel-soil-

pile interaction addressing some key issues raised in this chapter.


41

Figure 2.1 Simulation technique of tunnelling using high density polystyrene foam (After Sharma et al., 2001 and Feng, 2003)

Figure 2.2 Simplified tunnel lining deformation with time by simulation technique of tunnelling using high density polystyrene foam (After Ran, 2004)

Brass lining

Polystyrene foam


42

Figure 2.3 Simulation technique of tunnelling - applying compressed air (After Grant & Taylor, 2000)

Figure 2.4 Simulation technique of tunnelling - model tunnel infilled with oil (After Loganathan et al., 2000)


43

Figure 2.5 Simulation technique of tunnelling - model tunnel infilled with water (After Jacobsz, 2002)

Figure 2.6 Sand pouring in process during model preparation with model tunnel infilled with water (After Jacobsz, 2002)


44

Figure 2.7 Strong-box with two openings to fix the model tunnel in place (After Jacobsz, 2002)

Figure 2.8 Simulation technique of tunnelling - mechanical equipment of miniature shield tunneling machine (After Nomoto et al., 1994)

Opening to fix the model tunnel


45

Figure 2.9 Simulation technique of tunnelling - mechanical equipment of shield model machine (After Yasuhiro et al., 1998)

Figure 2.10 Simulation technique of tunnelling - mechanical model tunnel used to simulate the tunnel volume loss by decreasing the diameter of model tunnel under 1g

(After Lee and Yoo, 2006)


46

Figure 2.11 Gaussian curve approximating transverse surface settlement trough for MRT project C852, Singapore (After Cham, 2007)

Figure 2.12 Definition of parameters controlling tunnelling-induced settlement trough (After Standing & Burland, 2006)


47

Figure 2.13 Gaussian curve approximating transverse surface settlement trough (After Peck, 1969)

Figure 2.14 Variation in surface settlement trough width parameter with tunnel depth for tunnels in clay (After Lake et al., 1992)


48

Figure 2.15 Variation of trough width parameter K with depth for subsurface settlement profiles above tunnels in clay (After Mair et al., 1993)

Figure 2.16 Initial settlement trough at Grimsby increased from 1.5Z to a final value in excess of 4Z in long-term (After O’Reilly et al, 1991)


49

Figure 2.17 The maximum surface settlements at Grimsby increased significantly over the time (After O’Reilly et al, 1991)

Figure 2.18 The ratio of the maximum immediate settlement to maximum long-term settlement for Shanghai Metro Tunnel No.2 (After Zhang et al., 2004)


50

Figure 2.19 Normalized post-construction surface settlement troughs due to consolidation of soft clay (After Shirlaw, 1995)

Figure 2.20 Estimated trend of excess pore pressure in normally consolidated clay surrounding the tunnel (After Schmidt, 1989)


51

Figure 2.21 Changes in Pore pressure for Shanghai Metro (After Schmidt, 1989)

Figure 2.22 Change in pore pressure measured at Thunder Bay Sewer Tunnel (Adapted from data in Ng et al, 1986) (After Shirlaw et al, 1994)


52

Figure 2.23 Transverse movements in Toulouse subway line B were significantly increased over time, but stabilized after 15 days of tunnel excavation (After Emeriault

et al, 2005)

Figure 2.24 Horizontal soil movement for the Singapore’s effluent outfall pipeline in tunnel (After Balasubramanian, 1987)


53

-1.8

-1.5

-1.2

-0.9

-0.6

-0.3

0-25 -20 -15 -10 -5 0 5 10 15 20 25

Distance From Tunnel Centre-Line (m)

Settl

emen

t (m

)

Surface settlement trough in clay (measured) Surface Settlement Trough in Sand (measured)Gaussian curve (clay)Guassian curve (sand)

Figure 2.25 Comparisons of surface settlement troughs in sand (After Feng, 2003) and clay (After Ran, 2004)

-160

-120

-80

-40

0-25 -20 -15 -10 -5 0 5 10 15 20 25

Surf

ace

Settl

emen

t (m

m)

2 days

30 days

180 days

360 days

720 days

1080 days

Gaussian curve (2 days)

Gaussian curve (1080 days)

Distance from tunnel centre-line (m)

Figure 2.26 Ground surface settlement trough over time from a typical test (After Ran, 2004)


54

Figure 2.27 Normalised Vertical and horizontal soil movement profile at different subsurface elevations with best-fit curves: (a) 10mm below ground level; (b) 30mm below ground level; (c) 70mm below ground level; (d) 100mm below ground level;

(After Grant and Taylor, 2000)

Figure 2.28 Comparisons of measured surface settlement and analytical solutions (After Loganathan et al., 2000)


55

Figure 2.29 Plastic deformation mechanism for tunnels in clay (After Osman et al., 2006a)

Figure 2.30 Definition of GAP parameter (After Lee et al., 1992)


56

Figure 2.31 Oval-shaped soil displacement around tunnel boundary (After Loganathan and Poulos, 1998)

Figure 2.32 Boundary conditions of prescribed displacement (After Park, 2005)


57

Figure 2.33 Zone of influence due to Earth Pressure Balance (EPB) shield tunneling in London clay for Channel Tunnel Rail Link (CTRL) Contract 250, in Dagenham, Essex,

UK (After Selemetas et al., 2005)

Figure 2.34 A piled bridge pier foundation assessed during the CTRL project (After Jacobsz et al., 2005)


58

Figure 2.35 Typical section and instrumentation layout for pile-tunnel interaction study for MRT North East Line Contract 704 in Singapore (After Pang et al., 2005a)

Figure 2.36 Post-tunneling measurement of the development of axial force in pile P1 at Pier 20 showing the time-dependent behaviour of drag-load MRT North East Line

Contract 704 in Singapore (After Pang et al., 2005a)


59

Figure 2.37 Responses of pile foundation in terms of (a) axial force and (b) bending moment for MRT North East Line Contract 704 in Singapore (After Pang et al., 2005a)

Figure 2.38 Illustration of positions of existing instrumented piles relative to tunnels for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore. (After Cham, 2007)


60

Figure 2.39 Configuration of centrifuge tests (After Loganathan et al., 2000)

Figure 2.40 Tunneling-induced pile bending moments (After Loganathan et al., 2000)


61

Figure 2.41 Tunneling-induced pile axial loads (After Loganathan et al., 2000)

Figure 2.42 Variations of (a) induced pile bending moment profiles and (b) induced pile lateral deflection profiles with time in typical test (After Ran, 2004)

-2 0 2 4 6 8Pile lateral deflection (mm)

-25

-20

-15

-10

-5

0

Dep

th b

elow

GL

(m)

-250 -200 -150 -100 -50 0 50Pile bending moment (kNm)

-25

-20

-15

-10

-5

0

Dep

th b

elow

GL

(m)


62

Figure 2.43 (a) Induced pile axial force profile and (b) pile settlement profile at 2 days in typical test (After Ran, 2004)

Figure 2.44 Zone of influence around tunnel in which potential for large pile settlements exists (After Jacobsz et al., 2005)

0 50 100 150 200 250Pile axial force (kN)

-25

-20

-15

-10

-5

0

Dep

th b

elow

GL

(m)

5.85 5.9 5.95 6 6.05 6.1Pile settlement (mm)

-25

-20

-15

-10

-5

0

Dep

th b

elow

GL

(m)


63

Figure 2.45 Settlement, rotation and load distribution on triple pile group (After Jacobsz et al., 2005)


64

Figure 2.46 Layout of basic problem (After Chen et al., 1999)

Figure 2.47 Tunneling-induced pile responses and Greenfield soil movement (After Chen et al., 1999)


65

Figure 2.48 Numerical analysis of pile-group responses due to tunnelling (After Loganathan et al., 2001)

Figure 2.49 Typical 3D finite elements mesh to back-analyze a case history on the response of pile foundation subjected to shield tunnelling (After Pang et al., 2005b)


66

Figure 2.50 Prediction of responses of pile foundation in terms of axial force and bending moment using 3D finite element analysis (After Pang et al., 2005b)

Chapter 3 Experimental Set-up and Procedures

67

CHAPTER THREE

EXPERIMENTAL SET-UP AND

PROCEDURES

3.1 INTRODUCTION

This chapter presents the details of centrifuge modelling technique, experimental set-

up and procedures adopted in the present study. A technique of simulating tunnelling

in the centrifuge is developed and described in detail in this chapter. The preparation of

the strong box, soil specimens, fabrications and configurations of the instrumented

model piles as well as associated equipments are then elaborated. Finally, the reduced-

scale model set-up and test procedure are described in detail.

3.2 GEOTECHNICAL CENTRIFUGE MODELING

3.2.1 Principles of Geotechnical Centrifuge Modelling

Recently, there has been rapid development in geotechnical centrifuge modelling

technology world-wide, and centrifuge testing is now commonly used to study

geotechnical and geo-environmental problems. Geotechnical centrifuge modelling has

also been employed to complement conventional numerical analysis and field

monitoring (Schofield, 1998; Ng et al., 1998; Kimura, 1998). Each approach has its

own advantages in terms of quality of result, time and cost. Particularly in cases where

there are uncertainties in the applicability of a proposed design methodology, use of


68

more than one approach permits calibration of results against each other and

verification of conclusions drawn.

Conventional scaled physical models, in spite of their advantages of well-

controlled soil condition and extensive data monitoring, have significant limitations in

their usefulness as the fundamental mechanical behaviour of soil is highly non-linear

and stress-level dependent. However, by subjecting the 1/Nth scaled model in a

geotechnical centrifuge to an enhanced gravitational field N times the earth gravity, the

prototype stress can be reproduced in the reduced model, and the model test results can

be used to interpret the prototype behaviour in a rational manner.

An important principle of centrifuge modelling is to simulate the prototype

stress conditions in the model. This is done by subjecting the model components to an

enhanced body force, which is provided by a centrifugal acceleration of magnitude Ng,

where g is the acceleration due to the Earth gravity (i.e. 9.81 m/s2). Stress replication in

an Nth scale model is achieved when the imposed "gravitational" acceleration is equal

to Ng. Thus, a centrifuge is suitable for modelling stress dependent problems.

Moreover, significant reduction of test time for model tests such as consolidation time

can be achieved by using a reduced size model.

For centrifuge model tests, model scaling laws are generally derived through

dimensional analysis from the governing equations for a phenomenon, or from the

principles of mechanical similarity between a model and a prototype (Schofield, 1980,

Tan & Scott, 1985, Taylor, 1995a). Various commonly used scaling relations between

model and prototype can be deduced as summarized by Leung et al. (1991) in Table.

3.1. It can be readily deduced from Table 3.1 that the stress level of a 30-m deep clay


69

can be correctly modelled by subjecting a 30-cm deep clay model to an elevated

"gravitational" acceleration of 100g (i.e., N=100). Also, a 4-hour centrifuge modelling

at 100g can correctly simulate a prototype soil settlement problem consolidated for

more than 4.5 years (i.e., 4xN2 or 4x1002 hours). Substantial time reduction and hence

cost savings can be achieved by adopting the centrifuge modelling technique.

3.2.2 NUS Geotechnical Centrifuge Facility

Figures 3.1 and 3.2 show the NUS centrifuge facilities. The centrifuge primarily

consists of a conical case, a driven shaft, and rotating arm, and two swinging platforms.

It has a capacity of 40,000 g-kg and operates up to a maximum g-level of 200g,

implying that the allowable payloads at 200g and 100g are 200 kg and 400 kg,

respectively. The structure of the centrifuge is based on the conventional dual swing

platform design.

The model package is normally loaded onto one of the swing platforms with

the opposing platform counter balanced by either counterweights or the other model

package with identical weights. When fully spun up during test operation, the distance

from the axis of rotation to the base of the platform is 1.871 m. The centrifuge is

driven by a hydraulic motor delivering up to about 37 kW power. The swing platform

has a working area that measures 750 mm x 700 mm and headroom of 1180 mm. A

stack of electrical slip rings is mounted at the top of the rotor shaft for signals and

power transmission between the centrifuge and the control room.

DC voltage is transmitted through the slip rings to the transducers mounted on

the centrifuge or the model package from the control room. Similarly, registered


70

signals from the transducers are then transmitted via the slip rings. The signals are first

filtered by an amplifier system at 100 Hz cut-off frequency to reduce interference or

signal noise pick-up through the slip rings. The amplified signals are then collected by

a data acquisition system at a regular interval in the control room. Software called

Dasylab © is used to process the signals whereby the signals are smoothened using a

block average. Two closed circuit cameras, which are mounted on the centrifuge,

enable the entire in-flight test process to be monitored in the control room. The NUS

centrifuge is described in detail by Lee at al. (1991) and Lee (1992).

3.3 EXPERIMENTAL SET-UP

Figures 3.3 and 3.4 show the sketch and photograph of the model package for the

present study, respectively. The main features of the centrifuge model are described as

follows.

3.3.1 Model Tunnelling Technique

There are many modes of ground movement associated with tunnel construction. In a

situation where the tunnel excavation has passed a particular section is considered, the

vectors of the ground movement developed will be more or less in the plane

perpendicular to the tunnel axis. Consequently it is reasonable to assume that a plane

strain model of long tunnel section would be a good representation of tunnelling-

induced soil movements; this is usually referred to as a two-dimensional simulation

(Taylor, 1998).


71

In the present study, an innovative model tunnelling technique has been

developed to simulate the inward tunnel deformation due to over-excavation. An oval-

shape ground deformation pattern is imposed as the boundary condition and the gap

parameter (GAP) proposed by Lee et al. (1992) is used to quantify the amount of

tunnel over-cut. Loganathan & Poulos (1998) and Park (2005) evaluated that an oval-

shape deformation pattern is in reasonable agreement with tunnel deformations

observed in the field, as reported in Chapter 2.

The longitudinal and cross section of the innovative model are shown in

Figures 3.5 and 3.6, respectively. The model tunnel is made of a circular rigid outer

plate and a hollow metallic circular tube of 60 mm diameter, simulating a 6-m

diameter prototype tunnel at 100g. The rigid plate helps to maintain a uniform GAP for

the entire model tunnel. The two radial bearings inside the model tunnel help to

facilitate a smooth movement of the sliding rod and provide support to the solid

aluminium sliding rods. There are nine small rods which are inserted into the

respective holes of the model tunnel. A rigid circular plate is then used to encircle the

model tunnel and an oval-shape GAP is created between the rigid circular plate and the

point of contact of nine small rods. The whole mechanism works as such when there is

a force pushing the aluminium sliding rod, the small rods will fall onto the three

thinner parts of sliding rod of smaller cross-sectional area. As such, the GAP in cross-

sectional view will close up and this simulates the inward tunnel deformation of the

oval-shape GAP.


72

3.3.1.1 Advantages of Model Tunnel

The present model tunnel is able to simulate the precise volume loss during the process

of tunnelling. The percentage of volume loss has been calibrated by calculating the

area of surface settlement against the GAP (see section 2.3.3 )created in the model

tunnel at the undrained stage.

Moreover, the circular rigid outer plate shown in Figure 3.6 can provide a very

uniform oval-shaped of the GAP throughout the entire length of the model tunnel. This

ensures that the volume loss is constant along the model tunnel.

For the innovative mechanism created with the control of hydraulics system,

the closure of the GAP between the tunnel linings can be more effectively controlled.

With a switch of hydraulics pump, the hydraulic force can immediately push the

sliding rod inside the model tunnel forward and the small rods will then fall onto the

thinner part of the sliding rod. This causes an ‘immediate’ closure of the GAP between

the tunnel linings and simulating the volume loss.

The accuracy and repeatability of volume loss control are good, as the model

tunnel is mechanically controlled. The test has been repeated and consistent test results

are obtained as in each test, the settlement trough is measured and the volume loss is

validated.


73

3.3.1.2 Limitations of Model Tunnel

The model tunnel is a two-dimensional plane strain model. The three-dimensional

effects of tunnelling before tunnel approaching and after tunnel passing by cannot be

modelled. Nevertheless, the results of the present study are still a good representation

of tunnelling-induced soil movements as elaborated by Taylor (1998).

Owing to the constraints and difficulties faced in the model set-up in the

centrifuge, the model tunnel is pre-installed at 1g instead of in-flight tunnel excavation.

This does not simulate the real tunnel excavation process whereby the process of

tunnelling shall include excavation, installation of tunnel lining, jet grout etc. Although

the ‘rigid’ outer aluminum lining may exert some stress around lining during

centrifuge spinning up, it is believed that this effect is unlikely to be significant as the

test is properly conducted after the forced acceleration field is stabilized. The wish-in-

place model tunnel is a simplification and idealization of a cavity contraction, but with

an oval-shaped GAP and well controlled volume loss to simulate the soil movements

induced by tunnelling when the tunnel excavation has passed a particular section.

The minimum volume loss that can be simulated by the present model tunnel is

only 3% in order to maintain the accuracy of the experiment results. The accuracy of

volume control will be demonstrated in Section 4.3.2. To achieve the volume loss of

3% in 100g, a GAP parameter of 100-mm in prototype scale is needed. This means

that the GAP is as small as 1-mm in model scale. Few attempts have been carried out

to model lower volume losses in centrifuge but the results so far are not satisfactory.

Nevertheless, the magnitude of volume loss depends primarily on the method of

tunnelling and soil conditions. Although improvements in tunnelling technology have


74

significantly reduced the volume loss due to tunnel excavation, The Civil Design

Criteria for Road and Rail (LTA, 2009) recommends the contractor to demonstrate the

suitability of the selected volume loss values in relation to the values of volume loss

that would occur during tunnelling. The typical values for tunnels up to 6.6-m diameter

in marine clay are in the range of 2% to 3.5%, depending on the tunnelling method

(15% volume loss should be considered if using TBM with compressed air) (LTA,

2009). In view of the above, a volume loss of 3% is simulated in the present study. To

evaluate the detrimental effect of higher volume loss, 6.5% is also simulated.

3.3.2 Instrumented Model Piles

Two different types of instrumented model piles are used in the present study to

examine the responses piles due to tunnelling. They were fabricated using square

aluminium tubes of 9.53 mm external width and 6.35 mm internal width. Ten pairs of

strain gauges were attached along the pile shafts to measure the bending moments

along one type of pile (termed ‘bending’ pile) and axial forces along the other type of

pile (termed ‘axial pile’), see Figure 3.7. The strain gauges were protected by a thin

layer of epoxy resin for waterproofing. The final external width of each pile shaft is

12.6 mm corresponding to 1.26 m in prototype scale.

The strain gauges were wired and then connected with a TDS-300 strain meter

mounted on the centrifuge to form a full Wheatstone bridge circuit utilizing the

dummy strain gauges provided in the strain meter to produce a temperature

compensation system. The ‘bending’ and ‘axial’ piles were connected to the strain

meter with half-bridge mode and full-bridge mode, respectively. The detailed

connection principles and load-output relations were elaborated in Feng (2003). A very


75

thin and light PVC plate with smooth and dark surface was attached to the ‘bending’

pile to facilitate reflection of laser-rays, for the purpose of measuring the pile

deflection. Conical tip was chosen to minimise the deviation of the piles from the

vertical during installation.

The flexural rigidity, EI, of the model pile, is 3.97x106 kNm2 at 100g, which is

equivalent to that of a 1300-mm diameter Grade 40 concrete bored pile. It should be

noted that it is not possible to correctly simulate the pile axial rigidity, EA, and

flexural rigidity, EI of a prototype pile simultaneously. The flexural rigidity is more

crucial as the pile bending moments and lateral deflection are more sensitive. A pile

with a higher flexural rigidity tends to attract larger bending moments but a lower pile

deflection. On the other hand, the pile axial force and settlement is less sensitive to

pile axial rigidity. For example, the settlement of a loaded pile will be mainly due to

the compression of soil while the elastic shortening of the pile shaft is normally

negligible as long as the pile axial rigidity is relatively high as in the present case.

The calibration of the pile bending moment and axial force was conducted

separately prior to the tests. ‘Bending’ pile was calibrated by fastening the pile head

with a G-clamp and hanging mass centrally at the pile tip, the strain gauge outputs

were then related to the calculated bending moments. The ‘axial’ pile was calibrated

by applying incremental loads on the top of the pile resting on a digital balance. The

corresponding strain gauge outputs were then related to the axial force.

The model piles are installed in 1g and positive excess pore water pressures are

generated during installation. Hence, the reconsolidation of the soil before simulating

the tunnel excavation is deemed to be necessary in order to recover the initial stress


76

level of the soil and to allow the full dissipation of excess pore water pressure. Pore

pressure transducers are installed to monitor and ensure that the equilibrium state is

achieved before tunnel excavation.

3.3.3 Model Pile Cap

The model pile cap is made of aluminium with a thickness of 25 mm or 2.5 m thick at

100g. Two types of pile caps were fabricated for the 2- and 6-pile group configurations.

This is an improved design compared to previous pile caps fabricated at NUS

(Lim, 2001; Ong, 2004) as the pile caps were specifically designed to enable each pile

in the group to be tightened individually by tow rows of bolts in both directions. Thus,

the rotation or movement of the pile-pile cap connection can be minimized. Figure 3.8

shows the pile caps used in this study.

The prototype pile cap bending rigidity for the 2 groups is 3.24 x 108

kNm2 .

For the 6-pile group, the pile cap bending rigidity depends on the configuration of the

piles facing the tunnel excavation. If the 6-pile group consists of 3 rows of 2 piles per

row (2 piles x 3 rows) facing the excavation, the prototype EI of the pile cap is similar

to the 2-pile group case. However, in a 6-pile group of 3x2 configurations, the EI of

the cap is 1x 109 kNm2.

3.3.4 Strong Box

The strong box is made of stainless steel alloy to contain the soil specimens. It has

internal dimensions of 525 mm × 200 mm × 490 mm (length ×width ×height). One


77

sidewall of the strong box is made of a 75-mm thick transparent Perspex plate, which

allows image acquisition by a video camera mounted to the centrifuge platform. A

measuring tape is attached to the Perspex wall to provide reference co-ordinates in

order to check the depth of the clay. Both the front (Perspex plate) and back walls of

the strong box can be removed to facilitate the installation of model tunnel and

transducers during the model set-up. To minimize the soil/strong box friction, all the

inner walls of the strong box are heavily greased. This would help to ensure the

deformation of the model ground is under plane strain condition.

3.3.5 Kaolin Clay

The physical properties of Malaysian kaolin clay used in the present study are

summarized in Table 3.2 (Goh, 2003). It has a liquid limit (LL) of 80%, plastic limit

(PL) of 40 % and hence a plasticity index (PI) of 40%, and a specific gravity, Gs, of

2.65. The coefficient of consolidation Cv and permeability at pressure of 100 kPa, are

40 m2/year and 2×10-8 m/s, respectively. The effective internal friction angle, φ’, is

23o . Kaolin clay has critical state parameters λ of 0.244, average κ of 0.053, N of 3.35

and M of 0.9.

Figure 3.9 shows the measured in-flight undrained shear strength profile of the

Kaolin clay used at NUS, using miniature T-bar developed by Stewart and Randolph

(1991). The undrained shear strength profiles from the five tests are consistent and

repeatable The profile indicates an over consolidated layer down to 40 mm, below

which the shear strength increases nearly linearly with depth, and is consistent with

that for normally consolidated clay.


78

3.3.6 Toyoura Sand

The sand that underlies the clay serves as drainage channel and socket for the pile. The

physical properties of Toyoura sand are listed in Table 3.3 (Teh et al., 2005). It has an

average particle size of 0.2 mm and specific gravity, Gs, of 2.65. The minimum and

maximum density of the sand is 1335 kg/m3 and 1645 kg/m3, respectively. The critical

state friction angle is 32o.

3.3.7 Potentiometers

Potentiometers (model LP-50F-61) were used to measure the surface settlements and

pile head settlements during the tests. This model has a measuring range of 50 mm and

an independent linearity of ±0.2 %. The working part of the instrument consists of a

resistant and a rod whose stretch can alter the resistance of the resistor and hence the

output voltages. The output voltages are then linearly translated to the measured

distance. A round plastic plate is attached to the tail end of the rod to stop it

penetrating into the clay.

3.3.8 Pore Pressure Transducers (PPT)

Druck PDCR81 miniature pore pressure transducers (PPT) were used to monitor the

variations in pore water pressures during the centrifuge tests. Before the test was

carried out, the PPTs were de-aired using an electronic vacuum pump to release

trapped air bubbles in the PPTs to prevent acquisition of inaccurate readings. Each


79

PPT comes with its own manufacturer’s calibration factor and this is incorporated to

determine the magnitude of pore water pressure. To confirm the manufacturer’s factors,

a digital air pump and a multimeter were used to calibrate the PPTs. The calibration

check was conducted by pumping air into the PPTs and recording simultaneously the

air pressure as well as the PPTs output voltage readings measured by the multimeter.

3.3.9 Non-Contact Laser Transducers

NAIS micro laser sensor LM10 (model ANR1250) were used to measure the lateral

pile head deflections during and after tunnel excavation. This model of sensors has a

centre point distance (distance between sensor and target) of 50 mm and a measurable

range of ±10 mm within the centre point distance. The light source comes from a laser

diode and has a wavelength 685 nm and beam dimension of 0.6 mm x 1.1 mm at the

centre point distance. It has a linear resolution of 0.5 μm, which translates to a linear

error of 0.5 mm at prototype scale.

The laser sensor has three main components, namely, the sensing body, the

relay cable and the controller/display unit. The sensing body houses the laser diode and

its function is to emit laser beam upon connected to a power supply of 24V DC. The

relay cable connects the sensing body to the DC power supply. The controller/display

unit is used to control and set the measuring limit of the sensor.

Calibration was carried out by securely attaching a 100-mm travel

potentiometer to the sensing body of the laser sensor. The transducer was connected to

a multimeter so that the digital display of the voltage could be displayed. The

transducer could take up to a maximum of 10 V. Hence, a direct relationship between


80

displacement and voltage could be established, i.e. 1 V per 10 mm movement of the

transducer. The laser sensor has a specified optimum range of measurement to ensure

accuracy of the reading. However, readings outside this optimum range can still be

measured by the laser sensor but to a lesser degree of accuracy. Therefore, calibration

is ensured to lie only within this optimum range. As such, the transducer serves as an

indication or a ‘ruler’ for the calibration of the laser sensor. The output voltage reading

on the laser sensor display unit varies with the displacement. Each set of readings of

the transducer and the laser sensor were recorded at every specified displacement

intervals so that correlation between displacement and voltage could be established.

3.4 IMAGE ACQUISITION SYSTEM

An advanced technique of image analysis has been developed at NUS as a method of

acquiring soil movement profiles from high-solution images captured in the centrifuge

model tests. Subsequently, Particle Image Velocimetry (PIV) is used to process the

resulting images (White et al., 2003; Zhang et al., 2005).

3.4.1 High Resolution Camera

JAI ©CV-A2 progressive scan camera coupled with a Tamron © lens was mounted in

front of the Perspex window of the model container, as shown in Figure 3.10. The

camera’s maximum grabbing speed is 15 frames per second (fps). However, the

capturing rate was set at 0.5 fps during tunnel excavation and at slower rate during post

tunnelling.


81

3.4.2 Lighting System

Lighting system is important in producing high quality images for the purpose of post

processing the data. As shown in Figure 3.10, two spot lights, each with a 50 W

halogen bulb, were each mounted at the frame arm to provide uniformly distributed

lighting across the soil sample. The florescent lights inside the centrifuge enclosure

were turned off during the capturing of images while the centrifuge was in operation.

3.4.3 On-Board and Command Computers

The progressive scan camera is connected to a computer installed on-board of the

centrifuge, as shown in Figure 3.11. The on-board computer is capable of sustaining

high gravitational force without being damaged. All captured images during an in-

flight centrifuge test were stored in this on-board computer, in which the hard disk of

the on-board computer was specially designed to provide greater resilience to physical

vibration, shock and extreme temperature fluctuations.

The on-board computer was remotely controlled by another commands

computer in the control through a wireless connection to activate the capturing of live

images of the soil movement during an in-flight centrifuge test. All captured images

could be retrieved from the on-board computer after tests. Figure 3.12 displays the

picture captured by the JAI ©CV-A2 progressive scan camera, which will be

processed subsequently.


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3.4.4 Post-Processing of Images

New technique of image analysis, particularly Particle Image Velocimetry (PIV)

(White et al., 2003; Zhang et al., 2005) has been recently applied to geotechnical

centrifuge modelling. The PIV technique increases the details and precision of

deformation measurements. Take & Bolton (2004) and Zhang et al. (2005) confirmed

that this high precision can be achieved in centrifuge conditions. The GeoPIV software

(White and Take, 2002) has been used to implement the PIV technique for post

processing of the images captured during the centrifuge test.

The principles of PIV analysis are summarized in Figures 3.13 and 3.14. PIV

operates by tracking the texture (i.e. the spatial variation of brightness) within an

image of soil through a series of images. The initial image is divided into a mesh of

PIV test patches. Consider one of these test patches, located at coordinates (u1,v1) in

image 1. To find the displaced location of this patch in a subsequent image, the

following operation is carried out. The correlation between the patch extracted from

image 1 (time = t1) and a larger patch from the same part of image 2 (time = t2) is

evaluated. The location at which the highest correlation is found indicates the

displaced position of the patch (u2,v2). The location of the correlation peak is

established to sub-pixel precision by fitting a bicubic interpolation around the highest

integer peak. This operation is repeated for the entire mesh of patches within the image,

and then repeated for each image within the series, to produce complete trajectories of

each test patch. Details are presented in White and Take (2002), White et al. (2003)

and Zhang et al. (2005).


83

3.4.5 Assessment of Effectiveness of Image Processing System

The image processing technique, PIV is used to track the soil markers so that the soil

movement due to tunnelling can be quantified. For clay such as kaolin, there is not

enough natural texture (e.g. sand) for the application of PIV technique (Zhang et al.,

2005). Hence, it is necessary to provide an artifical texture on the clay surface as PIV

technique requires random and unique textures of soil patches within the images for an

accurate analysis.

A typical reconsolidation stage of a centrifuge test has been conducted to assess

the effectiveness of the image processing system, as well as the performance of flocks

and beads as material in creating artificial textures for PIV analysis purpose.

Two methods of measuring the surface settlement are adopted. The first

method is the direct measurement of the surface settlement using potentiometer and the

second method involves measuring the soil settlement by tracking the movement of

flocks and beads using the image processing method.

On the left handside of the model tunnel, the exposed clay surface was

sprinkled with black and gray flocks, while on the right handsied of the model tunnel,

1-mm diameter black/blue/red beads were randomly embedded on the surface of plain

white clay. The set-up of the test is shown in Figure 3.15. Figure 3.16 shows the

settlement measured by these methods. It is observed that the measured settlement

analysed by image processing method agrees well with the direct measurement of the

surface settlement using potentiometer, with an error less than 5%. Amongst these two

different materials, beads demonstrate a higher accuracy compared with the flocks. It


84

is probably the beads are embedded in the soil and move freely together with the soil,

but the flocks are only spread on the surface of clay and hence influenced by the

greased applied on the Perspex windows.

In view of this, the image processing analysis can be considered a competent

and reliable method to measure the soil movements in the present study, and different

colours of beads are used to create unique textures.

3.5 EXPERIMENTAL PROCEDURE

3.5.1 Preparation of the Soil Sample

The model ground was remoulded from Malaysian kaolin clay powder and water at a

weight ratio of 1 to 1.2 in a de-airing mixer. After 4 hours of mixing, the clay sample

was carefully poured into the strong box, in which a 30-mm thick sand layer is placed

at the bottom. To prevent air trap in the clay, water is poured into the strong box before

the clay is poured. Three rubber tubes are used to act as drainage for pore water of the

soil under consolidation. The lower part of the tubes is covered by the sand. A thin

geotextile was placed on top of the sand to separate the sand and clay. The clay slurry

was then carefully scooped into the strong box by immersing into water body to

prevent air trap until it reaches a predetermined height. The underlying sand and

rubber tubes together act as drainage function in this case. After that, the soil container

was shifted to a pneumatic loading frame. The clay was then pre-loaded under a

pressure of 20 kPa. During the loading process, some of the pore water may come out

from the rubber tubes, which further helps to act as drainage.


85

3.5.2 Pre-Consolidation Process

Pore pressure and soil settlement throughout the in-flight consolidation and

reconsolidation were monitored by PPTs and potentiometers. Since the position of the

PPT was fixed and the elevation of the free water was known from the potentiometer

attached with floating ball, the corresponding hydrostatic pressure at PPT can be

calculated from the difference between the two measurements.

The pore pressure and soil surface settlement of a consolidation test over 13

hours are recorded so as to optimise the duration of spinning and for the ease of the

analysis. The results are shown in Figure 3.17. The pore pressure and soil surface

settlement appear to be stabilized after 8 hours of consolidation, whereby further

analysis using Asaoka’s method (1978) depicts that the final settlement after 8 hours

approaching 100% degree of consolidation (see Figure 3.18). Hence, this can eliminate

the uncertainty in term of long term self-weight soil consolidation in the analysis of the

test data after tunnelling.

The same monitoring was carried out at the reconsolidation stage for every test

to ensure that complete consolidation was restored, as swelling of the soil sample was

noted due to stress release during the set-up installation at 1g. Based on the

observation, a total of 6 hours are required to restore the final soil elevation during the

earlier preconsolidation stage.


86

3.5.3 Installation of Model Tunnel and PPTs in 1g

Following the completion of self-weight consolidation, the back wall of the strong box

was opened. Another layer of grease is again applied to the back wall to make sure that

the clay’s boundary will act as free-roller support. The wooden tunnel installation

guide was designed in such a way that it can be used to make sure that the model

tunnel is inserted into the clay perpendicularly so as to minimize any soil disturbance.

This will enhance the accuracy of the whole experiment. The tunnel installation guide

will be fixed in position by 2 G-clamps. Along the tunnel installation guide, the

stainless steel tube (60 mm in diameter, 0.8 mm wall-thickness) was inserted into the

clay and excavates a cylindrical cavity through the two openings. The model tunnel

was then subsequently inserted into the cylindrical cavity. The model tunnel end cap is

connected to the hydraulic hose through an aluminium tube. The aluminium tube,

which was secured by a small frame, was then be connected to the hydraulic hose,

whereby the hydraulic force will be used to push the sliding rod inside the model

tunnel so as to simulate the closure of the oval-shaped tunnel gap.

Similarly, PPTs installation guide was used to ensure that the PPTs can be

carefully inserted into the clay perpendicular to the soil surface. PPTs will then be used

to monitor the change in pore water pressure throughout the entire experiment.

3.5.4 Preparation Works for PIV Analysis

After installed the model tunnel and PPTs, the back wall was then fixed back and the

front wall was then removed. As shown in Figure 3.19, different colours of 1-mm


87

diameter beads were randomly embedded on the surface to produce an artificial texture

for the subsequent analysis of PIV. These beads are made of light PVC so that they

could move with the soil freely. The beads were pushed into the soil by the highly

greased Perspex window (see Figure 3.20) of the strong box to ensure a full perfect

contact and the beads can move together with the soil. Permanent control markers dots

with known centre to centre distance were marked on the Perspex window in order to

provide reference points to the subsequent image analysis by PIV.

3.5.5 Installation of Model Pile at 1g

Using the pile installation guide, the model piles were then carefully installed at the

predetermined distance at 1g. The pile installation guide was designed in such a way to

ensure that the model pile could be inserted vertically into the clay and this also

minimizes the disturbance to the clay. The pile guide is made of Perspex.

Two transducers were placed on the pile head to measure the model pile

settlement. Two non-contact laser transducers were used to measure the lateral

deflection of the pile head. The distance between the laser transducers and the bending

pile is about 50-mm. The transducers were attached to a stainless steel holder mounted

tightly onto the top of the container.

3.5.6 Test Procedures

The completed set-up of the entire model package is shown in Figure 3.21. The model

package was then spun up to 100g in 10 steps at 5 minute intervals for reconsolidation


88

of the clay. After about 6 hours, the total pore pressure would be restored to the same

state as that in the consolidation stage and the test then began when the hydraulic valve

was switched on and the hydraulic force would push the sliding rob inside the

mechanical tunnel forward. When the sliding rod was pushed forward, the small rods

lying on the sliding rod would drop onto the thinner part of the sliding rod with smaller

cross-sectional area. This caused the gap between the rigid aluminium plate and the

model tunnel to close and the inward tunnel deformation at the tunnel spring line was

thus simulated in this way. The model tunnel was left in place to simulate the tunnel

lining to study the post-excavation ground deformation and pile responses. The

centrifuge would be kept at 100g for 3 hours after the completion of tunnel excavation.

All instruments were monitored regularly throughout the test.


89

Table 3.1 Scaling relation of centrifuge modeling (After Leung et at, 1991) Parameter Prototype Centrifuge Model at Ng

Linear dimension 1 1/N

Area dimension 1 1/N2

Volume dimension 1 1/N3

Density 1 1

Mass 1 1/N3

Acceleration 1 1/N

Velocity 1 1

Displacement 1 1/N

Stress 1 1

Strain 1 1

Force 1 1/N2

Time (viscous flow) 1 1

Time (seepage) 1 1/N2

Flexural rigidity 1 1/N4

Axial rigidity 1 1/N2

Bending moment 1 1/N3


90

Table 3.2 Physical properties of Malaysian kaolin clay (After Goh, 2003) Property Value

Liquid limit, LL 80%

Plastic limit, PL 40%

Specific gravity, Gs 2.65

Coefficient of consolidation at 100 kPa, cv 40 m2/year

Coefficient of permeability at 100 kPa, k 2×10-8 m/s

Angle of internal friction, φ' 23o

Particle size* 3.0~5.5 μm

Modified Cam-clay parameters:

M 0.9

λ 0.244

κ 0.053

N 3.35

* Manufacturer data Table 3.3 Physical properties of Toyoura sand (After Teh et. al, 2005) Property Value

Specific gravity, Gs 2.65

Average particle size, d50 0.2 mm

Particle size, d10 0.163 mm

Minimum dry density, minρ 1335 kg/m3

Maximum dry density, maxρ 1645 kg/m3

Critical state (constant volume)

Friction angle, critφ

32o


91

Figure 3.1 Schematic diagram of NUS geotechnical centrifuge

Figure 3.2 Photograph of NUS geotechnical centrifuge with the model package

mounted on the platform

Rotating arm Drive

shaft

Bearings

Conical base

(Static position)

(In-flight position)

Swing platform

Counter Weight

Slip Rings

Payload

Balance Arm

Strain Meter

On-board camera

Conical Base


92

Figure 3.3 Sketch of a typical centrifuge model package (All dimensions in mm)

Figure 3.4 Photograph of a typical centrifuge model package

Tunnel

Beads

Lasers Potentiometers

Camera


93

Figure 3.5 Longitudinal view of model tunnel set up

Figure 3.6 Cross-section of model tunnel

GAP

GAP


94

1525

2525

2525

2525

2525

2550

70

Plate to measure

Strain gauge

9.53

12.6

End cap

coatingEpoxy

Aluminumtube

deflections by lasers

40

Figure 3.7 Instrumented model pile (All dimensions in mm)

Figure 3.8 Model pile caps


95

0

2

4

6

8

10

12

14

16

18

20

22

24

0 5 10 15 20 25 30 35 40

Undrained Shear Strength (kPa)

Dep

th (m

)

TEST 1TEST 2TEST 3TEST 4TEST 5

Figure 3.9 In-flight undrained shear strength of clay

Figure 3.10 Image acquisition system

Lighting system

Camera CCTV

Acer

Highlight


96

Figure 3.11 On board set-up

Figure 3.12 Picture captured by JAI ©CV-A2 progressive scan camera for PIV analysis

On board Imaging PC

Strain meter

Wireless system

On-board Camera

Hydraulic hose for connection to tunnel

Control marker

Texture clay

Model tunnel


97

Figure 3.13 Image manipulation during PIV analysis. (After White et al., 2003)

Figure 3.14 Evaluation of displacement vector from correlation plane, Rn(s): (a) correlation of Rn(s); (b) highest correlation peak (integer pixel); (c) sub-pixel

interpolation using cubic fit over ± 1 pixel of integer correlation. (After White et al., 2003)


98

Figure 3.15 Experimental set-up for assessment of effectiveness of image processing

system and comparison of performance of flocks and beads

Figure 3.16 Results of assessment of effectiveness of image processing system and comparison of performance of flocks and beads

Flocks Beads

LVDT LVDT

Control marker

Flocks PIV -662mm Error=-4.5%

Beads PIV -690mm Error=-1.5%

Potentiometer 1-694mm

Potentiometer 2-702mm


99

Model time- consolidation (hours)

0

100

200

300

400

0:00 1:00 2:00 3:00 4:00 5:00 6:00 7:00 8:00 9:00 10:00 11:00 12:00 13:00 14:00

Tota

l por

e pr

essu

re (k

Pa)

PPT 11403 T

Hydrostatic

Model time- consolidation (hours)

0

1

2

3

4

5

0:00 1:00 2:00 3:00 4:00 5:00 6:00 7:00 8:00 9:00 10:00 11:00 12:00 13:00 14:00

Mod

el s

ettle

men

t (cm

)

Figure 3.17 Pore pressure dissipation and settlement during consolidation stage

Figure 3.18 Estimation of ultimate settlement by Asaoka’s method (1978)

0

1

2

3

4

5

0 1 2 3 4 5

Settlement Si-1 (cm)

Set

tlem

ent,

Si (

cm)

Sult= β0 / (1-β1)

β0=2.4615038β1=0.99925896

Sult= 3.325 cm

U=Sf/Sult= 3.325/3.325=100%


100

Figure 3.19 Different colours of beads were randomly embedded on the surface of clay

Figure 3.20 The Perspex window is highly greased to ensure free movement of soil

Control marker

Texture clay

Model tunnel


101

Figure 3.21 Set-up of the entire model package in 1g (top view)

Lasers

Model tunnel

Potentiometer

Model pile

Chapter 4 Basic Test on Volume Loss

102

CHAPTER FOUR

BASIC TESTS ON VOLUME LOSS

4.1 INTRODUCTION

In order to interpret the pile behaviour due to tunnelling-induced soil movement, it is

important to examine the mechanism of tunnel-soil interaction. Particle Image

Velocimetry (PIV) technique (White et al., 2003; Zhang et al., 2005) has been used in

centrifuge model tests to obtain more accurate and detailed information on soil

displacements, as described in Chapter 3. With a better understanding and results

obtained in free-field experimentally in the present study, further evaluations can then

be made on tunnelling-induced pile responses.

4.2 TEST PROGRAM

The first half of this chapter presents the results of study on free-field soil movements

due to tunnelling. The test results on the effects of tunnelling with the same volume

loss on single free head piles are then presented in the second half of this chapter. The

complete test program presented in this chapter is shown in Table 4.1. The centrifuge

tests were performed under 100g. Unless otherwise stated, the test configurations and

results are presented in prototype scale hereinafter. For all tests, the thickness of kaolin

clay layer and the underlying Toyoura sand layer is 24 m and 3.5 m, respectively. The

tunnel cover C (distance from ground surface to tunnel crown) and tunnel diameter D


103

is 12 m and 6 m, respectively. The schematic and digital image of a typical test are

shown in Figures 4.1 and 4.2, respectively.

In the present study, the terminology “short-term (ST)” refers to the stage in

which tunnel excavation has just been completed, under undrained condition. On the

other hand, “Long-term (LT)” refers to the stage when the soil has completed

consolidation due to tunnelling. After 720 days, it was observed that changes in the

ground movement and pile responses are negligible. Hence the time of 720 days after

tunnel excavation is taken as reaching the long-term stage.

As discussed in Section 3.1.3.1.2, the magnitude of volume loss depends

primarily on the method of tunnelling and soil conditions. The typical volume loss

design values for tunnels of up to 6.6m diameter in marine clay are in the range of 2%

to 3.5%, depending on the method of tunnelling (15% volume loss should be

considered if using TBM with compressed air) (LTA, 2009). In view of the above, a

volume loss of 3% (Test 1) is simulated in the present study. To evaluate the

detrimental effects of higher volume loss, 6.5% (Test 2) is also simulated.

Test 1 has been repeated to evaluate the repeatability and consistency of the test

with and without the presence of pile. During both tests (with and without presence of

pile), the beads were randomly embedded on the surface to produce an artificial texture

for the subsequent analysis of PIV. Images captured in days 2, 180, 360, 540 and 720

were analyzed. It is observed that the results are consistent for both tests as long as the

pile is installed far enough behind the Perspex window of the strong box. Both Tests 1

and 2 presented in this chapter refer to centrifuge model tests conducted without pile.


104

In addition, Tests 3 and 4 were carried out in series 1 to investigate the

behaviours of long pile (pile tip below tunnel invert) under two different tunnel volume

losses. The pile-to-tunnel distance and pile embedment length are kept constant as 6 m

and 22 m, respectively, in these tests.

4.3 TUNNELLING-INDUCED SOIL MOVEMENTS (TESTS 1 & 2)

4.3.1 Cumulative Soil Movements

The cumulative soil displacement vectors and contours at different times after tunnel

excavation can be obtained using the PIV technique. Figures 4.3 and 4.4 show the

cumulative soil movement contours and vectors over time for both Tests 1 and 2,

respectively. For ease of comparison, the contour of cumulative soil movement of 10

mm is highlighted as bold dash lines in the plots. The 10 mm movement is selected as

the bench mark because the maximum allowable settlement for shallow foundation as

specified by the Civil Design Criteria for Road and Rail (LTA, 2006) is 20 mm and 10

mm is often set as the alert level. In the short-term, principal soil movements are

concentrated within a zone indicated in Figure 4.3(a). This zone may be identified as

the ‘Immediate Shear Zone’ as the soil within this zone has likely been ‘unloaded’ due

to tunnel excavation. For clay, the soil does not settle as a rigid body but gradually

deforms by arching, causing the radial stress in the immediate shear zone to be reduced

due to stress relief. This leads to the observed soil movement pattern and the settlement

trough at the ground surface. On the other hand, the zone outside the immediate shear

zone may be identified as the ‘Support Zone’, as the circumferential soil stresses

increase within this zone to support the arches formed in the immediate shear zone.


105

Qualitatively, it is expected that volumetric soil strain in the long-term would

increase due to soil consolidation. This might cause the soil movements to increase in

both the horizontal and vertical directions, as observed in Figures 4.3 to 4.4. It can be

observed from the figures that the shear zone propagates with time and becomes wider

over time due to post-tunnelling soil reconsolidation. This observation demonstrates

that with relatively large volume loss (>3%), the long-term effects of soil movement

induced by tunnelling could be substantial.

4.3.2 Soil Surface Settlement Troughs

The surface settlement trough along a plane transverse to the tunnel can be described

by the Gaussian curve (Peck, 1969). The popularity of the Gaussian curve as a

prediction tool for the magnitude of surface settlement due to tunnelling lies in its

simplicity and efficiency. The surface settlement curve, S, is given in Equation 2.1.

Figures 4.5 and 4.6 show the measured surface settlement troughs over time

obtained from PIV and potentiometers with volume loss of 3% and 6.5%, respectively.

It is evident that in the short-term (2 days), the surface settlement troughs follow a

Gaussian distribution curve with a maximum ground surface settlement of 41 mm for a

volume loss of 3% and 92 mm for a volume loss of 6.5%. This corresponds to the

respective imposed volume loss with simulated tunnel opening of approximate GAP =

100 and 200 mm (refer to definition of GAP in Section 2.3.3). Based on the above

findings, there is further evidence that the accuracy of volume control of the model

tunnel is good and reliable. As expected, the volume loss at the ground surface is close

to the tunnel volume loss under such undrained condition. The point of inflection, i, is


106

determined from the settlement trough at the point when the change of gradient is zero.

The point of inflection, i, of the settlement trough is determined to be approximately

7.5 m for both tests. This value is identical to the prediction of 7.5 m by Peck (1969),

using a trough width parameter k of 0.5 suggested by Mair et al. (1993) for tunnels in

clay. Thus it can be established that the observed settlement trough in the short-term

can be reasonably predicted using existing methods.

In the long-term, the ground settlement continues to increase with time, as

shown in Figures 4.5 & 4.6. It should be noted that the soil has practically completed

its self-weight consolidation before tunnel excavation, as illustrated in Figure 3.19.

The remaining self-weight consolidation settlement of the soil should be very small.

Hence the long-term soil movement is mainly due to stress relief of clay due to tunnel

excavation and the settlement trough S are noted to become wider over time. Shirlaw

(1993) presented case studies of long-term settlements and reported that the ground

settlement due to tunnelling and the extent of settlement trough can increase

significantly in the long-term in some cases. In contrast, Gaussian distribution curve is

found to be inappropriate for representing the long-term surface settlement trough with

a wider parabolic shape. Nevertheless, although the magnitude of maximum long-term

ground settlement is larger, the differential settlement for a wider settlement trough is

not as significant as that in the short-term.

Figure 4.7 clearly shows that that the maximum surface settlements measured

from potentiometers and derived from PIV match well, confirming the accuracy and

effectiveness of the PIV image processing technique. Similar good comparisons


107

between the settlement measurements obtained by potentiometers and PIV are

demonstrated in Figures 4.5 & 4.6 as well.

The maximum surface settlement often occurs at the tunnel crown for a single

tunnel case. Figure 4.7 demonstrates that the maximum surface settlements for both

Tests 1 & 2 increase over time. The rate of increase in settlement is significantly

reduced after a period of 360 days and becomes very small after approximately 720

days. This finding implies that for tunnel in soft clay with relatively large volume loss,

the surface settlements in the field should be monitored for the first 1 to 2 years after

tunnelling.

4.3.3 Subsurface Vertical Soil Movements

The vertical soil movements can provide clues on the mechanisms associated with

tunnel-soil-pile interaction, particularly on the induced pile axial forces and settlements.

Figures 4.3(a) and 4.4(a) indicate that in the short-term (ST, 2 days), the largest

vertical soil movements are spotted in the immediate shear zone above the tunnel.

However, this zone becomes wider in the LT as shown in the contour plots over time

in Figures 4.3(b) to (e) and 4.4(b) to (e). The propagation of vertical soil movement

trough seems to be an inverted ‘half-ripple’. This large vertical deformation zone is

critical and must be taken into consideration.

Figures 4.8 and 4.9 show the ST surface and subsurface settlement troughs for

Tests 1 and 2, respectively, in comparison with existing predictive methods proposed


108

by Mair et al. (1993) and Loganathan and Poulos (1998). Mair et al. (1993) proposed

that at a depth z below the ground surface, and above a tunnel depth of zo, the trough

width parameter for tunnels constructed in clays are given by Equations 2.3 and 2.4.

The solution of vertical displacement around a tunnel excavation proposed by

Loganathan and Poulos (1998) is given in Equation 2.7. It is noted that the method

proposed by Mair et al. (1993) yields a better prediction as compared to the method

proposed by Loganathan and Poulos (1998). In addition, the influence zone predicted

by Loganathan and Poulos (1998) is much greater than the measured data and

prediction by Mair et al. (1993). Hence, care should be exercised when employing

quasi-analytical methods to predict soil displacements due to tunnelling as certain

conditions in the derivation of analytical solutions may not be valid (e.g. volume loss

may not be conserved (Loganathan and Poulos, 1998)).

For the subsurface settlement troughs at various depths, the subsurface

settlement profiles generally follow the prediction by Mair et al. (1993). It should be

noted that the maximum subsurface settlements measured in the experiments,

especially when close to the tunnel, may not be accurate. This is mainly due to the

over-sizing of the tunnel end cap to prevent water seepage. The over-sized tunnel end

cap greatly influenced the tracking of soil displacements which were subsequently

analysed by PIV. Despite the above shortcoming, the back analysis generally validates

the use of Mair et al.’s (1993) method to predict the subsurface settlements in the

short-term.

Figure 4.10 compares the measured short-term and long-term inflection point, i

at different depths with the empirical method proposed by O’Reilly and New (1982)


109

and Mair et al (1993). Both methods are based on Equation 2.3, but O’Reilly and New

(1982) assumed K=0.5 while Mair et al (1993) assumed that K varies with depth as

given by Equation 2.4.

For the present tests in clay, the measured distribution of inflection point (i)

with depth can be described by a straight line and the results are consistent with the

prediction of Mair et al (1993). The distribution of inflection point (i) with depth in

clay can be simplified as Equations 4.1 and 4.2 for the method proposed by O’Reilly

and New (1982) and Mair et al (1993), respectively.

Zo-Z = 2i (4.1)

Zo-Z = 3i-8 (4.2)

In the present study, a finding on the long-term behaviour is established

revealing that the distribution of inflection point (i) with depth in clay can be

reasonably approximated by a straight line parallel to Equation 4.2 that yields

convergence points lower than those in the short-term. Hence the proposed distribution

of the inflection point (i) with depth in clay in the long-term can be presented as

Zo-Z = 3i-12 (4.3)

It is worth examining more closely the relationship between inflection point (i)

with depth in the short-term and long-term. Figure 4.11 shows the ratio of iLT / iST at


110

different depths. It can hence be deduced that iLT is approximately between 1.21 to

1.29 times iST.

4.3.4 Subsurface Horizontal Soil Movements

The short-term and long-term lateral soil movements at various distances from tunnel

centre-line are plotted in Figures 4.12 & 4.13. The proportion of horizontal to vertical

movements at the surface is considerably greater than that at greater depths, especially

when the distance from the tunnel centre-line increases. This observation is similar to

the finding obtained from the centrifuge model tests conducted by Grant and Taylor

(2000).

As expected, the horizontal soil movement caused by tunnelling diminishes

with increasing distance away from the tunnel. It is noted that the lateral soil

movements form a bulb shape at the tunnel spring line. However, the soil movements

diminish rather rapidly in the horizontal direction and become negligible at distance of

approximately 1.5D from the tunnel circumference, i.e 12 m from tunnel centre-line.

The results from the analytical solution proposed by Loganathan and Poulos

(1998) are also presented in the figures. However, the predictions by Loganathan and

Poulos (1998) do not agree well with the measured data. This may be attributed to the

condition that volume loss has not been conserved for undrained cases in their

formulation and other factors.


111

4.3.5 Qualitative Assessment on Excess Pore Pressure Responses

Pore water pressure changes in the ground are monitored using pore pressure

transducers (PPTs) during Test 1, as described in Chapter 3. To minimize the effect of

reinforcement that the PPTs have on the ground, only 2 PPTs were used, of which one

PPT is located within the immediate shear zone and the other one is located outside the

zone. Figure 4.14 shows the schematic location of the PPTs placed in the clay near the

tunnel lining and the trend of the pore water pressure changes obtained from the PPTs

throughout the test.

For the first 50 minutes of the test, the pore water pressure increases in 10 steps.

This is because the acceleration of the centrifuge from 0g to 100g is divided into 10

steps with an interval of 5 minutes per step. Subsequently, the pore water pressure

starts to drop and stabilize. This is because the excess pore water pressure induced by

the increased acceleration field dissipates. This process continues until the effective

stress in the ground is equivalent to the preconsolidation pressure. At this state, the soil

sample is normally consolidated. As PPT1 is at a higher elevation, the initial pore

pressure at PPT1 is lower than that at PPT2.

Tunnel excavation causes stress relief on the clay surrounding the tunnel lining

and thus a sharp drop in the pore water pressure is observed immediately after the

tunnelling process for PPT1 which is located inside the immediate shear zone.

Subsequently, the pore water pressure gradually increases over time due to dissipation

of pore water pressure. In contrast, an opposite trend is observed for PPT 2 located

outside the immediate shear zone. It is observed that additional excess pore pressure is


112

being induced in the clay, as indicated by a sharp increase in pore water pressure

immediately after tunnel excavation, caused by the shearing process of the affected soil

due to soil arching.

Further observation suggests that the pore water pressure stabilizes about after

two and a half hours (720 days in prototype scale) after the tunnel excavation. This

observation shows that the excess pore water pressure due to tunnelling has practically

fully dissipated and approaches the steady state pore pressure.

The above changes in the pore water pressure regime once again confirm that the

behaviour of clay can be time-dependent due to low permeability of the clay sample.

The soil will continue to deform with time as a result of dissipation of excess pore

pressures. This observation reiterates the importance of studying the long-term

behaviour of tunnelling-induced soil movement and pile responses for tunnels with

relatively large volume loss.

4.4 TYPICAL TUNNELLING-INDUCED PILE RESPONSES

The detailed results of Test 3 are reported here as an illustrative example of test results

from the beginning to the end of a typical test. For the sign convention used in the

present study, positive lateral soil movements refer to soil movement towards the

tunnel. Likewise, the deflection of pile towards tunnel is taken as positive. Bending

moment inducing pile shaft curvature towards the tunnel is considered as positive.

Lastly, downward vertical movement is regarded as positive.


113

In order to study the post-excavation ground deformation and pile responses,

the centrifuge would be kept at 100g for 3 hours after the completion of tunnel

excavation. All instruments were monitored regularly throughout the entire test. In

Test 3, both the axial and bending piles used are long floating piles (pile tip lower than

tunnel invert) with free heads and tips. The ultimate axial capacity of the 22-m long

model pile is estimated to be 2300 kN in soft clay having an almost linearly increasing

undrained shear strength profile, as shown in Figure 3.9. The ultimate bending moment

capacity of the pile is determined to be 3000 kNm.

4.4.1 Induced Axial Force and Settlement

The induced pile axial forces are directly measured from the readings of semi-

conductor strain gauges installed along the piles. The four semi-conductor strain

gauges are bonded on the external surface of the aluminium rod at an appropriate

elevation to form a full-bridge configuration to measure the induced axial force on the

pile. A strain meter is used to record the strain gauge signals along the model pile.

Figure 4.15(a) shows the induced pile axial force profile at 2 days, 180 days,

360 days and 720 days after tunnel excavation. It is noted that the induced pile axial

force increases with depth and reaches a maximum value approximately at the tunnel

spring elevation, after which the induced axial force gradually decreases till the pile tip.

The observed trend is consistent with the field data reported by Pang et al. (2005a) for

the MRT North East Line Contract 704 and 3D finite element analysis by Cheng

(2003). It is evident that the settling soil drags the pile down and induces negative skin

friction on the pile. This is consistent with the observed downward vertical soil


114

movement above the tunnel spring line due to soil over-cut in the process of tunnelling,

as observed in Figure 4.16. The neutral plane elevation becomes deeper over time. The

plot of maximum pile axial force with time shown in Figure 4.17 (a) reveals that the

drag load along the upper pile shaft increases with time and reaches a maximum

magnitude after about 720 days. This observation is consistent with that for Test 1

where the soil settlement does not increase further after 720 days (Fig. 4.7). The

readings reveal that there is a noticeable increase in maximum axial force in the long-

term, from 198 kN after 2 days to 370 kN after 720 days. The total increment is about

90%.

Figure 4.15(b) shows field data reported by Cham (2007) for MRT Circle Line

Stage 3 Contract 852 in Singapore and the results are compared with those from Test 3

(Fig. 4.15(a). Although the general configuration of tunnel-pile and soil condition for

both cases is not identical, the general induced axial force profiles are consistent for

both cases. The soil settlement due to tunnelling would induce negative skin friction on

the pile shaft and maximum negative skin friction occurs at the tunnel axis. It can also

be observed that a smaller increase in down-drag force acts on the upper 20m of the

pile shaft. This is an indication of the effectiveness of the de-bonding system in

Contract 852. As non-zero value is observed near to the head of each pile, the

tunnelling-induced pile settlements could have resulted in some re-distribution of

structural loads on the piles after tunnel advancement. This is possible as the piles were

connected by transfer beams and slab. However, , the down-drag forces measured near

to the pile head are negligible in Test 3.

Figure 4.16 shows the observed free-field vertical soil movement profile over

time at the pile location obtained from Test 1 and the corresponding pile head


115

settlement shows the relationship between soil movement and pile settlement. From

the subsurface soil settlement observed through the marker beads movement analyzed

by PIV in Test 1 and the measured pile settlement in Test 3, it can be deduced that in

the short-term, the neutral plane is at a depth of about 14.2 m with subsurface soil

settlement very close to the pile settlement. However, in the long term, the neutral

plane shifts to a lower depth of 16.1 m. The results also illustrate that the measured

pile head settlement for Test 3 increases substantially from 6 mm in the short-term to

17 mm in the long term. These data prove that the pile responses are time-dependent,

as shown in Figure 4.17(b). The significant increase in pile settlement is likely due to

the pile tip floating in the soft clay. Nevertheless, the pile undergoes much smaller

settlement than the soil. It is observed that the pile continues to settle after the

completion of tunnel excavation until long-term ground movement has been stabilized

at about 720 days after tunnel excavation.

4.4.2 Induced Bending Moment and Deflection

The induced pile bending moments are determined from the readings of strain gauges

attached along the pile shaft with the pile calibration conducted at 1g. The strain

gauges were connected to the strain meter with half-bridge mode and were properly

calibrated so that the relationship between microstrain, με (registered by the strain

gauges) and the induced pile bending moment (result of hanging dead weights at the

tip of the cantilevered pile) could be established. Two appropriate faces of the

instrumented square pile were calibrated. As such, during centrifuge experiments, the

microstrain readings acquired from the data acquisition system can be readily related


116

to its corresponding bending moment from the appropriate scaling law of centrifuge

modelling.

Figure 4.18(a) shows the induced pile bending moment profiles at 2 days, 180

days, 360 days and 720 days after tunnelling. It is shown that the pile bending moment

increases with depth and the maximum induced bending moment toward the tunnel

occurs approximately at the tunnel central axis for long piles with tips well beneath the

tunnel. As expected, the bending moment at the pile head and tip are both zero as they

are not restrained.

The experimental results show that in the short-term, tunnel excavation induces

a maximum bending moment of 47 kNm. The induced bending moment increases

significantly by 98% to 93 kNm after 720 days. It is also noted that the shape of the

bending moment profile remains fairly constant over time and the maximum bending

moment remains practically unchanged beyond 720 days after tunnel excavation

Figure 4.17(c), revealing that the induced bending moment has stabilized. The findings

are consistent with the trend of soil movements induced by tunnelling as discussed in

Section 4.3.

The field measured bending moment profiles reported by Cham (2007) plotted

in Fig. 4.18(b) are compared to those obtained from Test 3. In addition, the measured

bending moment responses of a pile in pier 20 for C704 NEL (Pang, 2005) are also

included in Figure 4.18(b) for comparison. It is noted that the bending moment profiles

reported by Cham (2007) and Pang (2005) are similar to those observed in Test 3. In

general, the maximum transverse bending moment is noted to be at the tunnel axis

level.


117

The pile head deflection at the ground surface is obtained by geometry from the

two displacement readings obtained at 2 different pile elevations above the ground.

The free-field lateral soil displacements can be obtained from Test 1 whereas the pile

lateral displacement profile can be obtained by integrating the bending moment

profiles twice with two specified boundary conditions using the measured pile head

displacements at 2 elevations for the present study. The lateral pile displacement is

related to the free-field lateral soil movements, pile bending stiffness and pile-soil

interaction. A comparison of pile and free-field horizontal soil displacements for the

tests is shown in Figure 4.19. As expected, both the pile and soil move towards the

tunnel.

It is evident that the lateral soil movement profile has a roughly similar trend as

the pile deflection profile, showing that the pile basically deforms with the soil in a

similar fashion. As the pile can be considered a rigid body, the magnitude of pile

deflection is much smaller than that of the soil. The pile deflection profile is also

‘smoother’ than the soil movement profile due to the large pile bending rigidity. For

Test 3, the magnitude of pile deflection increases with time and the maximum lateral

pile deflection occurs at the pile head having a magnitude of 5 mm in the short-term

and 12.1 mm in the long-term.


118

4.5 TEST SERIES 1 - EFFECTS OF VOLUME LOSS (TESTS 3, 4)

In Test 4, the simulated tunnel opening for this model has a GAP of approximately 200

mm in prototype scale with an equivalent imposed volume loss of 6.5%, as compared

to that of 3% for Test 3. The pile-to-tunnel distance and pile length are kept constant in

these tests.


Figure 4.20 compares the induced pile axial force due to tunnel excavation in Tests 3

and 4. Generally, the maximum induced axial force on the pile increases with tunnel

volume loss and time. It is observed that the pile axial load transfer profiles along the

piles in Tests 3 and 4 are similar in trend, with a larger magnitude of negative skin

friction for the test with a larger volume loss. For Test 4 with a volume loss of 6.5%,

the maximum negative skin friction is 265 kN in the short-term (2 days after

excavation) and 422 kN in the long term (720 days after excavation). When compared

with Test 3, the increment in negative skin fraction is only 34% in the short-term and

14% in the long term despite an increase in volume loss of over 100% (i.e. from 3% to

6.5%).

Figure 4.21 shows the observed short- and long-terms free-field vertical soil

movement profile at the pile location obtained from PIV analysis from Tests 1 & 2,

and the measured pile head settlements from Tests 3 & 4. The results illustrate a

significant increase in pile head settlement when the volume loss increases from 3% to

6.5%. In the short-term, the pile head settlement increases significantly from 6 mm to


119

17 mm (increment of 183%) and drastically from 14.7 mm to 51.3 mm (increment of

250%) in the long-term when the volume loss increases from 3% to 6.5%.

The moderate increment of negative skin friction as compared to drastic

increment in settlement may be due to the fact that 3% volume loss has already

induced substantial relative pile-soil movement for the full mobilisation of negative

skin friction. As such, further ground settlement does not induce further negative skin

friction significantly when volume loss increases. For the induced base resistance in

the short- and long-terms, the pile base load obtained from Test 4 is higher than that of

Test 3. This is because the moderate increase in down drag has to be resisted by the

base resistance. In additional, when vertical soil movement increases with a larger

volume loss, additional pile settlement is necessary to mobilise sufficient positive shaft

resistance and base resistance to maintain pile equilibrium.

4.5.2 Induced Bending Moment and Deflection

Figure 4.22 shows the short- and long-terms induced pile bending moment profiles

obtained from Tests 3 and 4. The results demonstrate that the induced pile bending

moment profile has a double curvature with the moment magnitude increasing over

time. The bending moments at the pile head and tip are zero as they are not restrained.

The maximum induced bending moment occurs approximately at the tunnel springline.

The trend of the bending moment profile remains fairly constant over time and the

maximum bending moment remains practically unchanged after 720 days of tunnel

excavation. Owing to tunnel over-cut, the soil moves towards the tunnel spring

elevation, resulting in the pile bending towards the tunnel.


120

In the short-term, the maximum induced bending moment on the pile is 47

kNm, and increases to 93 kNm after 720 days for Test 3. The pile bending moment for

Test 4 is larger than that of Test 3 at all times, especially in the long term due to a

higher volume loss and hence larger soil movements. The maximum induced short-

term pile bending moment of 136 kNm from Test 4 is almost triple of that observed in

Test 3. The long term pile bending moment of 316 kNm is more than triple of that

observed in Test 3. It is evident that the maximum induced pile bending moments

increase significantly when the volume loss increases from 3% to 6.5%.

Figure 4.23 shows the observed short- and long-terms free-field lateral soil

displacement profiles at the pile location. The displacement profiles are obtained from

the PIV analysis of Tests 1 and 2, while the corresponding measured pile deflection

profiles are from Tests 3 and 4. The results reveal that the pile deflection increases

with volume loss, due to increase in lateral soil movement with larger volume loss and

time. As expected, the pile deflection is much smaller than that of the soil. The largest

lateral soil movement occurs at the pile head location, hence induces the largest pile

deflection at this elevation. The magnitude of lateral soil movement decreases with

depth and so did the pile deflection. The pile deflection profiles are similar for both

tests and the pile moves toward the tunnel, with the largest pile deflection observed at

the pile head. This deflection profile is due to increasing soil stiffness with depth.

Besides that, both the lateral soil movement and lateral pile deflection increase with

time. Comparing the pile deflection with horizontal soil movement, the pile deflection

has a smaller magnitude due to the large bending stiffness of the pile. Hence, the

measured pile head deflection in Test 3 is only 5 mm in the short-term and 12.1 mm in

the long-term, as compared to a much larger pile head deflection of 10 mm in the shor-


121

term and 28 mm in the long-term for Test 4. The increment of pile deflection is

significant when the volume loss increases from 3% to 6.5%.

Figures 4.24(a) to (d) show a summary of maximum pile axial force, pile head

settlement, pile bending moment and pile head deflection with volume loss for Tests 3

and 4. A consistent trend is observed that all pile responses increase with volume loss

and time. Nevertheless, it is observed that in the present floating pile condition, the

excessive pile movement (settlement and deflection) are the critical pile responses. On

the other hand, the significant increment of bending moment under a large volume loss

is detrimental to the structural integrity of the pile, especially if the pile foundation

supported the existing building only designed to resist the compression load as

illustrated in Figure 1.2. Thus, it is important to keep the tunnel volume loss as small

as possible in order that the long term induced pile responses are not a concern.

Figure 4.25 shows the long-term to short-term ratio of pile responses (pile axial

force, pile bending moment, pile head settlement, pile head deflection) for the two

volume losses. The results reveal that the pile responses increase over time and the

ratio of all long-term/short-term (LT/ST) pile responses except axial force also

increases with volume loss.


122

4.6 CONCLUDING REMARKS

4.6.1 Tunnelling-Induced Soil Movement

The centrifuge model tests with the application of PIV have provided useful data to

examine the patterns of soil movements induced by tunnelling in soft clay. The main

aim is to investigate the induced soil movement patterns over time.

The surface settlement trough in clay generally follows the Gaussian

distribution curve in the short-term. The magnitude of maximum ground surface

settlement increases with time and tunnel volume loss. The settlement magnitude is

larger in the long-term and the settlement trough is wider as compared to that in the

short-term. The data confirmed that the empirical equation proposed by Mair et al.

(1993) is applicable in the prediction of the subsurface settlement troughs in clay in the

short-term. Empirical equations in the short-term and long-term were proposed for the

distribution of inflection point in soft clay. On the other hand, an immediate shear zone

with large soil movement above the tunnel can be identified in the short-term. In the

long term, the significant soil movement zone extends much wider. In addition, soil

settlement is noted to be more dominant than lateral soil movement in the long term.

Qualitative assessment on the excess pore pressure responses has provided an

understanding on the development of negative excess pore pressure in the immediate

shear zone and positive excess pore pressure in the support zone.


123

4.6.2 Tunnel-Soil-Piles Interaction

Two centrifuge model tests with different volume loss have been performed to

investigate the effects of tunnelling on single free head piles. Owing to downward soil

movement, negative skin friction and bending moment are induced on the pile and the

magnitude of negative friction and moment increases with time and tunnel volume loss.

Both the neutral plane and maximum induced pile moment take place about the tunnel

centre-line elevation. The lateral soil movement profile has a similar trend as the pile

deflection profile, but the magnitude of the pile deflection is much smaller than that of

the soil. The pile deflection profile is ‘smoother’ than the soil movement profile due to

the large pile rigidity.

Test Series 1 studies the effects of volume loss on pile performances. The test

results shed light on the actual performance of single floating pile due to tunnelling

with volume loss of 3% to 6.5%. It is found that the induced pile bending moment

triples and the pile settlement and deflection increase by almost 2.5 times when

volume loss increases from 3% to 6.5%, in this particular case.

Chapter 4 Basic Tests on Volume Loss

124

Table 4.1 Test program and parameters for the basic tests on volume loss

Phase 1 (Free-field soil movement) -Effects of volume loss

Test No. Configuration Common parameters Individual parameters

1 Volume loss =

3%

2

C = 12 m

D = 6 m

Volume loss =

6.5%

Phase 2 -Effects of tunnelling on single piles

Test series 1 Effects of volume loss

Test No. Configuration Common

parameters Individual parameters

3

Typical Volume loss = 3%

4

C = 12 m

D = 6 m

L = 22 m

X= 6 m

Volume loss = 6.5%

XD

C

L

‘Axial’ Pile‘Bending’ Pile

2m

D

C

Kaolin clay

Toyoura Sand

24m

3.5m

24m

3.5m


125

Figure 4.1 Schematic of viewing area in tunnel-soil interaction tests (all dimensions in mm)

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Figure 4.2 Example of digital images taken during test for PIV analysis

30

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520

Water

Kaolin Clay

Toyoura Sand

Control marker

Model Tunnel

Control marker

Texture clay

Model tunnel


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2 days (from PIV)90 days (from PIV)180 days (from PIV)360 days (from PIV)720 days (from PIV)2 days (from Potentiometer)90 days (from Potentiometer)180 days (from Potentiometer)360 days (from Potentiometer)720 days (from Potentiometer)2 days (Gaussian curve)

inflectionpoint, 'i"

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Set

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Volume loss = 6.5%

Figure 4.5 Surface settlement troughs over time (Test 1) Figure 4.6 Surface settlement troughs over time (Test 2)


135

-300-280-260-240-220-200-180-160-140-120-100

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00 180 360 540 720 900 1080 1260 1440

Time (days)

settl

emen

t (m

m)

Test 1 (from PIV)Test 1 (from Potentiometer)Test 2 (from PIV)Test 2 (from Potentiometer)

Figure 4.7 Maximum surface settlements over time (Tests 1 & 2)


136

(a) (b)

Figure 4.8 Settlement troughs at surface, 4.3m and 9.3m depths (Test 1): (a) comparing with Mair et. al (1993) (b) comparing with Loganathan and Poulos (1998)


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Surface settlement, Test 1 (PIV)

Surface settlement, Mair et al (1993)

4.3m, Test 1 (PIV)

4.3m, Mair et al (1993)

9.3m, Test 1(PIV)

9.3m, Mail et al (1993)

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Surface settlement, Loganathan & Poulos (1998)

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Surface settlement, Mair et al (1993)

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5m, Mair et al (1993)

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10.9m, Mair et al (1993)

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Surface settlement, Loganathan & Poulos (1998)

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t (m

m)

Figure 4.9 Settlement troughs at surface, 5m and 10.9m depths (Test 2): (a) comparing with Mair et. al (1993) (b) comparing with Loganathan and Poulos (1998)

Short-term (VL=6.5%) Short-term

(VL=6.5%)


138

-20

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-4

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8

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0 1 2 3 4 5 6 7 8 9 10 11 12

i (m)

O'Reilly and New (1982)Mair et al.(1993)Test 1, 2 days (ST)Test 2, 2 days (ST)Test 1, 720 days (LT)Test 2, 720 days (LT)Proposed long-term equation

Zo-Z

(m)

Proposed long-termequiation, Zo-Z = 3i-12

Figure 4.10 Distribution of inflection point ‘i’ with depth in short- and long-term (Tests 1 & 2)

0.6

0.7

0.8

0.9

1

1.1

1.2

1.3

1.4

1.5

1.6

0 1 2 3 4 5 6 7 8 9 10 11 12

Depth, Z(m)

Test 1 (PIV) Test 2 (PIV)

Rat

io o

f i LT

/ iST

Figure 4.11 Comparison of ratio of iLT/iST at different depths (Tests 1 & 2)


139


4m from tunnel

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Loganathanet al 1998

720 days

Soil movement (mm)

6m from tunnel

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Tunnel


140


4m from tunnel

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141

-60-40-20

020406080

100120140160180200220240

0 500 1000 1500 2000 2500 3000 3500 4000 4500 5000 5500 6000 6500

Time (day)

Exce

ss p

ore

pres

sure

(kPa

)

PPT Point 1PPT Point 2

A

B

C

DE

F

Point1Point2

Point 1 (shear zone)

Point 2 (support zone)

4)Post-tunnelling

5)Spinningdown

2) Consolidation at 100g1)Spinning upfrom 1g to 100g

3)Tunnelling

Figure 4.14 Pore pressure changes due to tunnelling (Test 1)


142

Axial Force (kN)

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00 100 200 300 400 500 600

Dep

th (m

)

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360 days 720 days

Tunnel

ApproximateNeutral Plane

Figure 4.15(a) Tunnelling-induced pile axial force (Test 3, 3% free-head floating long pile)

Figure 4.15(b) Tunnelling-induced pile axial force (Pile BP1-G) for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore (After Cham, 2007)

Volume loss = 0.7% Tunnel ID=5.8m Pile dia. =600mm Pile Length =31m


143

Settlement (mm)

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00 20 40 60 80 100 120

Dep

th (m

)

2 days 180 days

360 days 720 days

Tunnel


Pile head settlement (Test 3)

Free-f ield soil settlement (Test 1, PIV)

Figure 4.16 Tunnelling-induced pile head settlement (Test 3) and observed free-field soil movement at pile location (Test 1, PIV)


144

Figure 4.17 Tunnelling-induced (a) maximum pile axial force (b) maximum pile head settlement and soil surface settlement (Test 1) (c) maximum pile bending moment (d)

maximum pile head deflection and soil surface lateral movement (Test 1)

150

200

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0 180 360 540 720 900 1080 1260 1440Time (days)

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. pile

axi

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Pile head settlement

Settl

emen

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202530

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Free-f ield lateral soil movement (Test 1)

Pile head deflection

Late

ral m

ovem

ent (

mm

)(a)

(b)

(c)

(d)


145

Bending Moment (kNm)

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Tunnel

Figure 4.18 (a) Tunnelling-induced pile bending moment (Test 3, 3% free-head floating long pile)

Figure 4.18 (b) Tunnelling-induced pile bending moment (Pile BP2-E) for MRT Circle Line Stage 3 (CCL3) Contract 852 in Singapore (After Cham, 2007)

Volume loss = 0.7% Tunnel ID=5.8m Pile dia. =800mm Pile Length =33.4m


146

Lateral deflection (mm)

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2 days 180 days

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Tunnel

Pile def lection (Test 3)

Free-field lateral soil movement (Test 1, PIV)

Figure 4.19 Tunnelling-induced pile deflection (Test 3) and free-field lateral soil movement at pile location (Test 1)

Axial Force (kN)

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Dep

th (m

)

ST (Test 3) LT (Test 3)


Tunnel

Figure 4.20 Variation of pile axial force with volume loss (Tests 3 and 4)


147

Settlement (mm)

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00 50 100 150 200 250

Dep

th (m

)



Tunnel

Test 3 Pile head settlement

Free-field soil settlement


Figure 4.21 Variation of pile head settlement (Tests 3 and 4) and observed free-field soil movement at pile location (Tests 1 and 2) with volume loss


148


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Dep

th (m

)



Tunnel

Figure 4.22 Variation of pile bending moment with volume loss (Tests 3 and 4)


-30

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00 5 10 15 20 25 30 35 40 45

Dep

th (m

)





Tunnel

Pile deflection

Free-f ield lateral soil movement

Figure 4.23 Variation of pile deflection profiles (Tests 3 and 4) and observed free-field lateral soil movement at pile location (Tests 1 and 2) with volume loss


149

(a) (b)

(c) (d)

Figure 4.24 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with volume loss (Tests 3 and 4)

0

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)

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)

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ile a

xial

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N)


150

0

0.5

1

1.5

2

2.5

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3.5

4

Pile axial force Pile headsettlement

Pile bendingmoment

Pile headdeflection

Test 3 (VL=3%) Test 4 (VL=6.5%)

LT/S

T ra

tio Long-termeffect

Figure 4.25 Long-term to short-term ratio of pile responses for different volume losses (Tests 3 and 4)

Chapter 5 Effects of Tunnelling on Single Piles

151

CHAPTER FIVE

EFFECTS OF TUNNELLING

ON SINGLE PILES

5.1 INTRODUCTION

In this chapter, the results of further centrifuge model tests conducted to investigate the

effects of tunnelling on single piles in clay are presented. The detailed test program

and configurations in prototype scale is given in Table 5.1 and the pile tip positions

investigated in the parametric studies are schematically illustrated in Figure 5.1.

5.2 TEST SERIES 2- EFFECTS OF PILE TIP & HEAD CONDITIONS

5.2.1 Effects of Pile Tip Condition (Tests 3, 9 And 10)

Tests 3, 9 and 10 were performed to simulate three different pile tip conditions, namely

a “floating” pile in Test 3 (presented in Chapter 4), a “socketed” pile in Test 9 and an

“end-bearing” pile in Test 10. For the test on “floating” pile (Test 3), the 22-m long

pile is entirely embedded in the 24-m thick soft clay layer. The soft clay was underlain

by a layer of 3.5-m thick sand layer. For the test on “socketed” pile (Test 9), the soft

clay was underlain by a 8-m thick sand layer and the pile was embedded 3m into the

underlying sand layer. For Test 10, the underlying sand thickness of 3-m is the same as

that for Test 3 but the 27.5 m long pile is resting on the rigid base of the container.


152

Strictly speaking this pile is not a pure “end-bearing” pile, as there will be some load

transfer in the 3 m thick sand layer just above the pile tip. All tests have the same

tunnel volume loss of 3% and the same pile-to-tunnel distance, i.e. 1D or 6 m. The pile

heads are free.

Figures 5.2(a) & (b) show the induced pile axial force profile of Tests 3, 9

and 10 at the end of tunnel excavation (short- term at 2 days) and in the long-term (720

days), respectively. It can be seen that the induced axial force in all tests increases

downwards from the pile head. However, the respective maximum axial force occurs

approximately at the tunnel spring elevation for Test 3 but slightly lower than the

tunnel spring elevation for Tests 9 and 10. This is as expected because the pile in Tests

9 and 10 are socketed into the sand layer and hence the pile settlement is much smaller

than that for a floating pile (Test 3). As such, the larger soil movement relative to pile

settlement induces much larger drag loads on the socketed pile (Test 9). The negative

skin friction is highest for the end-bearing pile (Test 10) as the pile is resting on a rigid

base with negligible measured pile settlement. The elevation of neutral plane in the

socketed pile (Test 9) is lower than that of the floating pile (Test 3). The neutral plane

for end-bearing pile in Test 10 shifts even much lower as compared to Test 9. The

axial force is mostly transferred to the pile socket in Tests 9 and 10, as evidenced from

the steep gradient of positive skin friction shown in Figure 5.2. Intuitively, the neutral

plane for a pure end-bearing pile should be at the pile tip. This is not so for Test 10 as

positive skin friction is mobilised to transfer the load in the sand layer as shown in

Figure 5.2.


153

Owing to lower part of pile socketed in stiff soil, the pile settlement for Test

9 (socketed pile) is only 2 mm in the short- term and 3 mm in the long term. As

expected, the pile settlement for Test 10 (end-bearing pile) is negligible. Owing to the

floating condition of the pile tip, the settlement of the floating pile (Test 3) increases

by 183% from 6 mm in the short- term to 17 mm in the long-term. However, these

magnitudes are still much smaller than the observed ground surface settlement. The

above observations reveal that pile settlement is more critical for a floating pile while

induced negative skin friction is more critical for both socketed and end-bearing piles.

Figures 5.3(a) & (b) show the variation of induced pile bending moment

profiles of Tests 3, 9 and 10 in the ST and LT, respectively. The bending moment

profiles in Tests 3 and 9 share a similar trend as the maximum bending moment occurs

close to the tunnel axis with double curvature. Although the volume loss for both tests

is the same, the socketed pile in Test 9 exhibits a larger maximum bending moment, as

compared to a floating pile (Test 3). This is due to socketing of the pile into sand, thus

sufficiently restricting movement of the lower portion of the pile. The restraint at the

pile tip would restrict pile movement and hence result in slightly larger pile bending

moments. As a result, the induced bending moment on the socketed and end-bearing

pile is more critical than that on the floating pile. On the other hand, the end-bearing

pile (Test 10) exhibits triple curvatures profile with the maximum bending moment

occurring close to the tunnel axis as well. The maximum bending moment is the largest

among the three tests. This is probably because the longer span of pile in Test 10 is

exposed to more soil movements induced by tunnelling and hence the pile length effect

can be one of the factors contributing to the magnitude of induced pile bending

moment. This aspect will be further examined in Section 5.3.


154

Figures 5.4(a) & (b) show the pile deflection profiles for Tests 3, 9 and 10.

The pile head deflection for the socketed pile (Test 9) is only 2.1 mm and 3 mm in the

short- and long-term, respectively. This is much smaller than the corresponding pile

head deflection of the floating pile of 5 mm (ST) and 12.1 mm (LT). From the pile

deflection profile exhibited in the socketed pile, the pile head is noted to bend towards

the tunnel, but there is almost no deflection from the mid-pile shaft to the pile tip. This

is because the lower part of the pile is restrained and hardly moves. In contrast, much

larger deflection is noted for the floating pile. Nevertheless, the pile head deflection

for the end-bearing pile (Test 10) is slightly larger than that of socketed pile, with 2.8

mm and 6 mm in the short- and long-terms, respectively. This is likely due to the

higher elevation of the underlaying sand layer for the socketed pile. As such, the ‘end-

bearing’ pile in Test 10 is exposed to more lateral soil movements and its lower

elevation of pile toe fixity point causes it to deflect more than the socketed pile in Test

9. Moreover, it is noted that the mid-pile shaft in end-bearing pile (Test 10) being

pushed away from the tunnel. This is probably due to the relatively large lateral soil

movement in the immediate shear zone pushing the pile head while the pile tip is being

restrained, and hence causing the mid-pile shaft being bent away from the tunnel. In

short, the induced pile deflection is more critical for a floating pile as compared to

socketed and end-bearing piles.

5.2.2 Effects of Pile Head Condition (Tests 10, 13)

Tests 10 and 13 were performed to study the effects of pile head condition. Test 13 is

modelled such that the pile head is totally fixed in position with no vertical or lateral

movements allowed. This simulates the condition where the pile cap is tied rigidly


155

with the ground beams. In both tests, the 27.5-m long model pile is fully embedded

into the 24-m thick soft clay layer and 3.5 m thick sand layer, and resting on the rigid

base of the model container, simulating an end-bearing pile.

The induced pile axial force profiles of Tests 10 and 13 are shown in Figure 5.2.

It is observed that the axial load transfer profiles are similar for both the free-head pile

and fixed-head pile, with the exception of the development of tensile force in fixed-

head piles. In a completely fixed head condition, the pile is not allowed to settle,

resulting in tensile force induced along the upper portion of pile. The maximum

negative skin friction is observed at an elevation lower than the tunnel spring line or at

approximately 17.5 m (Short-term) and 20 m (Long-term). The general trend of

development of tensile force is observed by Mroueh and Shahrour (2002) and Pang

(2006) as well. Pang (2006) reported that the pile is significantly affected when the pile

cap is restrained. Tensile force is observed along the top 15 m of the pile while the

maximum drag load is reduced compared to the pile without pile cap restraint. For a

fixed-head pile, engineers may need to evaluate the connection between the pile and

the pile cap in resisting the tensile force.

In addition, it is noted that the reduction of maximum drag load in fixed-head

pile is approximately 36% and 33% in the short- and long-term, respectively. However,

the trade-off is that tension force would be induced near to the pile head. It might be

due to the load transfer curve essentially offset towards the tension side due to the total

restraint at pile head (fixed-head condition) applied on the pile. The measured

settlement of end-bearing pile is less than 2mm as the pile tip is rested on a rigid base.


156

Figure 5.3 shows the tunnelling-induced pile bending moment for the free- and

fixed-head end-bearing piles (Tests 10 & 13). The pile bending moment profile is

similar for both cases, where triple curvature is induced with negative bending

moments at the upper and lower portions of the pile body, whilst positive pile bending

moment occurs approximately at the tunnel spring line. As expected, no bending

moment is induced in the free pile head which is allowed to move, while relatively

large negative bending moment is induced at the fixed pile head with zero rotation and

displacement. It is worth noting that the observed negative bending moment at the pile

head is larger than the positive bending moment at the mid-pile shaft. For the fixed-

head pile, the bending moment profile is offset towards the negative side as compared

to the free-head pile. The large magnitude of bending moments at the pile head needs

to be evaluated in practice.

As no pile head deflection is allowed for the fixed-head pile (Test 13), the

measured mid-pile shaft deflection of less than 0.2mm is much smaller than that for the

free-head pile, see Figure 5.4. This might be attributed to the fact that the pile head and

toe are fixed in placed and thus the movement of the pile at mid-pile shaft could be

purely due to the bending of the pile shaft due to the large rigidity of the pile. Since the

pile deflection is negligible, the effects of deflection induced by tunnelling toward the

existing pile in fixed-head pile are not a major concern as compared to the free-head

pile.

Figure 5.5 shows a summary of the variation of short-term and long-term pile

responses with tip and head conditions. As compared to floating pile, socketed and

end-bearing piles experienced smaller induced pile settlement and deflection; but


157

larger induced pile axial force and bending moment. Thus, in conclusion, a floating

pile (Test 3) will be mainly governed by pile settlement when tunnelling is carried out

nearby. For end-bearing piles and piles that are socketed into stiffer material (e.g.

dense sand), the piles will experience significant negative skin friction. In practice, it is

common that a pile is rested or socketed into hard strata and engineers should assess

whether the pile can resist the induced negative skin friction due to tunnelling.

The variation of pile responses for free- and fixed-head pile is now examined.

Since the comparison is made between end-bearing piles, pile movements (settlement

and deflection) are not a major concern, as any fixity in toe would substantially reduce

the pile movement as discussed before. However, the trade-off is the increase in pile

material stress (bending moment and axial force) due to fixity. Particularly, the

restriction of pile head which will cause negative bending moment and tension at the

pile head (Figure 5.5(c)), which may be detrimental to the pile (Figure 5.5(a)).

5.3 TEST SERIES 3 -EFFECTS OF PILE LENGTH (TESTS 3, 7, 8)

Test 7 was conducted using a short pile with embedment length of 11.4 m (ratio of pile

length over tunnel depth, L/H=0.76). The pile tip is located within the immediate shear

zone. In addition, Test 8 was conducted using a pile with embedment length of 15.6 m

(ratio of pile length over tunnel depth, L/H=1.04). The pile tip is located approximately

at the tunnel axis. The results of these 2 tests will be compared with the performance of

the long pile in Test 3, which has an embedment length of 22 m (pile tip below tunnel

invert) to study the effects of pile length. In this section, a short pile is referred to as a

pile with its tip elevation at or above the tunnel axis elevation. As such, Test 3 involves


158

a long pile case and Tests 7 and 8 involve short pile cases. The pile-to-tunnel distance,

tunnel volume loss and other parameters are kept constant in the tests.

Figure 5.6 shows the variation of induced pile axial force with pile length for

Tests 3, 7 and 8. The longest pile (22 m) in Test 3 experiences the largest negative skin

friction as compared with shorter piles in Tests 7 and 8. This is because the pile in test

3 has the greatest pile shaft area, thus allowing the development of a larger negative

skin friction. In addition, the pile tip is located far below the large soil displacement

immediate shear zone. The portion of the pile beneath the shear zone develops positive

skin friction and end bearing to resist the down drag force.

Figure 5.7 shows the variation of induced pile head settlement with pile

length for Tests 3, 7 and 8 and the free-field vertical soil movement at the respective

pile locations. The results demonstrate that the tunnelling-induced settlement on a short

pile is more critical than on a long pile for both short-term and long-term. Moreover,

the incremental short pile settlement in the long-term over short-term is also much

larger than that of the long-pile. The pile settlement in Test 7 is significantly larger

than the piles in Tests 3 and 8 whose tips are beneath the immediate shear zone. These

findings are consistent with the observed large settlement zone for tunnelling in sand

reported by Jacobsz (2002). However, as the trough width in clay is generally larger

than that in sand (Rankin, 1998; Ran, 2004), the large settlement zone in clay is noted

to be wider than that in sand.

Figure 5.8 shows the variation of induced bending moment with pile length.

The pile bending moment profile changes from double to single curvature as the pile

tip moves from L/H=1.5 to L/H=1.04 or at the tunnel axis level. It is observed that the


159

maximum bending moment increases with pile length in both short- and long-terms.

The maximum bending moment of the piles (L/H=1.04 and 0.76) occurs above the

elevation of the tunnel spring line and approximately at the middle of the piles, instead

of at the tunnel spring line in the case of long pile (L/H=1.5). In addition, the induced

pile bending moment is greater for a longer pile (pile base located outside the

immediate shear zone) than that of a shorter piles. This is due to restraint of the pile

beneath the immediate shear zone with relatively small or negligible soil movement.

The lateral soil movement profiles shown in Figure 5.9 provide clear evidence on the

changes of induced pile bending moment. As a result, a longer pile would induce a

larger bending moment as compared with a shorter pile.

Figure 5.9 shows the variations of pile deflection for Tests 3, 7 and 8 and the

corresponding free-field lateral soil movement profile obtained from PIV (Test 1). The

pile generally deflects in a similar fashion as the free-field soil displacement profile but

with much ‘smoother’ and smaller movements. The restraint in movement can be

attributed to the large bending stiffness (EI) of the pile body. In contrast to the induced

pile bending moment, the short pile head deflection is significantly larger than that of

the long pile. In the case of long pile, its deflection is found to be the smallest among

the three cases. When the pile length is reduced, the magnitude of pile deflection also

increases. This is due to the fact that the long pile is partially embedded in the support

zone with smaller induced soil movements while the short pile is embedded entirely in

the immediate shear zone which induces relatively larger soil movements. In

consequence, the long pile ‘bends’ towards the tunnel while for the short pile, it

‘shifts’ toward the tunnel with a maximum pile deflection at the pile head for both

cases.


160

Figure 5.10(a) shows the variation of maximum pile axial force with

normalised pile length over tunnel depth (L/H=0.76, 1.04, 1.5). The results clearly

show that the induced maximum pile axial force increases with pile length over tunnel

depth. It is noted that the maximum axial force increases linearly with normalised pile

length over tunnel depth, from 25 kN (L/H=0.76) to 92 kN (L/H= 1.04) and finally to

198 kN when the pile tip is located below the tunnel invert (L/H=1.5). However, in the

long-term, the linear increment of axial force with normalised pile length over tunnel

depth has a much steeper gradient than when compared with the short-term. The axial

force increases from 45 kN to 92 kN and then 198 kN when the normalised pile length

over tunnel depth (L/H) increases from 0.76, 1.04 to 1.5, respectively. This is because

the piles whose tips are located in the zone of immediate shear zone would be

intensively dragged down by the large soil movements, resulting in large pile

settlement. On the other hand, piles with tips located out of the shear zone would

experience less settlement as the lower part of the pile shift is socketed in stiff soil.

Figure 5.10(b) shows the variation of pile head settlement with normalised

pile length over tunnel depth (L/H=0.76, 1.04, 1.5). The data clearly show that when

the pile tip is located within the zone of large displacements, the pile would settle

excessively with the tip in the short-term with a similar magnitude to the soil

displacement at the location, see Figure 5.7.

Figure 5.10(c) shows the maximum induced pile bending moment with

normalized pile length over tunnel depth in the short and long-terms. It is observed that

the maximum bending moment of the long pile is larger than that of shorter piles due


161

to the restraining effect discussed earlier in this section. Similar findings are also

reported by Chen et al. (1999) and Pang (2006).

Figure 5.10(d) shows the maximum pile head deflection with normalized pile

length over tunnel depth in short and long-terms. The maximum induced pile head

deflection of a short pile also exhibits marked increase compared with a long pile. In

the short- term, the pile head deflection of the long pile (Test 3, L/H=1.5) is 5 mm and

increases to 8 mm and 11 mm when the pile length over tunnel depth reduces to 1.04

and 0.76, respectively. This can be attributed to the lateral soil movement profile (Fig.

5.9) induced by tunnelling, where a long pile is restrained if the pile length extends

beneath the immediate shear zone. Hence, the pile deflection is smaller when the pile

length is longer. Similar trend is observed in the long term as well, in which the pile

head deflection increases from 12.1 mm to 18 mm and 22 mm when the pile length

over tunnel depth reduces from 1.5 to 1.04 and 0.76 respectively.

To further assess the effect of pile length over tunnel depth due to tunnelling,

the key pile responses are summarized in Figure 5.11. In a short pile, especially those

located in the immediate shear zone, the pile structure is less vulnerable as compared

to the long pile. However, there will be excessive pile movements (settlement and

deflection) because of lack of anchorage of pile into the stable support zone. In this

respect, a longer pile with extension of pile length into stabilised support zone tends to

provide more resistance to the pile movements but will experiences larger structural

stress in term of bending moment and axial force.


162

The figures illustrate that a short pile is less vulnerable in terms of tunnelling-

induced pile axial force and bending moment, while significant adverse effect is

observed in terms of pile head settlement and deflection for short pile. This

phenomenon is verified in the soil settlements and lateral soil movements shown in

Figures 5.7 and 5.9. It is demonstrated that relatively large soil movements are induced

within the immediate shear zone while relatively small or negligible soil movement

induced in the support zone, thus when the pile base is extended into the support zone,

smaller pile settlement and deflection are expected due to restraint. Paradoxically,

larger axial force and bending moment are induced for longer pile caused by the same

restraint. This finding provides evidence that tunnelling-induced displacements of

short piles (settlement and deflection) are more critical than that of long piles. In

contrast, the pile axial force and bending moment are crucial in the long pile upon

tunnel excavation. This implies that the effect of pile length over tunnel depth has

pronounced implications in tunnelling-induced pile responses.

5.4 EFFECTS OF DISTANCE OF PILE FROM TUNNEL

5.4.1 Test Series 4 - Free-Head Floating Piles (Tests 3, 5, 6 and 16)

In Test series 4, centrifuge model tests (Tests 3, 5, 6 and 16) were performed to study

the behaviours of long free-head floating piles with various pile-to-tunnels centre

distances. The pile-to-tunnel distance in Tests 3, 5, 16 and 6 is 6 m (or 1D), 9 m (or

1.5D), 10m (or 1.67D) and 12 (or 2D), respectively. Other parameters are kept

constant in all tests. For clarity, the results of Test 16 are not presented in Figures 5.12


163

to 5.15, but the maximum pile responses of this test are presented in Figure 5.21 for

comparison purpose.

Figure 5.12 shows the induced pile axial force due to tunnel excavation in

Tests 3, 5, 6 and 16. The test results reveal that all the induced axial load profiles are

similar with the maximum values taking place slightly higher than the tunnel spring

line in the short-term, with the neutral planes shifting lower over time. This implies

that the axial load transfer patterns of the piles are similar irrespective of the pile-to-

tunnel distance for a long pile. A similar trend of axial load variation with pile-to-

tunnel distance was also reported by Ran (2004) and Mroueh et al. (1999). It is

observed that the induced pile axial forces decrease with an increase in pile-to-tunnel

distance. Similar steady decrease is also observed in the long-term. This may be

attributed to the fact that the total contact area for piles in the immediate shear zone is

reduced when the distance of pile-to-tunnel increases. As such, a shorter portion of the

total pile length experiences negative skin friction as a pile is located further away

from the zone of large displacements. This can be attributed to the reduced shaft

contact area with the soils in the immediate shear zone when the distance of pile-to-

tunnel increases. Figure 5.13 shows the variation of pile head settlement and the

free-field vertical soil movement at the respective pile location. Similar gradual

decreases in vertical soil settlement with depth and pile-to-tunnel distance are observed.

The magnitude of pile head settlement also decreases with increasing pile-to-tunnel

distance. This is consistent with the observed variations of pile axial forces from the

three tests. Besides, the smaller magnitudes of soil settlement is expected to induce less

negative skin friction (Figure 5.12) for the case in which pile tip is below the tunnel

invert. Once again, the pile head settlement exhibits time-dependent behaviour and


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reaches its respective peak value after 720 days. The results suggest that the induced

axial pile responses are insignificant when the pile-to-tunnel distance is larger than 2D

in the present study.

Figure 5.14 shows the variation of induced bending moment with pile-to-tunnel

distance. Generally, the induced pile bending moment profiles are similar in all tests

with the maximum bending moment occurring approximately at the tunnel spring

elevation. This is consistent with the numerical predictions reported by Cheng (2003)

and field measurements reported by Pang (2006). As expected, the maximum induced

bending moment generally decreases with increasing pile-to-tunnel distance for both

short and long-terms. Hence, it would be reasonable to assume that induced bending

moment are generally small beyond a horizontal offset of 2D from the tunnel centre as

magnitudes are less than 50 kNm (long-term) even with a relatively large tunnel

volume loss of 3%. In addition, it is evident that regardless of pile-to-tunnel distance,

the induced maximum bending moment increases for some time after the completion

of tunnel excavation in all tests, exhibiting the time-dependent behaviours as described

earlier. The pile responses peak at 720 days after excavation. Figure 5.15 shows the

variations of pile deflection and free-field lateral soil movement profiles from these

three tests. As expected, the pile deflection decreases with increasing pile-to-tunnel

distance. Moreover, it is observed that the pile deflection drops rapidly from 1D to

1.5D, with a much smaller decrease from 1.5D to 2D. The pile lateral responses are

best explained by the soil deflection profiles obtained from Test 1 as shown in Figure

5.15. It is noted that the lateral soil movements in the three tests increase with time and

decrease with increasing distance of pile location to the tunnel, as the soil movement

decreases when the pile-to-tunnel distance increases. It is also noted that at the distance


165

of one tunnel diameter (1D), the lateral soil displacement is prominent at the tunnel

spring elevation and surface. However, when the distance is large enough, for instant

at a distance of 2D, the lateral soil displacement profile reveals significant soil

deflection at the ground surface while the soil movement at the tunnel spring elevation

becomes negligible. This can be related to the soil movement pattern as observed in the

immediate shear zone and thus results in the observed pile bending moment and

deflection shown in Figures 5.14 and 5.15. This finding demonstrates that when the

pile-to-tunnel distance increases, a shorter portion of the pile length is inside the

immediate shear zone, resulting in a smaller tunnel-pile interaction.

5.4.2 Test Series 5 - Free-Head End Bearing Piles (Tests 10, 11, 12)

In this test series, Tests 10, 11 and 12 were performed to study the effects of pile from

tunnel for free-head end bearing piles. The pile-to-tunnel distance in Tests 10, 11 and

12 is 6 m, 10 m and 14 m, respectively. Other parameters are kept constant in all tests.

Figure 5.16 shows the variation of induced free-head end-bearing pile axial force due

to tunnel excavation with pile-to-tunnel distance for Tests 10, 11 and 12. Generally,

the test results illustrate that all the induced axial load profiles are similar with the

maximum values taking place slightly lower than the tunnel spring line in the short-

term, with the neutral plane becomes deeper over time. This implies that the axial load

transfer patterns of the piles are essentially the same irrespective of pile-to-tunnel

distance for an end-bearing pile.

Similar to test series 4 (effects of distance of pile from tunnel for free-head

floating piles), the induced axial force decreases when the pile-to-tunnel distance


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increases. This may be attributed to the fact that the total contact area for piles in the

immediate shear zone is lessened when the distance of pile-to-tunnel increases, as

noted in Section 5.4.1.

An important feature of an end-bearing pile is that the pile would undergo

minimal settlement as the pile tip is rested on very stiff ground. As such, pile

settlement is not a major concern. Although the pile may undergo elastic shortening,

the magnitude is small and thus negligible.

Figure 5.17 shows the variation of induced free-head end-bearing pile

bending moment with distance of pile from tunnel centre. The induced positive and

negative bending moments generally decrease, as expected, with increase in pile-

tunnel distance. This trend is similar to that observed in test series 4 for free-head

floating piles.

Figure 5.18 illustrates the variation of induced pile deflection profiles for

Tests 10, 11 and 12. The magnitude of pile head deflection decreases when the pile-

tunnel distance increases. Likewise, the mid-pile shaft also deflects in a similar trend.

5.4.3 Test Series 6 - Fixed-Head End Bearing Piles(Tests 13, 14A, 14B)

Tests 13, 14A and 14B aim to investigate the effects of pile location from tunnel for

fixed-head end bearing piles. The pile-to-tunnel distance in Tests 13, 14A and 14B is 6

m, 10 m and 14 m, respectively. Other parameters are kept constant in all tests. It

should be noted that pile axial force is only investigated in Test 13 (See Figure 5.2) but


167

not in Tests 14A and 14B, as the axial force profiles for pile-to-tunnel distance of 10 m

and 14 m are expected to be similar to that of Test 13 but with smaller magnitudes.

Nevertheless, the pile bending moment in Tests 14A and 14B are worthy for further

study as the changes are more significant and the results can also be served as the

bench mark for comparison with the pile group responses presented in Chapter 6.

Figure 5.19 shows the variation of induced fixed-head end-bearing pile bending

moment with pile distance from tunnel centre for Tests 13, 14 and 15. In all cases, the

negative bending moment is larger than the positive bending moment due to the

restraint at pile head. The results indicate that both positive and negative bending

moment decreases when the pile-tunnel distance increases.

Figure 5.20 shows the pile deflection profiles which are derived from the pile

bending moment profiles. It is shown that the pile deflection is generally very small

(less than 0.02 mm). The pile deflection profile reveals that the mid-pile shaft deflects

toward the tunnel, but the magnitude decreases when the pile distance to tunnel

increases.

5.4.4 Comparison of Results from Test Series 4, 5 and 6

Figure 5.21(a) shows the variation of maximum pile axial force with pile-to-tunnel

distance for Test Series 4, 5 and 6. It should be noted that for Test Series 6, there were

no axial piles monitored in Tests 14A and 14B. With the test range of pile location of 6

m to 14 m from the tunnel centre, the induced pile axial forces are observed to

decrease fairly linearly with an increase in pile-to-tunnel distance for Test Series 4 and


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5. Similar steady decrease is also observed in the long-term. This may be attributed to

the fact that the total contact area for piles in the immediate shear zone becomes

smaller when the distance of pile-to-tunnel increases, as illustrated in Figure 5.22. As

such, a shorter portion of the total pile length experiences smaller negative skin friction

as the pile moves away from the zone of large displacements. It is also observed that

the induced axial force in Test Series 5 (end-bearing piles) is much larger than that of

Test Series 4 (floating piles) because the larger soil settlement relative to pile

settlement would induce a much larger axial force on the end-bearing pile. In addition,

the most significant difference in Test Series 5 (free-head) and Test Series 6 (fixed-

head) is that tension force is induced in the fixed-head pile due to the total fixed

condition at the pile head.

Similarly to pile axial force, consistent responses in the corresponding soil

surface settlement (Test 1) and the maximum pile head settlement in Test Series 4

(Tests 3, 5, 6 and 16) are observed from Figure 5.21(b). The soil surface and pile

settlements are observed to decrease with an increase in pile-to-tunnel distance in both

short- and long-term. This is simply because when the pile-to-tunnel distance increases,

the soil settlements within the immediate shear zone decreases, as shown in Figure

5.13. The result seems to suggest that the pile settlement are insignificant for pile-to-

tunnel distance larger than 2D in Test Series 4. It should be noted that the pile

settlements of end-bearing piles in Test Series 5 and 6 are practically negligible as

shown in the figure.

In contrast to the pile vertical responses, the pile lateral responses are different.

Figure 5.21(c) shows the induced maximum pile bending moment with pile-to-tunnel


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distance in the short and long-terms for Test Series 4, 5 and 6. Generally, the

maximum induced bending moments decrease reasonably linearly with increasing pile-

to-tunnel distance when the magnitude is relatively small (see for instant pile bending

moment in the short-term and positive bending moment in series 6 (long-term)).

However, the bending moments decrease exponentially when the magnitude is

relatively large (for example long-term bending moment in series 4, 5 and negative

bending moment in series 6). Generally, the induced bending moments in end-bearing

piles (Test Series 5) are larger than the floating piles (Test Series 4). This is probably

because the restraint at the pile toe would restrict the pile lateral movement and induce

a larger bending moment. On the other hand, the data reveals that negative bending

moments are induced at pile head due to total fixity condition for Test Series 6 (fixed-

head) as compared to the free-head piles in Test Series 5. Nevertheless, since the

bending moments are offset toward the negative bending moment, the trend reveals

that the positive bending moment in a fixed-head pile (Test Series 6) is consistently

smaller than that of the free-head pile (Test Series 5), regardless of pile-tunnel position.

In addition, it would be reasonable and safe to assume that induced bending moments

are generally small beyond a horizontal offset of 2D from the tunnel centre as

magnitudes are less than 50 kNm (long-term) even with a tunnel volume loss of 3%. It

is evident that regardless of pile-to-tunnel distance, both induced maximum bending

moments increase for some time after the completion of tunnel excavation in all the

tests, exhibiting the time-dependent behaviours described earlier. The pile responses

peak at 720 days after excavation. This further illustrates that the induced pile bending

moments are small as the lateral soil movements are not significant when the pile-to-

tunnel distance increases beyond 2D, as illustrated in Figures 5.22 and 5.23. As

discussed in Section 4.3.1, the soil within the ‘Immediate Shear Zone’ is ‘unloaded’


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due to tunnel excavation and gradually deforms by arching, causing the radial stress in

the immediate shear zone to be reduced due to stress relief. This leads to the observed

soil movement pattern which subsequently affects the pile responses as observed in

this chapter. On the other hand, circumferential soil stresses increase within the

‘Support Zone’ to support the arches formed in the immediate shear zone. Thus, Figure

5.22 illustrates that the pile responses for different pile-tunnel distance is greatly

influenced by the changes of the stress in both the Immediate Shear Zone and Support

Zone.

Figure 5.21(d) illustrates the variations of pile head deflection for the floating

piles (Test Series 4) and end-bearing piles (Test Series 5). Generally, it is observed that

the pile deflection reduces rapidly from 1D to 1.5D, with a much smaller decrease

from 1.5D to 2D for both Series 4 and 5. This is probably because the lateral soil

movement decreases with increasing distance of pile from to the tunnel. Nevertheless,

it is expected that the pile head deflection for end-bearing piles (Test Series 5) is

smaller than that for floating piles (Test Series 4), regardless of pile-tunnel distance

because the lower portion of pile is restrained and cannot move. On the other hand, the

measured pile head deflection of Test Series 6 (fixed-head) is negligible due to totally

fixed condition at pile head.

The observed variation of pile bending moment and deflection with pile-to-

tunnel distance can be explained by the soil movemnt profiles obtained from Test 1,

which was analysed by PIV (see Figure 5.23). The results reveal that at locations near

the tunnel, the lateral soil displacement is prominent at the tunnel spring elevation.

However, when the distance is large enough, the lateral soil displacement profile


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reveals significant horizontal soil movement at the ground surface while the soil

movement at the tunnel spring elevation became negligible. This finding is consistent

with the relatively insignificant observed pile lateral responses when the pile-to-tunnel

distance is more than 2D. It thus further illustrates that when the pile-to-tunnel distance

increases, a shorter portion of the pile length is inside the immediate shear zone shown

in Figure 5.22.

5.5 EFFECTS OF TIME ON PILE RESPONSES IN SOFT CLAY

In order to further assess the long-term pile responses, Figure 5.24 shows a summary

of the ratio of long-term to short-term pile responses for all tests presented in this

chapter. The results show that the long-term to short-term pile responses (pile axial

force, pile bending moment, pile head settlement and pile head deflection) ranges from

1.34 to 3.5. It is noted that the soil movement is dominant in the vertical direction in

the LT. As such, the LT over ST ratio is comparatively important for pile settlement

which depends on the magnitude of downward soil movement before full pile slip. On

the other hand, the LT over ST ratio for pile axial force is comparatively small as the

axial forces might have been fully mobilized at an earlier stage. Although the

magnitude of lateral soil movement is much smaller than that of vertical soil

movement, the increase in lateral soil movement is significant and thus in turn, causing

both pile bending moment and pile deflection to increase over time. This study has

provided strong evidence that the long-term pile behaviour needs to be considered if

the volume loss due to tunnelling is large.


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5.6 COMPARISON OF PILE BEHAVIOUR DUE TO INWARD AND OUTWARD TUNNEL DEFORMATIONS

Ran (2004) carried out centrifuge model studies on the effect of tunnelling on single

piles in clay. In his study, outward tunnel deformation was simulated (ovalisation of

tunnel lining, i.e. tunnel springline moves outwards), as shown in Figures 5.25 and

5.26. Although outward tunnel deformation is not as commonly observed as inward

tunnel deformation, there have been some reports of outward tunnel deformation in the

field (George, 1981 and Yann and Alain, 1991). For example, the Handbook of Plastic

Pipe Institute (2003) reported that the deformation patterns of poly-material linings

under service load are of horizontal-oval shape. Furthermore, large deformations

experienced by some tunnel lining rings are encountered in some tunnelling projects.

In this section, the difference between inward and outward tunnel deformation patterns

and its influences to adjacent piles are investigated.

It should be noted that the volume loss for the outward tunnel deformation

simulation was 2%, while it is slightly larger at 3% for the inward tunnel deformation

simulation in the present study. Some similarities and differences can be drawn in the

behaviours of soil and single piles induced by both inward and outward tunnel

deformations in clay.

5.6.1 Tunnel-Soil Interaction

Figure 5.27 shows the development of subsurface soil movements at 2 days and 720

days after tunnel excavation for the case of outward tunnel deformation, which was

conducted by (Ran, 2004). Subsurface soil movement was traced from high resolution


173

photographs of the marker beads and analyzed using Computer Program OPTIMUS

instead of PIV used in this study.

5.6.1.1 Similarities (Tunnel-Soil Interaction)

Figures 5.28 and 5.29 show the variation of surface soil settlement troughs with tunnel

deformation over time. It is interesting to note that the measured short-term surface

settlement trough follows the Gaussian distribution curve fairly well with the inflection

point (i) at approximately 7.5 m for both cases, despite the difference in tunnel

deformation. This finding is consistent with the observation made by Verruijt and

Booker (1996). The void created by ovalisation deformation at the tunnel crown is

similar to the clearance at the upper half of the tunnel for the case of contraction

deformation. Hence the soil above the tunnel crown in both cases settles by a similar

vertical distance. However, Verruijt and Booker (1996), (2000) reported that for cases

with significant tunnel ovalisation, the surface settlement trough would be narrower

than the Gaussian curve.

The Gaussian curve is also found to be inappropriate in depicting the measured

long-term surface settlement troughs for both cases. In addition, it is observed that the

soil continues to settle with time and the settlement rate decreases with time. This

exhibits time-dependent behaviour of clay for both cases. An examination of soil

settlement at the pile location (i.e. 6m or 1D from tunnel centre-line) shown in Figure

5.30 reveals that the soil settlement profile is fairly similar for these 2 situations.


174

5.6.1.2 Differences (Tunnel-Soil Interaction)

The most distinct difference in the soil behaviour between the 2 different tunnel

deformations is that the soil moves intensely away from the tunnel due to spring line

expansion in the case of outward tunnel deformation and a much larger deformation

zone is formed. In the case of inward tunnel deformation, the soil moves towards the

tunnel and the immediate shear zone is defined as discussed in Section 4.3 which is

smaller than that of the deformation zone for the outward tunnel deformation. Thus,

qualitatively, the soil movement above the tunnel crown is prominent in the case of

inward tunnel deformation; whereas the lateral soil movement near tunnel axis is

prominent in the case of outward tunnel deformation. This is illustrated in Figure 5.31

that the soil moves in opposite directions at the pile location for these 2 situations due

to differences in the deformation of the tunnel lining. However, the magnitude is much

larger in the case of outward tunnel deformation, despite the volume loss is smaller at

only 2% as compared to 3% for the inward tunnel deformation case.

5.6.2 Tunnel-Pile Interaction

In both tunnel deformation studies, the pile-to-tunnel distance is similar, i.e. 6 m from

the tunnel centre-line. However, the pile length in the case of outward tunnel

deformation is 23.5 m whereas it is slightly shorter at 22 m in the case of inward tunnel

deformation. The tunnel depth remains at 15 m in both cases. The long-term pile axial

forces were not presented by Ran (2004) as the pile axial forces were measured by

‘quarter bridge’ strain gauge circuits. Strain gauge readings obtained with ‘quarter


175

bridge’ circuits have a tendency to drift with time due to temperature changes, whereas

this does not happen for ‘full bridge’ circuits in the present study. Similar to the

tunnel-soil interaction studies, the volume loss in the outward tunnel deformation and

inward tunnel deformation simulations is 2% and 3%, respectively.

5.6.2.1 Similarities (Tunnel-Pile Interaction)

For both tunnel lining deformations, the induced pile axial forces (Figure 5.32) exhibit

similar profiles, as the piles experience negative skin friction due to settling soil

around the pile shaft. Also, the neutral plane is found to be located approximately at

the tunnel axis for both cases. Despite the slightly lower volume loss in the case of

outward tunnel deformation, the pile axial forces in both cases have similar magnitudes.

This is because the relatively large amount of tunnel expansion at the springline for the

outward tunnel deformation causes large settlement, as clearly shown in Figure 5.30.

In the same way, Figure 5.33 reveals that the pile vertical settlement is also time-

dependent. As the piles are ‘floating’ in the soft clay instead of being socketed into the

hard stratum for both cases, the pile settlement depends very much on the vertical soil

movement along the pile shaft. Hence, the piles continue to settle with the soil until

full pile slip, regardless of tunnel deformation patterns.

5.6.2.2 Differences (Tunnel-Pile Interaction)

It is worth noting that totally different pile bending moment profiles are induced under

the two different tunnel deformation patterns. It is observed that outward tunnel

deformation causes negative maximum induced bending moment (bending away from


176

tunnel) whereas inward tunnel deformation causes positive maximum induced bending

moment which bends toward tunnel (Figures 5.34 and 5.35).This is because in the case

of outward tunnel deformation, the soil moved away from the tunnel at the tunnel axis

due to the protrusion of the tunnel lining at the tunnel spring line, hence pushing the

pile away from the tunnel. On the other hand, in the case of inward tunnel deformation,

the soil moves towards the tunnel axis due to volume loss caused by excavation over-

cut, thus drawing the pile towards the tunnel. The relatively large magnitude of

induced pile bending moment in the outward tunnel deformation simulation is mainly

due to the large expansion of the tunnel lining at the springline. In the case of inward

tunnel deformation, the gap above the tunnel crown is the main cause of soil

movements (Leung, 2006). The induced pile deflection profiles shown in Figure 5.36

demonstrate the pushing of the mid-pile shaft away from the tunnel due to tunnel

protrusion at the tunnel spring elevation. In the case of inward tunnel deformation, the

pile moves towards the tunnel, with the pile head deflecting more than the pile tip.

These findings are consistent with the soil movements observed in Figure 5.31.

Similar to the study on deep excavations in clay at NUS (Leung et al.,

2006), the clay continues to move over time after the completion of tunnel excavation.

Thus, the induced pile bending moments, head settlement and deflection in both tunnel

deformation simulations also change with time, as illustrated by the long-term to short-

term ratio of pile responses in Figure 5.37.

The above comparison of pile behaviours due to inward and outward tunnel

deformations illustrate that the trend of pile axial force and pile settlement behaviour

and profile are essentially similar regardless of tunnel deformation pattern. However,


177

the outward tunnel formation would induce higher pile responses as compared to the

inward tunnel deformation under the same volume loss. On the other hand, the pile

lateral responses (bending moment and deflection) are totally opposite for both inward

and outward tunnel deformations, respectively, in terms of profiles and magnitude.

This is due to the fact that the different tunnel deformation patterns would induce

different soil movement profiles and patterns, which in turn changes the pile behaviour

significantly. Nevertheless, considerable engineering judgement and experiences are

required to first determine the tunnel deformation pattern, in a case by case basis.

Subsequently, the analysis of soil movements is necessary to evaluate the pile

responses due to tunnelling.


A total of thirteen centrifuge model tests (Tests 3 to 16) have been performed to

examine the fundamental mechanisms behind the effects of tunnelling on single piles.

Three main items were investigated: (1) pile vertical and lateral responses in different

tunnel-pile configurations (six test series), (2) tunnelling-induced response of a single

pile in relation to tunnelling-induced response of free-field ground, and (3) comparison

of soil and single pile behaviours due to inward and outward tunnel deformations.

Three different pile tip conditions, namely “floating” pile, “socketed” pile and

“end-bearing” pile were investigated to study the effects of pile tip condition. It is

noted that a floating pile is mainly governed by pile settlement when tunnelling is

carried out adjacent to it. On the contrary, socketed piles are likely governed by the

material stress of the pile. On the other hand, some opposite trends are observed in the

fixed-head when compared to free-head. It is noted that tensile force and relatively


178

large negative bending moments are induced at the pile head due to total fixity.

Nevertheless, these responses have led to the reduction in drag load and positive

bending moment at the mid-pile shaft.

Different lengths of piles are deployed to further assess the effect of pile

length over tunnel depth due to tunnelling. In a short pile, especially those located in

the immediate shear zone, the pile structural responses are less vulnerable as compared

to those of long pile. However, there will be excessive pile movements (settlement and

deflection) because of lack of anchorage of pile into the stable support zone. In this

respect, a longer pile with pile length in the stabilised support zone tends to provide

more resistance to the pile movements but will attract more bending moment and axial

force.

Test Series 4, 5 and 6 examine the effects of pile-to-tunnel distance for

different pile head and tip conditions. Generally, it is observed that the pile responses

decrease with increase in pile-to-tunnel distance. The induced pile axial forces are

observed to decrease fairly linearly with increase in pile-to-tunnel distance for Test

Series 4 (floating piles) and 5 (end-bearing pile) and the axial force in Test Series 5 is

always much larger than that of Test Series 4 for all pile to tunnel distances. In

addition, the most significant difference in Test Series 5 (free-head) and Test Series 6

(fixed-head) is that tension force is induced in the fixed-head pile due to total fixed

condition at the pile head. Similarly to pile axial force, consistent responses are also

observed in the corresponding soil surface settlement (Test 1) and the maximum pile

head settlement in Test Series 4.


179

In contrast to the pile vertical responses, the pile lateral responses are different.

Generally, the maximum induced bending moments decrease fairly linear with

increasing pile-to-tunnel distance when the magnitude is relatively small (for instant

pile bending moment in the short-term and positive bending moment in series 6, but

the bending moments decrease exponentially when the magnitude is relatively large.

Generally, the induced bending moments in end-bearing piles (Test Series 5) are larger

than the floating piles (Test Series 4) due to the restraint at the pile toe would restrict

the pile lateral movement and induce a larger bending moment. On the other hand, the

results reveal that negative bending moments are induced at pile head due to total

fixity condition for Test Series 6 (fixed-head) as compared to the free-head piles in

Test Series 5. As the bending moment profile is offset toward the negative bending

moment for fixed-head pile, the positive bending moment in fixed-head pile (Test

Series 6) are consistently lower than that of free-head pile (Test Series 5). It can be

established from the test results that induced bending moments are generally small

beyond a horizontal offset of 2D from the tunnel centre. Generally, it is observed that

the induced pile deflection reduces rapidly from 1D to 1.5D, with a much smaller

decrease from 1.5D to 2D for both Series 4 and 5. This is because the lateral soil

movements decrease with increasing distance of pile location to the tunnel. The pile

head deflection for end-bearing piles (Test Series 5) is smaller that of floating piles

(Test Series 4), regardless of pile-tunnel distance, as the lower portion of the pile is

restrained and cannot move.

Some similarities and differences were drawn in the comparisons of soil and

single pile behaviour in the cases of both inward (present study) and outward (Ran,

2004) tunnel deformations. It is noted that the measured short-term surface settlement


180

trough follows the Gaussian distribution curve fairly well with the inflection point (i)

at approximately 7.5 m for both tunnel deformation cases. The most distinct difference

in the soil behaviour is that the soil moves significantly away from the tunnel in the

case of outward tunnel deformation, whereas in the case of inward tunnel deformation,

the soil moved towards the tunnel. It is revealed that the pile axial force and pile

settlement behaviour and profile are essentially similar regardless of tunnel

deformation pattern, but the outward tunnel formation would induce larger pile

responses as compared to the inward tunnel deformation under the same volume loss.

On the other hand, the pile lateral responses (bending moment and deflection) are

opposite in direction for both inward and outward tunnel deformations, respectively, in

terms of profiles and magnitude.

Chapter 5 Effects of Tunneling on Single Piles

181

Table 5.1 Test program and prototype parameters in Phase 2 study Phase 2 -Effects of tunnelling on single piles

Test series 2 Effects of pile tip & head conditions Test No. Configuration

Common parameters

Individual parameters

3 Typical

Floating free-head pile.

Thickness of sand, S=3.5m

Pile embedment length, L=22m

9

Socketed free-head pile.

S=8.5m. Pile socketed 3m

in sand L=22m

10

End-bearing

free-head pile. S=3.5m

L=27.5m

13

C = 12 m

D = 6 m

X = 6 m

Volume loss = 3 %

End-bearing fixed-head pile.

S=3.5m L=27.5m

C

L


S

D

3m

D

C

L


S 2m

XD

C

L


XD

C

L

‘Bending’ Pile ‘Axial’ Pile

24m

24m

8.5m

3.5m

19m

24m

Tie Beam

24m

3.5m

3.5m


182

Test series 3 Effects of pile length for free-head piles



3

Typical

L=22m

7

L=11.4m

8

C = 12 m

D = 6 m

X = 6 m

Volume loss = 3 %

L=15.6m

Test series 4 Effects of distance of pile from tunnel (a) free-head floating piles



3 Typical

X = 6 m

5

X = 9 m

16

X = 10m

6

C = 12 m

D = 6 m

L = 22 m

Volume loss = 3 %

X= 12 m

XD

C

L


XD

C

L


24m

3.5m

24m

3.5m


183

Test series 5 Effects of distance of pile from tunnel (b) free-head end bearing piles



10

X = 6 m

11

X = 10 m

12

`

C = 12 m

D = 6 m

L = 27.5 m

Volume loss = 3 %

X= 14 m

Test series 6 Effects of distance of pile from tunnel (c) fixed-head end bearing piles



13

X = 6 m

14A

X = 10 m

14B

C = 12 m

D = 6 m

L = 27.5 m

Volume loss = 3 %

X= 14 m

XD

C

L


XD

C

L

‘Bending’ Pile ‘Axial’ Pile

Tie Beam

24m

3.5m

24m

3.5m


184

Figure 5.1 Pile base position investigated in the parametric studies (not to scale)

Pile-tunnel configuration (Pile base position)

Tunnel D=6m

L/H=0.76

L/H=1.04

L/H=1.5

H=15m

Toyoura Sand

Kaolin Clay X/D=1 X/D=1.5 X/D=2

Notes: Volume loss for all tests is 3%, except Test 4 (Vol. loss=6.5%) L = Pile length H = Tunnel depth X = Distance between tunnel axis and centre of pile D = Tunnel diameter

Test 7

Test 8

Tests 3, 4, 9 Tests 5, 16 Tests 6

Free-head- Test 10 Test 11 Test 12Fixed-head- Test 13 Test 14A Test 14B


185

Axial Force (kN)

-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-200 0 200 400 600 800 1000

Dep

th (m

)

Test 3-f loating, free-head

Test 9-socketed, free-head

Test 10-end-bearing, free-head

Test 13-end-bearing, f ixed-head

Tunnel

(a) Short-term

Axial Force (kN)

-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-200 0 200 400 600 800 1000

Dep

th (m

)





Tunnel

(b) Long-term

Figure 5.2 Variation of pile axial force with tip condition in (a) Short-term (b) Long-

term (Tests 3, 9, 10 and 13)


186


-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-200 -150 -100 -50 0 50 100 150 200 250 300

Dep

th (m

)

Test 3-f loating, free-headTest 9-socketed, free-headTest 10-end-bearing,free-headTest 13-end-bearing,f ixed-head

Tunnel

(a) Short-term


-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-200 -150 -100 -50 0 50 100 150 200 250 300

Dep

th (m

)

Test 3-f loating, free-headTest 9-socketed,free-headTest 10-end-bearing,free-headTest 13-end-bearing,f ixed-head

Tunnel

(b) Long-term

Figure 5.3 Variation of pile bending moment with tip condition (a) Short-term (b)

Long-term (Tests 3, 9, 10 and 13)


187


-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-4 -2 0 2 4 6 8 10 12 14

Dep

th (m

)





Tunnel

(a) Short-term


-27.5

-22.5

-17.5

-12.5

-7.5

-2.5

-4 -2 0 2 4 6 8 10 12 14

Dep

th (m

)





Tunnel

(b) Long-term

Figure 5.4 Variation of pile deflection with tip condition (a) Short-term (b) Long-term (Tests 3, 9, 10 and 13)


188

(a) (b)

(c) (d)

Figure 5.5 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with tip and head conditions (Tests 3, 9, 10

and 13)

0

4

8

12

16

20

24

Test 3-floating, free-

head

Test 9-socketed,free-head

Test 10-end-bearing, free-

head

Test 13-end-bearing, fixed-

head

ST

LT

Pile

hea

d se

ttlem

ent (

mm

)

-200

-100

0

100

200

300

400

500

600

700

800

Test 3-floating,

free-head


Test 10-end-

bearing,free-head

Test 13-end-

bearing,fixed-head

Test 13-end-

bearing,fixed-head

(at pilehead)

ST

LT

Max

imum

pile

axi

al fo

rce

(kN

)

-200

-150

-100

-50

0

50

100

150

200

Test 3-floating,

free-head


Test 10-end-

bearing,free-head

Test 13-end-

bearing,fixed-head

Test 13-end-

bearing,fixed-head

(at pilehead)

ST

LT

Max

imum

pile

ben

ding

mom

ent (

kNm

)

0

4

8

12

16

Test 3-floating, free-

head


Test 10-end-bearing, free-

head

Test 13-end-bearing, fixed-

head

ST

LT

Pile

hea

d de

flect

ion

(mm

)


189

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600 700 800

Dep

th (m

)

ST (Test 3, L=22m)LT (Test 3, L=22m)ST (Test 7, L=11.4m)LT (Test 7, L=11.4m)ST (Test 8, L=15.6m)LT (Test 8, L=15.6m)

Tunnel

Figure 5.6 Variation of pile axial force with pile length (Tests 3, 7 and 8)

Settlement (mm)

-30

-25

-20

-15

-10

-5

00 20 40 60 80 100 120

Dep

th (m

)

ST (Test 1, Soil settlement)LT (Test 1, Soil settlement)ST (Test 3, L=22m)LT (Test 3, L=22m)ST (Test 7, L=11.4m)LT (Test 7, L=11.4m)ST (Test 8, L=15.6m)LT (Test 8, L=15.6m)

Tunnel

Figure 5.7 Variation of pile head settlement and soil settlement profile (Test 1) with pile length (Tests 3, 7 and 8)


190


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250 300

Dep

th (m

)

ST (Test 3, L=22m)LT (Test 3, L=22m)ST (Test 7, L=11.4m)LT (Test7, L=11.4m)ST (Test 8, L=15.6m)LT (Test 8, L=15.6m)

Tunnel

Figure 5.8 Variation of pile bending moment with pile length (Tests 3, 7 and 8) Lateral deflection (mm)

-30

-25

-20

-15

-10

-5

00 10 20 30 40

Dep

th (m

)

ST (Test 1, Soil lateral deflection)LT (Test 1, Soil lateral deflection)ST (Test 3, L=22m)LT (Test 3, L=22m)ST (Test 7, L=11.4m)LT (Test 7, L=11.4m)ST (Test 8, L=15.6m)LT (Test 8, L=15.6m)

Tunnel

Figure 5.9 Variation of pile head deflection and free- field lateral soil displacement (Test 1) with pile length (Tests 3, 7 and 8)


191

(a) (b)

(c) (d)

Figure 5.10 Variation of (a) maximum pile axial force (b) pile head settlement (c) pile bending moment (d) pile head deflection with normalized pile length over tunnel depth

(Tests 3, 7 and 8)

0

50

100

150

200

0 0.5 1 1.5 2Normalised pile length over tunnel depth, L/H

ST

LT

Max

imum

pile

ben

ding

mom

ent (

kNm

)

0

5

10

15

20

25

30


ST

LT

Pile

hea

d de

flect

ion

(mm

)

0

10

20

30

40

50

60

70

80


ST

LT

Pile

hea

d se

ttlem

ent (

mm

)

0

100

200

300

400

500


ST

LT

Max

imum

pile

axi

al fo

rce

(kN

)


192

0

0.5

1

1.5

2

2.5

3

3.5

4

4.5

5

Pile axial force Pile bendingmoment

Pile headsettlement

Pile headdeflection

ST-Test 7 (L/H=0.76) LT-Test 7 (L/H=0.76)ST-Test 8 (L/H=1.04) LT-Test 8 (L/H=1.04)

Shor

t pile

/Lon

g pi

le (L

=22m

, Tes

t 3) r

atio

Negativeeffect forshorter pile

Positiveeffect forshorter pile

Figure 5.11 Short pile to long pile ratio of pile responses for different pile length over tunnel depth (Tests 3, 7 and 8)


193

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600

Dep

th (m

)

ST (Test 3, X= 6m) LT (Test 3, X= 6m)


ST (Test 6, X=12m) LT (Test 6, X=12m)

Tunnel

Figure 5.12 Variation of pile axial force with pile-to-tunnel distance for free-head

floating piles (Tests 3, 5 and 6)

Settlement (mm)

-30

-25

-20

-15

-10

-5

00 20 40 60 80 100 120

Dep

th (m

)

ST (Test 1, X= 6m) LT (Test 1, X= 6m)ST (Test 1, X= 9m) LT (Test 1, X= 9m)ST (Test 1, X=12m) LT (Test 1, X=12m)

Tunnel





Figure 5.13 Variation of pile head settlement for free-head floating piles (Tests 3, 5 and 6) and free-field soil settlement (Test 1) with pile-to-tunnel distance


194


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160 200 240

Dep

th (m

)

ST (Test 3, X= 6m)

LT (Test 3, X= 6m)

ST (Test 5, X= 9m)

LT (Test 5, X= 9m)

ST (Test 6, X=12m)

LT (Test 6, X=12m)

Tunnel

Figure 5.14 Variation of pile bending moment for free-head floating piles with pile-to-

tunnel distance (Tests 3, 5 and 6)


-30

-25

-20

-15

-10

-5

00 5 10 15 20 25 30 35

Dep

th (m

)

ST (Test 1, X= 6m) LT (Test 1 X= 6m)


ST (Test 1, X=12m) LT (Test 1, X=12m)



ST (Test 6, X= 12m) LT (Test 6, X=12m)

Tunnel

Free-field lateral soil deflection

Pile head deflection

Figure 5.15 Variation of pile deflection for free-head floating piles (Tests 3, 5 and 6) and free-field lateral soil displacement profile (Test 1) with pile-to-tunnel distance


195

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 200 400 600 800 1000

Test 10, X= 6m (ST) Test 10, X= 6m (LT)

Test 11, X=10m (ST) Test 11, X=10m (LT)

Test 12, X=14m (ST) Test 12, X=14m (LT)

Tunnel

Dep

th (m

)

Figure 5.16 Variation of pile axial force for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)


196


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250

Test 10, X= 6m (ST)Test 10, X= 6m (LT)Test 11, X=10m (ST)Test 11, X=10m (LT)Test 12, X=14m (ST)Test 12, X=14m (LT)

Dep

th (m

)Tunnel

Dep

th (m

)Tunnel

Figure 5.17 Variation of pile bending moment for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)


-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4 5 6 7 8

Dep

th (m

)

ST (Test 10, X= 6m)LT (Test 10, X= 6m)ST (Test 11, X=10m)LT (Test 11, X=10m)ST (Test 12, X=14m)LT (Test 12, X=14m)

Tunnel`

Figure 5.18 Variation of pile deflection for free-head end bearing piles with pile-to-tunnel distance (Tests 10, 11 and 12)


197


-30

-25

-20

-15

-10

-5

0-200 -150 -100 -50 0 50 100 150 200

Test 13, X= 6m (ST) Test 13, X= 6m (LT)

Test 14A, X=10m (ST) Test 14A, X=10m (LT)

Test 14B, X=14m (ST) Test 14B, X=14m (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel

Figure 5.19 Variation of pile bending moment for fixed-head end bearing piles with

pile-to-tunnel distance (Tests 13, 14A and 14B) Lateral deflection (mm)

-30

-25

-20

-15

-10

-5

0-0.01 0 0.01 0.02 0.03

Dep

th (m

)

Test 13, X= 6m (ST)Test 13, X= 6m (LT)Test 14A, X=10m (ST)Test 14A, X=10m (LT)Test 14B, X=14m (ST)Test 14B, X=14m (LT)

Tunnel`

Figure 5.20 Variation of pile deflection for fixed-head end bearing piles with pile-to-tunnel distance (Tests 13, 14A and 14B)


198

(a) (b)

(c) (d)

Note:- Test Series 4 – Free-head floating piles Test Series 5 – Free-head end-bearing piles Test Series 6 – Fixed-head end-bearing piles

Figure 5.21 Variation of (a) maximum pile axial force (b) maximum pile head

settlement and soil surface settlement (Test 1) (c) pile bending moment (d) maximum pile head deflection for Test Series 4, 5 and 6 with pile-to-tunnel distance

-160

-120

-80

-40

0

40

80

120

160

0 2 4 6 8 10 12 14

Distance between tunnel centre and pile (m)

Series 4 (ST)

Series 4 (LT)

Series 5 (ST)

Series 5 (LT)

Series 6 (ST)

Series 6 (LT)

Series 6, head BM(ST)

Series 6, head BM(LT)

Max

imum

pile

ben

ding

mom

ent (

kNm

)

0

2

4

6

8

10

12

14

16

0 2 4 6 8 10 12 14Distance between tunnel centre and pile (m)

Series 4 (ST)

Series 4 (LT)

Series 5 (ST)

Series 5 (LT)

Series 6 (ST)

Series 6 (LT)

Def

lect

ion

(mm

)

-200

-100

0

100

200

300

400

500

600

700

800

0 2 4 6 8 10 12 14

Distance between tunnel centre and pile (m)

Series 4 (ST)

Series 4 (LT)

Series 5 (ST)

Series 5 (LT)

Series 6 (ST)

Series 6 (LT)

Series 6,tension (ST)Series 6,tension (LT)

Max

imum

pile

axi

al fo

rce

(kN

)

0

10

20

30

40

50

60

70

80

90

100

0 2 4 6 8 10 12 14Distance between tunnel centre and pile (m)

Series 4 (ST)

Series 4 (LT)

Series 5 (ST)

Series 5 (LT)

Series 6 (ST)

Series 6 (LT)

Soil surface settlement (LT), Test 1

Soil surface settlement (LT), Test 1

Settl

emen

t (m

m)


199

Figure 5.22 Assessment of pile responses for different pile-to-tunnel distance (Tests 3, 5 and 6)

Figure 5.23 Lateral soil displacement profiles at different pile-to-tunnel distance (Tests 3, 5 and 6)

‘Support Zone’

Tunnel

‘Immediate Shear Zone’

Toyoura Sand

Kaolin Clay

4m from tunnel

-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0

2 days


720 days

Soil movement (mm)

6m from tunnel

-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0

Soil movement (mm)

Dep

th b

elow

GL

(m)

9m from tunnel

-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0

Soil movement (mm)

Dep

th b

elow

GL

(m)

12m from tunnel

-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0

Soil movement (mm)

Dep

th b

elow

GL

(m)

15m from tunnel

-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0

Soil movement (mm)

Dep

th b

elow

GL

(m)

Tunnel


200

0

0.5

1

1.5

2

2.5

3

3.5

4

Pile axial force Pile head settlement Pile bending moment Pile head deflectionTest 3 (Typical) Test 4 (VL=6.5%) Test 5 (X=9m)Test 6 (X=12m) Test 7 (L/H=0.76) Test 8 (L/H=1.04)Test 9 (Socketed) Test 10 (free-head, X=6m) Test 11 (free-head, X=10m)Test 12 (free-head, X=14m) Test 13 (fixed-head, X=6m) Test 14A (fixed-head, X=10m)Test 14B (fixed-head, X=14m) Test 16 (X=10m)

Long

-term

/ Sho

rt-te

rm ra

tio

Long-termeffect

Figure 5.24 Long-term to short-term ratio of pile responses for all tests (Tests 3, 4, 5, 6, 7, 8, 9, 10, 11, 12, 13, 14A, 14B and 16)


201

Figure 5.25 Comparison of (a) ovalisation of tunnel lining by Ran (2004); and (b) over-cut of tunnel in the present study

Figure 5.26 Simplified tunnel lining ovalisation with time (not to scale) (Test 1) (after Ran, 2004)


202

-20 -18 -16 -14 -12 -10 -8 -6 -4 -2 0

-25

-20

-15

-10

-5

0

10

30

50

70

90

110

130

Distance from tunnel central li ne (m)

Dep

th b

elow

grou

nd le

vel (

m)

(mm)

Ground surface

-20 -18 -16 -14 -12 -10 -8 -6 -4 -2 0

Distance from tunnel central line (m)

-25

-20

-15

-10

-5

0

Dep

th b

elow

gro

und

leve

l (m

)

10

20

30

40

50

60

Ground surface

Tunnel depth 15m

(mm)

(a)

(b)

Figure 5.27 Development of subsurface soil movements at (a) 2 days and (b) 720 days after tunnel excavation (after Ran, 2004)


203

-1

-0.9

-0.8

-0.7

-0.6

-0.5

-0.4

-0.3

-0.2

-0.1

00 2 4 6 8 10 12 14 16 18 20 22 24 26


Test 3, 2 days (VL=3%)Test 3, 2 days, Gaussian curve (VL=3%)Test 3, 720 days(VL=3%)Ran (2004), 2 days (VL=2%)Ran (2004), 720 days (VL=2%)Ran (2004), 2 days, Gaussion Curve (VL=2%)


Sv/S

max

Figure 5.28 Variation of surface soil settlement troughs with tunnel deformation

-1

-0.9

-0.8

-0.7

-0.6

-0.5

-0.4

-0.3

-0.2

-0.1

00 180 360 540 720 900 1080 1260 1440

Time (days)

Sv/ S

max

Inward tunnel deformatio (Test 3, VL=3%)

Outward tunnel deformation (Ran, 2004, VL=2%)

Figure 5.29 Variation of maximum surface soil settlement at tunnel central line with tunnel deformation


204

Settlement (mm)

-30

-25

-20

-15

-10

-5

00 20 40 60 80 100 120 140 160 180 200

Dep

th (m

)

Inward tunnel def ormation, Test 3, 2 day s

Inward tunnel def ormation, test 3, 720 day s

Outward tunnel def ormation, Ran (2004), 2 day s

Tunnel


Figure 5.30 Variation of vertical soil settlement at pile location with tunnel deformation


-30

-25

-20

-15

-10

-5

0-40 -30 -20 -10 0 10 20 30 40 50

Dep

th (m

)

Inward tunnel deformation, Test 3, 2 days (VL=3%)

Inward tunnel deformation, Test 3, 720 days (VL=3%)

Outward tunnel deformation, Ran (2004), 2 days (VL=2%)

Outward tunnel deformation, Ran (2004), 720 days (VL=2%)

Tunnel

Figure 5.31 Variation of soil deflection at pile location with tunnel deformation


205

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600 700 800

Dep

th (m

)

Inw ard tunnel deformation, Test 3, 2 days

Inw ard tunnel deformation, Test 3, 720 days

Outw ard tunnel deformation, Ran (2004), 2 days

Tunnel


Figure 5.32 Variation of pile axial force with tunnel deformation

0

5

10

15

20

25

30

Inw ard Tunnel Deformation (Test 3)

Outw ard Tunnel Deformation(Ran,2004)

ST LT

Pile

hea

d se

ttlem

ent (

mm

)

Figure 5.33 Variation of pile head settlement with tunnel deformation


206


-30

-25

-20

-15

-10

-5

0-350 -250 -150 -50 50 150 250 350

Dep

th (m

)

Inw ard tunnel deformation, Test 3, 2 daysInw ard tunnel deformation, Test 3, 720 daysOutw ard tunnel deformation, Ran (2004), 2 days Outw ard tunnel deformation, Ran (2004), 720 days

Tunnel

Figure 5.34 Variation of pile bending moment with tunnel deformation

-300

-250

-200

-150

-100

-50

0

50

100

150

200

250

300

0 180 360 540 720 900 1080 1260 1440

Time (days)

Inward tunnel deformation, Test 3

Outward tunnel deformation, Ran (2004)

Max

imum

pile

ben

ding

mom

ent (

kN)

Figure 5.35 Variation of tunnelling-induced maximum pile bending moment

over time for different tunnel deformation


207


-30

-25

-20

-15

-10

-5

0-5 0 5 10 15 20

Dep

th (m

)

Pile head deflection, 2 days

Pile head deflection, 720 days

Ran (2004), Pile head deflection, 2 days

Ran (2004), Pile head deflection, 720 days

Tunnel

Figure 5.36 Variation of pile deflection with tunnel deformation

0

0.5

1

1.5

2

2.5

3

3.5

4

Pile bending moment Pile head settlement Pile head deflection

Inward tunnel deformation, Test 3 Outward tunnel deformation, Ran (2004)

LT/S

T ra

tio Long-termeffect

Figure 5.37 Long-term to short-term ratio of pile responses over time for different tunnel deformation

Chapter 6 Effects of Tunnelling on Pile Groups

208

CHAPTER SIX

EFFECTS OF TUNNELLING

ON PILE GROUPS

6.1 INTRODUCTION

The results on the effects of tunnelling on single piles presented in Chapters 4 and 5

provide valuable insights. As piles are commonly installed in groups in practice, the

centrifuge model study on single piles is extended to pile groups in the same soil

conditions. The study aims to address the following issues:

1. Effects of tunnelling on a floating pile group as compared to single floating

pile.

2. Effects of tunnelling on an end-bearing pile group as compared to single

end-bearing pile with capped-head and fixed-head conditions.

3. Effects of size of pile group due to tunnelling.

4. Effects of tunnelling on pile groups with capped-head and fixed-head

conditions.

5. Effects of time on pile group responses in soft clay.

The test procedure for pile group is similar to that for single piles and the

schematic plan and elevation views for the five pile group tests (Tests PG1 to 5) are

shown in Table 6.1. In conjunction with the results of Tests 3 and 16 reported in the

previous chapter, the responses of a floating capped-head 2-pile group obtained from


209

Test PG1 are compared with a single floating free-head pile. Test PG2 investigates the

responses of an end-bearing capped-head 2-pile group and the results are compared

with those of a single end-bearing free-head pile obtained from Tests 10 and 11. Test

PG 3 evaluates the responses of an end-bearing fixed-head 2-pile group and the results

are compared with those of a single end-bearing fixed-head pile obtained from Tests

13 and 14.

In order to study the effects of size of pile group due to tunnelling, the

behaviours of 2-pile group and 6-pile group are compared (for capped-head - Tests

PG2 and 4; for fixed-head - Tests PG3 and 5). Finally, the effects of tunnelling on pile

groups with capped-head and fixed-head conditions are investigated (for 2-pile group -

Tests PG2 and 3; for 6-pile group - Tests PG4 and 5).

6.2 FLOATING PILE GROUP

Test PG1 was carried out on capped-head floating 2-pile group with the front pile at 6

m and the rear pile at 10 m from the tunnel centreline. The centre-to-centre pile

spacing is hence approximately three times pile diameter, similar to that recommended

by BS8004 (BSI, 1986). The results of Test PG1 are compared with those of single

free-head floating piles Tests 3 and 16 (presented in Chapters 4 and 5). Unfortunately

it was not possible to conduct single capped-head floating pile in the present centrifuge

model setup. In the capped-head pile groups, the cap is connected to the individual pile

heads at about 200 mm above the ground level to avoid interaction between the pile

cap and the soil following the approach adopted by Bransby and Springman (1997).


210

Leung et al. (2003) established that the individual pile responses are similar if the 2

piles are aligned parallel to the induced soil movement direction. As such, the 2 piles

are aligned perpendicular instead of parallel to the tunnel for Test PG1.


Figure 6.1 shows the tunnelling-induced axial force for the front and rear piles in a

capped-head 2-pile group. It is observed that the front pile in the pile group

experiences a larger pile axial force than that of the rear pile which is further away

from the tunnel. The results reveal that the trend of axial load transfer along the front

and rear piles are similar. Figure 6.2 compares the short-term (ST, 2 days after tunnel

of excavation) and long-term (LT, 720 days) axial load transfer of the front pile of the

2-pile group located at 6 m from tunnel centre and that of a single free-head pile

located at the same distance. It is evident that the magnitude of the induced axial forces

for capped-head pile is significantly reduced. Similar observation is noted when

comparing the axial pile load profiles of the rear pile of the 2-pile group located at 10

m from tunnel centre and that of a single pile at the same location, see Figure 6.3. As

the front and rear piles are connected by a rigid pile cap, the 2 piles are forced to act in

unison when subjected to tunnelling induced soil movement. The induced axial force

on the front pile, which experiences larger soil movement as presented in chapter 4,

would be moderated by the rear pile via the rigid pile cap. The reduction of negative

skin friction for the front pile is about 33 % (ST) and 28 % (LT) of that of single pile at

the same location. On the other hand, the corresponding reduction in negative skin

friction for the rear pile is 33 % (ST) and 27 % (LT). Such positive pile group effect is


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consistent with the findings by Kuwabara and Poulos (1989) and Loganathan et al.

(2001).

Figure 6.4 compares the tunnelling-induced pile head settlement for the front

and rear piles of the capped-head 2-pile group and that of corresponding single piles.

The front pile settlement is 53% and 56% smaller than that of corresponding single

piles at 6 m from tunnel centre, in the short-and long-terms, respectively. However, it

is worth noting that the rear pile settlement is slightly larger than that of the

corresponding single pile. The rear pile settlements are 2.2 mm (short-term) and 7.1

mm (long-term), as compared to the 2 mm (short-term) and 6.9 mm (long-term) for the

corresponding single pile. This can be explained by the interaction between the pile

and pile cap, in which the front pile in the rigidly connected pile group is moderated by

the rear pile and hence the pile settlement is smaller than that of the corresponding

single pile. However, as the rear pile is being dragged by the front pile via the rigid

pile cap, the rear pile settlement becomes slightly higher than that of a single pile at the

same location. It is noted that since the front pile settlement is larger than that of rear

pile, the pile cap has tilted slightly. This is evident by the measured pile cap deflection

which will be further discussed later.

6.2.2 Induced bending moment and deflection

Figure 6.5 shows the tunnelling-induced pile bending moment for the front and rear

piles in a capped-head 2-pile group. It was postulated by Ong (2005) that a relaxation

in the fixity of the pile cap may have occurred with the fabricated rigid pile cap used in

his study of excavation-induced soil movement on piles. Hence, for this study, a new


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and improved rigid pile cap has been fabricated to include double layers of bolts

instead of one as used by Ong (2005). With this new rigid pile cap, an almost perfect

fixity can be provided between the pile and the pile cap.

The results shown in Figure 6.5 demonstrate that positive and negative bending

moments were induced for both the front and rear piles. The maximum pile bending

moment occurs approximately at the tunnel spring line for the front pile while the

magnitude of maximum positive or negative bending moments for the rear pile is

similar. The rear pile generates a larger negative bending moment at the pile cap level

as compared to the front pile. As before, since the pile cap is tilted, the pile-cap-pile

interaction causes the front pile responses to be moderated by the rear pile via the rigid

connecting pile cap. Thus, the upper bending moment profiles are different between

the front and rear pile because backward dragging force is exerted on the front pile by

the rear pile through the tie beam. This finding is consistent with the observation of

pile group due to excavation-induced soil movements as reported by Leung et al. (2003)

and Ong et al. (2009).

Figures 6.6 and 6.7 compare the front and rear pile responses with the

corresponding free-head single piles at the same location. Although the trend of the

pile bending moment profiles of 2-piles group and the corresponding free-head single

piles are similar, both front and rear piles of the capped-head pile group demonstrate

negative bending moment due to the presence of pile cap while zero bending moment

is recorded for single free-head pile whose head can move freely. The results further

reveal the shadowing effects of the front pile over the rear pile from the soil movement,

resulting in a smaller measured positive bending moment along the rear pile. On the


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contrary, Loganathan et al. (2001) observed that the bending moment profiles for piles

in a group and single free-head pile are almost the same, except for a small difference

at the pile cap location due to fixity condition.

Since the pile cap is tilted due to differential pile settlement, the pile deflection

profiles for the front and rear piles in a capped-head 2-pile group (Test PG1) is

different, as shown in Figure 6.8. The deflection of each individual pile head is

identical as the pile groups are capped. The pile cap deflection is 1.9 mm in the short-

term and increases continuously with time until the end of the tests with a final

deflection of 3.5 mm. In addition, the pile deflection profiles are slightly different for

the front and rear piles. The front pile, which is subjected to a larger soil movement, is

dragged back towards the rear pile due to the connecting pile cap. For the rear pile, the

lateral soil movement on the pile is smaller but the rigid pile cap drags the rear pile

deflecting towards the tunnel resulting in a different pile deflection profile. The

observed lateral pile deflection profiles are similar to those reported by Leung et al.

(2003) on pile groups subject to excavation-induced soil movement. This demonstrates

that the interaction of pile-cap-pile has a significant effect on pile group for excavation

and tunnelling works.

Figures 6.9 and 6.10 compare the pile deflection profile of the front and rear

pile, respectively, with their corresponding single piles. Since the two piles at various

distances are being capped by a rigid pile cap, considerable interaction between two

piles is expected through the rigid pile cap. The results illustrate that the magnitudes of

the front pile deflection of 3.2 mm (short-term) and 6.1 mm (long-term) for the 2-pile

group (Test PG1) are smaller than that of a corresponding single pile at 6 m (Test 3)


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from tunnel centre with pile deflection of 5 mm (short-term) and 12.1 mm (long-term).

However, they are larger that that of a corresponding single pile at 10 m (Test 16) from

tunnel centre with pile deflection of 2.8 mm (short-term) and 5.4 mm (long-term).

Nonetheless, the magnitude of the deflection of the capped-head pile group is smaller

than the average of the deflections of the two single piles at 6 m and 10 m from the

tunnel centre. This data further suggests that pile cap plays a vital role in the pile group

in resisting the lateral deflection induced by soil movement. In addition, the results

indicate that the pile deflection profile in the capped pile group is different from that of

free-head single pile, especially at the pile head. The front pile in the 2-pile group,

which is subjected to larger soil movements, is restrained at the pile head via the rigid

pile cap and hence the upper portion of the front pile is being dragged back as oppose

to the rigid body translation of pile deflection in the corresponding single pile due to

free head condition. On the other hand, the deflection profile of the rear pile in the 2-

pile group is similar to the corresponding single pile, which the pile tends to bend

towards the tunnel due to the unloading process of tunnel excavation.

Figures 6.12 and 6.13 show the ratio of pile responses for a single pile over pile

group ratio for the front and rear piles in capped-head pile and the corresponding

single pile. Generally, the group effect is beneficial to the front pile in all aspects

except for the negative bending moments. Figure 6.12 provides evidence that the pile

group effect is significant for the front pile, which is subjected to larger soil movement,

particularly bending moment, settlement and deflection. The induced substantial

bending moment at the pile head in the 2-pile group is mainly due to the restraint

provided by the pile cap. However, for the rear pile, the group effect is only beneficial

in axial force and positive bending moment. Adverse effects are observed in the


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negative bending moment, pile head settlement and pile head deflection. The pile

group effect can be attributed to the significant pile-cap-pile interaction as the

individual piles in a group are “forced” to act in unison when subject to different

magnitudes of soil movement (Ong, 2005). As such, the front pile would be moderated

by the rear pile via the rigid pile cap. In addition, the shadowing effect of the front pile

on the rear pile reduces the detrimental effects experienced by the rear pile, thus

resulting in an overall positive effect for the pile group.

6.3 END-BEARING PILE GROUP

6.3.1 Capped-Head

The configuration of Test PG2 is similar to that of Test PG1 except the pile length is

increased from 22 m to 27.5 m and the piles are rested on the base of the strong box to

simulate end-bearing piles (see Table 6.1).

Figure 6.14 shows the tunnelling-induced axial force for the front piles in a

capped-head end-bearing 2-pile group. The results show that the front pile experiences

larger pile axial force than that of the rear pile, similar to that observed for the capped-

head floating 2-pile group. The results of the respective free-head single piles at the

same location from Tests 10 and 11 are compared with the 2 piles from Test PG2 in

Figures 6.15 and 6.16. Although the induced axial pile profiles are similar for single

piles and the 2-pile group, the magnitude of the induced axial forces for the capped-

head piles is significantly reduced in both short-and long-terms, respectively. This

reduction is caused by the presence of pile cap, as explained in Section 6.2.1. Unlike

the floating pile group (PG1) which experiences long-term pile settlement of 7.5 mm


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and 4.8 mm for the front and rear pile, respectively, the measured pile settlement of

end-bearing pile (PG2) is negligible as the pile tip is rested on very stiff ground.

Figure 6.17 shows the tunnelling-induced pile bending moment for both front

and rear piles in a capped-head end-bearing 2-pile group. The results demonstrate

triple curvature in the induced bending moment, whereby negative bending moments

are induced at the pile upper and lower portions of the pile shaft, whilst positive pile

bending moment occurs approximately at the tunnel spring line. Three major trends

have been observed in this study. Firstly, the bending moment profile for the front and

rear pile are different, especially at the upper part of the pile. Secondly, the induced

bending moment increases over time for both front and rear piles. Finally, the induced

maximum positive bending moments are always larger than the maximum negative

bending moments.

The bending moment profiles of end-bearing single piles in Tests 10 & 11.

(presented in Chapter 5) and the pile group test (Test PG2) are plotted in Figures 6.18

and 6.19. The trend of pile bending moment profiles of Test PG2 and those of the

corresponding free-head single piles are similar. However, positive bending moment is

induced near the tunnel axis and negative bending moment induced at the upper and

lower parts of the pile for the pile group. Owing to fixity for capped pile in Test PG2,

there is a significant difference between the capped-head pile group and free-head

single pile at the pile top. For the capped-head pile group, both the front and rear piles

experience negative bending moment due to presence of the pile cap, while zero

bending moment is recorded for the single free-head pile as the pile head can move

freely without any restraint. This is consistent with the capped-head floating pile


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responses in Test PG1. When compared to the corresponding single pile, the

magnitude of maximum induced positive bending moment for capped-head piles is

reduced by about 23% in the short-term and approximately 38 to 46% in the long-term,

demonstrating the positive pile group effect.

Figure 6.20 shows the tunnelling-induced pile deflection profiles for the front

and rear piles in a capped-head end-bearing 2-pile group (Test PG2). The deflection of

each individual pile head is essentially identical as the pile groups are capped. The

induced pile deflection is 1.9 mm in the short-term and increases to 3.5 mm in the

long-term. The pile deflection profiles along the upper portion is similar to the capped-

head floating 2-pile group (Test PG1) due to the capping and dragging effect as

explained in Section 6.2.2. As expected, the pile deflection profile at the mid-pile shaft

for end-bearing pile group is different from the floating pile group. The induced pile

deflection profiles shown in Figure 6.20 demonstrate the pushing of the mid-pile shaft

away from the tunnel due to underlying sand layer that restrains the pile toe movement

and lateral soil movement at pile head, as explained in Section 5.2.1.

Figures 6.21 and 6.22 compare the pile deflection of the front and rear pile with

their corresponding single piles. Two main findings are observed. Firstly, the pile

deflection profile along the upper portion for front pile is very different from the single

free-head piles due to presence of the pile cap. The front pile is dragged back by rear

pile via rigidly connected pile cap. For the rear pile, the profile is similar to the single

pile, which moves in rigid body translation mode. Secondly, the magnitude of the front

pile head deflection for the 2-pile group (Test TG2) of 1.9 mm (short-term) and 3.5

mm (long-term) is smaller than that of a corresponding single pile (Test 10, X = 6 m)


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of 2.8 mm (short-term) and 6 mm (long-term). The rear pile deflection is bigger when

compared to the corresponding single pile (Test 11, X = 10 m) of 1.5 mm (short-term)

and 3 mm (long-term) due to pile group effect (See Figure 6.23). Nevertheless, the

deflection of the pile group (Test PG2) again lies in between and smaller than the

average of deflections of two single piles at the same location, similar to the findings

for the floating 2-pile group (Test TG1).

Figures 6.24 and 6.25 show the single pile over pile group ratio for the front

and rear piles in capped-head condition with the corresponding single pile. The results

evidently reveal that the group effect is beneficial to the front in all aspects, except

negative bending moment which is induced due to pile cap condition. On the other

hand, positive group effects are only observed in the axial force and bending moment

for the rear pile. The pile negative bending moment and pile head deflection are found

to increase when the piles are capped in a group. This is consistent with the findings

for the floating capped-head 2-pile group condition.

6.3.2 Fixed-Head

Test PG3 was conducted with the pile head totally fixed in position having zero

vertical or lateral movements. This simulates the condition where the pile cap is tied

with a rigid pile cap and very strong/stiff ground beams. The test results are compared

with corresponding fixed-head single pile (Tests 13, 14A). However, as noted in

Chapter 5, the pile axial force response in Test 14A is not recorded.


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Figure 6.26 shows the tunnelling-induced axial force for the front and rear piles

of a fixed-head end-bearing 2-pile group. As discussed in Chapter 5, the upper pile

shaft of a fixed-head end-bearing pile is subjected to tensile force while compression

force is induced along the lower pile shaft. Since Test PG3 is a 2-pile group, the front

pile experiences higher tensile and compression forces than that of rear pile as it is

closer to the tunnel. When compared to single pile (Test 13), it is noted that the front

pile axial force responses are smaller than that of a single pile due to the group effect

of the pile, as shown in Figure 6.27. It is noted that pile axial force is not measured in

Test 14A, thus the comparison of rear pile in Test PG3 with a single pile at the same

location is not possible.

Similarly for pile bending moment responses (Figure 6.28), the rear pile

bending moment is smaller than that of front pile as the lateral soil movement acting

on the pile is much smaller. The trend of front and rear pile bending moment profiles

are similar, suggesting that pile-cap-pile interaction is less severe. Owing to the group

and shadowing effects, the rear pile experiences smaller bending moment.

Figures 6.29 and 6.30 compare the front and rear pile bending moments with

the corresponding single piles (Tests 13 & 14A) at the same location. The results

indicate similar pile bending moment profiles for all cases with a smaller maximum

bending moment for the 2-pile group due to shadowing effects.

The single pile over pile group ratio for the front and rear piles in fixed-head

condition (Test PG2) with the corresponding single piles (Tests 13 & 14A) are plotted

in Figures 6.31 and 6.32. The comparisons summarise the positive effects of the pile


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group, with improvement ratio of 1.05 to 1.5. In general, the reduction in pile axial

tensile force and pile head bending moment are more significant than that in pile axial

compression force and mid-pile shaft bending moment. This suggests that the restraint

due to fixed-head condition is dominant.

6.4 PILE GROUP SIZE

6.4.1 Capped-Head

In order to evaluate the effect of pile group size, capped-head end-bearing 6-pile group

(Test PG4, 2x3 configurations) was performed and compared with capped-head end-

bearing 2-pile group (Test PG2, 1x2 configuration). Following the finding by Leung et

al. (2003) on excavation-induced soil movement on pile group, it is expected that a 2x2

pile group would have similar behaviour with those of a 1x2 pile group (Test PG2)

except that the responses might be smaller due to greater number of piles. It is thus

decided to investigate the 2x3 configuration (Test PG4) as the effect of pile group size

is expected to be more significant and hence providing further insight on pile group

size effect. In Test PG4, two rows of piles were arranged in three columns at 6 m

(front), 10 m (middle) and 14 m (rear) perpendicular to the tunnel. Owing to the

symmetrical arrangement of the 2x3 pile group configuration, only one of each of the

three pairs of was instrumented. The minimum boundary clearance for this pile group

is 12 m to the edge of the container in the direction perpendicular to the tunnel, and 8

m to the edge of container in the direction parallel to the tunnel. Randolph and Wroth

(1978) established that it is reasonable to assume a maximum shear stress at the pile-

soil interface and the shear stress decreases with increasing distance from the pile.

Shen (2008) reported a minimum boundary clearance of 8 m would not cause


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significant container boundary effect. Thus the container boundary effect should not be

significant for the present study. As the pile settlement for end-bearing pile is

negligible as presented in Chapter 5, the pile settlement of Test PG4 will not presented

here.

Figure 6.33 shows the tunnelling-induced axial force for the front, middle and

rear piles in the fixed-head end-bearing 6-pile group (Test PG4). The results reveal that

as the distance between the pile and tunnel increases, the pile axial force decreases as

the magnitude of soil movement becomes smaller away from the tunnel. In the short-

term, the induced maximum drag load reducing from 321 kN (front pile) to 161 kN

(middle pile), and finally to 109 kN for the rear pile. The axial forces increase over the

time for all cases. The measured maximum drag load in the long-term are 510 kN

(front pile), followed by 261 kN (middle pile) and 186 kN (rear pile). The observed

trend appears consistent with the corresponding single piles reported in Chapter 5. This

finding suggests that pile-cap-pile interaction on the pile axial force is not significant.

The changes of the induced axial forces are mainly affected by the magnitude of the

tunnelling-induced soil movement at various locations from the tunnel.

To further compare the group effects, the results of front and middle piles in the

6-pile group are compared with the corresponding capped-head end-bearing 2-pile

group (front and rear piles) in Figures 6.34 and 6.35. In general, the piles in a bigger

pile group would be subjected to lower induced axial forces. This is reasonable as a

bigger pile group is likely to provide more shadowing effect on the piles behind the

front piles and reinforcing effects to other piles in the group. The reduction in the

maximum drag load for the front pile of the 6-pile group (Test PG4) is about 10% to


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11% while the corresponding reduction for the middle pile is slightly higher of

approximately 13% to 17%.

Figures 6.36(a) and (b) show the tunnelling-induced pile bending moment for

the front, middle and rear piles in a capped-head end-bearing 6-pile group, in the short-

and long-term, respectively. The bending moment profiles generally display triple

curvatures with negative bending moments at the upper and lower portions of the pile

body, whilst positive pile bending moments occur approximately at the tunnel spring

line. It is observed that the middle and rear piles have similar induced bending moment

profiles while the front pile shows a markedly different profile. Since the front, middle

and rear piles are connected via a rigid pile cap, the front pile is being dragged back

while the middle and rear piles being pulled by front pile toward the tunnel through the

connecting rigid pile cap.

The pile-cap-pile interaction would moderate the induced pile bending

moments among the piles within a pile group, as a result, the induced pile bending

moments in the middle row is smaller than that of rear row. This is contrary to the

induced lateral soil movements, in which the corresponding lateral soil movement on

the middle row of piles is larger than the corresponding lateral soil movement on the

rear row of pile. The observation is consistent with finding reported by Leung et al.

(2003), which suggesting that part of the bending moments of the pile in the middle

row is transferred to the rear piles due to the interaction through pile cap. Similarly, the

long-term pile bending moments are showing the consistent behaviours except the

magnitudes are increased over time.


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The comparison of bending moment responses of respective piles at the same

location for the 2-pile group (Test PG2) and 6-pile group (Test PG4) are plotted in

Figures 6.37 and 6.38. The shape of bending moment profile in the front pile is similar

for both 2-pile and 6-pile group. In addition, the bending moment profiles of rear pile

in 2-pile group are similar to that of middle row and rear row of piles in 6-pile group.

Furthermore, the reduction in maximum bending moment in 6-pile group (Test PG4)

as compared to 2-pile group (Test PG2) is more significant for the middle row piles

than the front row piles in 6-pile group, which is illustrated earlier that the part of

bending moments in the middle row piles are shared by the rear row of piles. The data

demonstrated that the induced bending moment is reduced by 21% to 30% for the front

row of piles and by a considerable reduction for the middle row of piles at 27% to 54%

from the above finding. As such, the pile-cap-pile interaction is more significant in

lateral pile responses for a bigger pile group. With more piles in a bigger pile group,

the larger pile-cap-pile interaction and shadowing of front piles over rear piles

significant affect the performance of the pile group.

Figures 6.39 (a) & (b) show the tunnelling-induced pile deflection profiles for

the front, middle and rear piles in a capped-head 6-pile group (Test PG4) in the short-

and long-term, respectively. Since all 6 piles at different position are capped by a rigid

pile cap, the pile head deflection is forced to act in unison and thus the lateral

deflection at pile head moves in a translation mode with the same deflection magnitude.

The measured pile head deflection increases from 1.3 mm (short-term) to 2.7 mm

(long-term). The data reveal that the pile deflection profiles demonstrate some

differences between the front, middle and rear piles. For the front pile, the pile head is

dragged back towards the rear pile similar to the Test PG2 (see Fig. 6.40). Moreover,


224

the middle pile in the 6-pile (PG4) group shares the same shape of the deflection

profile with the rear pile in the 2-pile group (PG2) (Fig. 6.41). The comparison of pile

head deflection for different pile groups is shown in Figure 6.23. The pile head

deflection reduces from 1.9 mm to 1.3 mm (short-term) and 3.5 mm to 2.7 mm (long-

term) when the pile group size increases from 2 to 6 piles. This further demonstrates

the positive effect of pile group increases with group size.

It is acknowledged that the pile head deflection is directly proportional to the

bending moment. In other words, if the pile head deflection increases, so would the

maximum negative bending moment at the pile head. The consistent responses of

deflection with bending moment observed in the present study show that the improved

pile cap has served its intended function of providing strong pile cap fixity, with

considerably less pile cap relaxation reported by Ong (2005).

Figures 6.42 and 6.43 show the ratio of the pile group responses of the 2-pile

group over 6-pile group for the respective piles at 6 m (front) and 10 m (middle piles in

6-pile group and rear pile in 2-pile group) in order to evaluate the effect of pile group

size. The pile group effect is positive if the ratio exceeds one as there is a larger

reduction in pile responses for the larger pile group. The measured ratio of between 1.1

and 2.1 clearly shows that a larger pile group would experience a greater positive

group effect.


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6.4.2 Fixed-Head

Test PG5 with a fixed-head end-bearing 6-pile group (2x3 configurations) was

conducted and compared with the fixed-head end-bearing 2-pile group (Test PG3, 1x2

configurations). The configurations of Test PG5 are similar to Test PG4 except for the

pile head condition.

Figure 6.44 shows the tunnelling-induced axial force for the front, middle and

rear piles for the fixed-head end-bearing 6-pile group (Test TG5). It is observed that

the shapes of pile axial force profiles are similar to those of the fixed-head 2-pile group.

The results reveal that as the distance of pile from the tunnel increases, both the

induced pile tensile and compression forces reduce. This is because a smaller soil

movement is induced at location further away from the tunnel. It is also noted that the

pile-cap-pile interaction in a totally fixed-head condition is not significant. The bigger

pile group is able to provide more significant reinforcing and shadowing effects. The

results of front and middle piles in the 6-pile group are compared with the

corresponding piles in 2-pile group (Test PG3) in Figures 6.45 and 6.46. Similar to the

capped-head, the fixed-head pile group also exhibit the same positive group effect.

For the front pile, the reduction in the maximum tensile force in the short-term

for the 6-pile group (Test PG5) is about 25% with a smaller reduction of 14% for the

maximum compression force when compared with the 2-pile group (Test PG3). On the

other hand, the reduction in tensile and compression force for the middle pile is about

the same, i.e. 13% to 14% when the size of pile group increases from 2 to 6 piles.


226

Figures 6.47(a) and (b) show the tunnelling-induced pile bending moment for

the front, middle and rear piles in a fixed-head end-bearing 6-pile group (Test PG5) in

the short-and long-terms, respectively. The results reveal that the bending moment

profiles display triple curvatures with negative bending moments along the upper and

lower portions of the pile body, whilst positive pile bending moment occur

approximately at the tunnel spring line. The shape of profile is similar to the fixed-

head end-bearing 2-pile group (Test PG3) as well as fixed-head single piles (Tests 13,

14A & 14B). The results reveal that the absolute magnitude of the negative bending

moment at the pile head is larger than the positive bending moment at mid-pile shaft

suggesting a significant fixed-head effect. As before, the induced maximum positive

bending moment reduces from the front to the middle finally to the rear pile as the

lateral soil movement reduces with increasing distance between the pile and tunnel. In

addition, the front row of pile provides shadowing to the trailing middle and rear rows

pile during tunnel excavation, resulting in smaller bending moment on the trailing piles.

A comparison of bending moment profiles of respective piles at the same

location for the 2-pile group (Test PG3) and 6-pile group (Test PG5) are shown in

Figures 6.48 and 6.49. As expected, a bigger pile group provides more shadowing and

reinforcing effect to other piles within the group, resulting an average smaller pile

response due to tunnelling. The comparison reveals that the reduction in maximum

negative bending moment at the pile head is more significant than the reduction in

maximum positive bending moment at mid-pile shaft for the 6-pile group. The data

illustrates that in the short-term; the induced negative bending moment reduces by

34% and 41% for the front and middle piles, respectively. On the other hand, the

corresponding reduction of 28% and 29% is smaller for the positive bending moment.


227

Figures 6.50 and 6.51 show the ratio of pile responses of the 2-pile group over

6-pile group for the respective piles at 6 m (front) and 10 m (middle piles in 6-pile

group and rear pile in 2-pile group). It is noted that the positive group effect is

generally higher in pile bending moment than axial force, being between 1.02 and 1.33

for axial force, between 1.3 and 1.7 for bending moment.

6.5 PILE CAP CONDITIONS

6.5.1 2-Pile Group

The effect of fixity between the pile and pile cap has been studied by many researchers.

However, the effect of fixity of pile cap itself has been rarely been studied before. In

reality, the behaviours of pile group heavily depend on the pile cap fixity. Hence, Tests

PG2 (capped-head) and PG3 (fixed-head) have been conducted to study the effects of

tunnelling on 2-pile groups with capped-head and fixed-head condition.

Figures 6.52 and 6.53 show the tunnelling-induced axial force respectively for

the front and rear piles of a fixed-head and capped head end-bearing 2-pile groups. The

most distinct difference observed is that both compression and tensile forces are

induced in fixed-head pile group (PG3), while only compression force (drag load) is

induced in capped-head pile group (PG2). This is due to the total restraint provided

by the pile head in fixed-head condition, in which no vertical or lateral pile head

movement is allowed, as explained in Chapter 5. As tensile force is induced along the

upper pile shaft due to total pile cap fixity in Test PG3, there is a reduction in the

maximum drag load of about 20% in the front pile and about 15% in the rear pile. Thus,


228

Figures 6.56 and 6.57 show the induced pile responses of capped-head pile over fixed-

head pile ratio for the front and rear piles. The figures reveal that a fixed-head is

beneficial for induced pile axial force, or reduction in maximum drag load, with the

ratio of axial force of capped-head pile over fixed-head pile of 1.15 to 1.24.

The induced 2-pile group bending moments in capped-head (Test PG2) are

compared with those in fixed-head (Test PG3) in Figures 6.54 and 6.55. For the front

pile, it is noted that the pile bending moment profiles for the 2 pile cap conditions are

not similar, particularly at the pile head. In fixed-head condition (Test PG3), the profile

is very much similar to those single fixed-head piles with the maximum bending

moment occurring at the pile head, but it is different for capped-head condition (Test

PG2) due to tilting of cap resulting in a downward shift of the location of maximum

bending moments. This is probably due to the different pile cap fixity condition,

whereby capped-head demonstrates a significant pile-cap-pile interaction while for

fixed-head condition, the pile-soil-pile interaction is dominant and thus suggesting that

pile-cap-pile interaction is less severe as the cap is totally fixed in the position. Owing

to total restraint at the pile head, the induced negative bending moment of fixed-head

pile group is much larger than that of capped-head pile group. The capped-head pile is

allowed to deflect freely and interact with other piles in a group and thus inducing a

smaller negative bending moment compared to the fixed-head pile. On the other hand,

the profiles of rear pile bending moments for capped- and fixed-pile are similar. It is

also worth noting that for fixed-head, the negative bending moment induced at pile

head is always larger than the positive bending moment induced at the mid-pile shaft

because the fixed-head condition has provided a very rigid restraint at the pile head

which dominates the lateral pile responses.


229

As shown in Figures 6.56 and 6.57, the maximum positive and negative

bending moments of capped-head pile over fixed-head pile ratio for the front and rear

piles experiencing opposite trends. Generally, the induced positive pile bending

moments near mid-pile shaft is beneficial to a fixed-head pile group, with the capped-

head pile over fixed-head pile ratio of 1.06 to 1.28. Paradoxically, at the pile head, the

negative pile bending moment ratio of capped-head pile over fixed-head pile is less

than 1, at about 0.4 to 0.5 for front pile and 0.7 for rear pile. This indicates an adverse

effect on the fixed-head pile group, which can be explained by the totally fixed-head

condition, which restraints the pile movement and thus induces the high bending

moment. It is thus important to check the adequacy of the steel reinforcement for the

different pile cap fixity, i.e. it is critical at the pile head for fixed-head pile group and

at the mid-pile shaft for the capped-head pile group.

6.5.2 6-Pile Group

The results of 6-pile group, i.e. Tests PG4 (capped-head) and PG5 (fixed-head) are

compared in this section. Figures 6.58, 6.59 and 6.60 show the tunnelling-induced

axial force for the front, middle and rear piles of a fixed-head and capped head end-

bearing 6-pile groups, respectively. Similar to 2-pile group, compression and tensile

forces are induced in fixed-head pile (PG5). Likewise, the compression forces induced

in the fixed-head pile (PG5) are always smaller than that of capped-head pile (PG4)

with a reduction ranging from 10% to 22%. Nevertheless, the trade-off for such

reduction in drag load is that tensile force is induced near to the pile head due to totally

fixed head condition.


230

Unlike axial force, the behaviour of bending moment is much more

complicated. The results of induced 6-pile group bending moments in capped-head

(Test PG4) are compared with those in fixed-head (Test PG5) in Figures 6.61, 6.62 and

6.63. The front pile (Fig. 6.61) exhibits different bending moment profiles for both

capped-head and fixed-head conditions, similar to the comparison of 2-pile group

discussed earlier. It is observed the upper portion of the front pile is being dragged

backward in Test PG4 (capped-head) while the profile of the front pile in Test PG5

(fixed-head) is similar to those single fixed-head piles. On the other hand, all of the

middle and rear row piles (Figs. 6.62 and 6.63) are noted to bend toward the tunnel

(similar to the front pile in the fixed-head pile group) regardless of pile cap conditions.

Despite the shape of bending moment profiles are similar in middle and rear row piles,

the comparison of capped- and fixed-head conditions illustrates a very different

behavior in transferring the induced bending movements within a bigger group of 6-

pile, as oppose to 2-pile group. Intuitively the induced lateral soil movement is largest

at the position of front pile, followed by middle and rear pile. It is thus expected that

the magnitude of the induced pile bending moment should follow a similar trend of

soil movements. Nevertheless, this trend is only observed in the fixed-head 6-pile

group (Test PG5) but not the capped-head 6-pile groups (Test PG4) because

moderation effect is negligible for fixed-head pile group and the profiles follow those

of single fixed-head pile with the magnitudes reduce with increase of tunnel-pile

distance. However, as discussed in Section 6.4.1, for Test PG4 (capped-head 6-pile

group), the induced pile bending moments in the middle row is smaller than that of

rear row. It is observed that since the pile cap is tilted slightly in the capped-head pile

group, the pile-cap-pile interaction would moderate the induced pile bending moments

among the piles within a pile group. As a result, part of the bending moments of the


231

pile in the middle row is transmitted to the rear piles due to interaction through pile cap.

In addition, the middle row piles are shielded in the 6-pile group and experiencing the

least induced pile bending moment.

The above-mentioned behaviors can be further illustrated in Figures 6.64 &

6.65. It appears that the maximum positive and negative pile bending moments reduce

almost linearly from the front row piles to the middle row piles and finally to the rear

row piles in Test PG5 (fixed-head). This is consistent with the trend of observed lateral

soil movement which reduces when the distance to tunnel increases. This suggests that

in a fixed-head 6-pile group (Test PG5), the pile-soil-pile interaction is dominant and

largely depends on the soil movement instead of pile-cap-pile interaction. Nonetheless,

for the capped-head 6-pile group in Test PG4, the front pile registered the highest

induced maximum bending moment, followed by the rear pile and finally the smallest

bending moment was noted for the rear pile. This observation further confirms that

pile-cap-pile interaction is significant in capped-head fixity, especially when the pile

group increases from 2-pile (Test PG-2) to 6-pile (Test PG4). The interaction among

the piles in a bigger group would moderate, transfer and share the induced responses

among the piles particularly for the pile bending moments.

Figures 6.66, 6.67 and 6.68 show the capped-head pile over fixed-head pile

ratio. It is revealed that maximum positive bending moments induced at the mid-pile

shaft for capped-head (PG4) is always larger than that of fixed-head (PG5), thus

registering a positive effect for fixed-head pile with the ratio ranging from 1.1 to 1.2

for the front and middle piles and a higher ratio ranging from 1.9 to 2.6 for the rear pile

due to the moderating effects in capped-head pile, whereby the rear pile in capped-


232

head is sharing some bending moment from middle pile while rear pile in the fixed-

head induced the smallest bending moment due to distance effect. Paradoxically, the

induced maximum negative bending moments induced near to pile head are expected

to be larger in the fixed-head pile (PG5) than capped-head pile (PG4) due to the

restraint enforced at the pile head. Thus the ratio of capped-head over fixed-head is

about 0.5 to 0.75. However, it is contrary for the rear piles, whereby a ratio of 1.5 to

1.6 is observed. This means that the induced negative bending moment at capped-head

is larger than that in fixed-head. It is attributed to the fact that the rear pile is inducing

higher bending moment than middle pile in the capped-head piles (PG4).

From the above findings, it is postulated that the bending moment transfer

mechanism for a capped-head 6-pile group (Test PG4) is very different from the fixed-

head pile. When the capped-head pile group is tilted due to tunnelling, the front row

piles tend to bend the most toward tunnel direction but the upper portion of the

bending moment profile is being dragged back by the by middle and rear row piles via

the connecting pile cap. It is postulated that the rear row piles behave like ‘passive’

pile when pile cap tilts and bends toward the tunnel. Thus, a pile cap tends to transmit

more bending moment from the middle pile to the rear pile which would otherwise be

less affected by the tunnelling-induced soil movement. Contrary, all of the piles in

fixed-head (Test PG5) 6-pile group behave like single piles standing side by side

without direct pile-cap-pile interaction, except that the magnitude is affected by the

total number of piles because the behavior is largely governed by the pile-soil-pile

interaction.


233

Figure 6.69 shows the summary of the ratio of pile responses of long-term to

short-term pile group responses for all tests presented in this chapter. Similar plot for

single pile responses have been presented in Section 5.5. The data showing the long-

term effects with the long-term over short-term ratio of 1.32 to 2.4 as oppose to the

single pile long-term over short-term ratio of wider range of 1.34 to 3.5. This

comparison again confirm the long-term time effects of pile responses due to

tunnelling, regardless of single or group, pile cap condition or group size.


This chapter presents the results of five centrifuge model tests on pile groups with

different number of piles, pile cap and pile tip condition.

In the case of a floating capped-head pile group (Test PG1), the pile group is

generally beneficial as the average pile group responses are smaller than the average of

those of single piles at the same locations. This is because more efforts are required to

drag or bend the entire pile group including the pile cap. When a pile group gets larger,

the induced pile responses become smaller, in which shadowing and reinforcing effects

are dominant, thus diminishing the effects of induced soil movements acting on the

piles. When the pile toe condition changes to end-bearing (Test PG2) with a short

socket, the behavior is totally different. The pile movements reduce substantially, but

the trade-off is that intensive structural responses are induced in term of axial forces

and bending moments at the upper shaft and pile cap.


234

The scenario becomes more complicated if different pile cap conditions are

modeled, as the head conditions play vital roles in dictating the pile responses.

Generally, capped-head piles (Test PG2 (2-pile group) & Test PG4 (6-pile group))

demonstrate significant pile-cap-pile interaction among the piles. Contrary, the fixed-

head piles (Test PG3 (2-pile group) and Test PG5 (6-pile group)) behave like single

piles standing side by side without direct pile-cap-pile interaction; expect that the

magnitude is affected by the total number of piles because the behavior is largely

governed by the pile-soil-pile interaction.

In addition, once the pile group size increases from 2 to 6 piles, the position of

the pile within a group demonstrates a totally different transfer mechanism in the

lateral pile responses, as compared to that of a single pile with responses reducing with

increasing distance of pile to tunnel. The pile-cap-pile interaction in capped-head 6-

pile group (Test PG4) would moderate the induced pile bending moments among the

piles within a pile group. As a result, the induced pile bending moments in the middle

row is smaller than that of rear row. This is contrary to the induced lateral soil

movements, in which the corresponding lateral soil movement on the middle row piles

is larger than the corresponding movement on the rear row piles. This suggests that

part of the bending moments of the middle row piles is transferred to the rear piles due

to the interaction through the pile cap. It is worth noting that the axial forces reduce

when the distance between the pile and tunnel increases. It is thus suggested that the

pile-cap-pile interaction in capped-head 6-pile group (Test PG4) is less significant in

the axial force as compared to bending moment and the induced pile axial forces are

mainly influenced by the soil settlement and the distance between tunnel and pile.


235

On the other hand, for the piles in fixed-head 6-pile group (Test PG5), the piles

behave like single piles in term of axial force and bending moment, except that the

magnitude is affected by the total number of piles. Moreover, the results show that the

fixed-head condition reduces the pile axial force, or in other words, a reduction in the

maximum drag load, but with tensile forces induced along the upper pile shaft due to

the total fixity at the pile cap. In contrast, there is an increment in the pile bending

moment. This is due to the totally fixed-head condition, which restrains the pile

movement causing a high pile bending moment.

A common trend has been observed for the long-term over short-term ratio of

pile responses for both single pile and pile group. The results reveal that soil and pile

responses increase over time with long-term over short-term pile responses ratio

ranging from 1.32 to 2.4, regardless of pile size, pile head and toe conditions.

Chapter 6 Effects of Tunneling on Pile Groups

236

Table 6.1 Test program and prototype parameters for pile group tests

Test Ref.

Plan View Elevation View Parameter Pile head & toe

conditionPG1

A=6m B=10m C=12m D=6m L=22m

Capped head Floating pile 2-pile group

PG2

A=6m B=10m C=12m D=6m L=27.5m

Capped head End-bearing pile 2-pile group

PG3

A=6m B=10m C=12m D=6m L=27.5m

Fixed head End-bearing pile 2-pile group

PG4

A=6m B=10m C=14m D=6m L=27.5m

Capped head End-bearing pile 6-pile group

PG5

A=6m B=10m C=14m D=6m L=27.5m

Fixed head End-bearing pile 6-pile group

AB

D

C

L

‘Axial’ Pile ‘Bending’ Pile

E

Tunnel

FP MP RP A

BC

AB

D

C

L


E

Tunnel

FP MP RP A

BC

AB

D

C

L

‘Axial’ Pile‘Bending’ Pile Tunnel

FP RP A

B

AB

D

C

L

‘Axial’ Pile ‘Bending’ Pile Tunnel

FP RP A

B

Tunnel

FP RP A

B AB

D

C

L


Tie Beam

Tie Beam

Tie Beam

Tie Beam


237

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600

Test PG1, Front (ST)

Test PG1, Front (LT)

Test PG1, Rear (ST)

Test PG1, Rear (LT)

Tunnel

Dep

th (m

)



238

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600

Test PG1, Front (ST)Test PG1, Front (LT)Test 3, Single, X=6m(ST)Test 3, Single, X=6m (LT)

Tunnel

Dep

th (m

)

Distance of pile fromtunnel centre = 6 m

Figure 6.2 Tunnelling-induced front pile (Test PG1) and corresponding single pile

(Test 3) axial force Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600

Test PG1, Rear (ST)

Test PG1, Rear (LT)

Test 16, Single, X=10m (ST)

Test 16, Single, X=10m (LT)

Tunnel

Dep

th (m

)


Figure 6.3 Tunnelling-induced rear pile (Test PG1) and corresponding single pile (Test

16) axial force


239

0

4

8

12

16

20

Single pile, front,Test 3 (X=6m)

Single pile, rear,Test 16 (X=10m)

2-pile group, front,Test PG1 (X=6m)

2-pile group, rear,Test PG1(X=10m)

ST (2 days) L (720 days)T

Pile

hea

d se

ttlem

ent (

mm

)

Figure 6.4 Tunnelling-induced pile head settlement (Tests PG1, 3, 16)


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250

Test PG1, Front (ST)Test PG1, Front (LT)Test PG1, Rear (ST)Test PG1, Rear (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel



240


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250

Test PG1, Front (ST)Test PG1, Front (LT)Test 3, Single, X=6m (ST)Test 3, Single, X=6m (LT)

Dep

th (m

)

Tunnel



(Test 3) bending moment


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250

Test PG1, Rear (ST)Test PG1, Rear (LT)Test 16, Single, X=10m (ST)Test 16, Single, X=10m (LT)

Dep

th (m

)

Tunnel


Figure 6.7 Tunnelling-induced rear pile (Test PG1) and corresponding single pile (Test

16) bending moment


241

Pile deflection (mm)

-30

-25

-20

-15

-10

-5

00 1 2 3 4 5 6 7 8

Dep

th (m

)

PG1, Front (ST)

PG1, Front (LT)

PG1, Rear (ST)

PG1, Rear (LT)

Tunnel

Figure 6.8 Tunnelling-induced pile deflection (Tests PG1)


242

Pile deflection (mm)

-30

-25

-20

-15

-10

-5

00 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

Dep

th (m

)

PG1, Front (ST)

PG1, Front (LT)



Tunnel



(Test 3) deflection Pile deflection (mm)

-30

-25

-20

-15

-10

-5

00 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15 16

Dep

th (m

)

PG1, Rear (ST)

PG1, Rear (LT)



Tunnel


Figure 6.10 Tunnelling-induced rear pile (Test PG1) and corresponding single pile

(Test 16) deflection


243

0

2

4

6

8

10

12

14

16

Single pile, front, Test 3 Single pile, rear, Test 16 2-pile group,Test PG1

ST (2 days) LT (720 days)

Pile

hea

d de

flect

ion

(mm

)

Figure 6.11 Tunnelling-induced pile head deflection (Test PG1, 3 & 16)


244

0

0.5

1

1.5

2

2.5

3

Pile axialforce

Pile headsettlement

Pile negativebendingmoment

Pile postivebendingmoment

Pile headdeflection


Sing

le p

ile/ p

ile g

roup

ratio Positive

effect ofpile group


0

0.5

1

1.5

2

2.5

3

Pile axialforce

Pile headsettlement

Pile negativebendingmoment

Pile postivebendingmoment

Pile headdeflection


Sing

le p

ile/ p

ile g

roup

ratio

Positiveeffect of pilegroup



245

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 200 400 600 800 1000 1200



Test PG2, Rear (ST)

Test PG2, Rear (LT)

Tunnel

Dep

th (m

)



246

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 200 400 600 800 1000 1200



Test 10, Single, X=6m(ST)


Tunnel

Dep

th (m

)



(Test 10) axial force Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 200 400 600 800 1000 1200

Test PG2,Rear (ST)

Test PG2, Rear (LT)



Tunnel

Dep

th (m

)



(Test 11) axial force


247


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160



Test PG2, Rear (ST)

Test PG2, Rear (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel



248


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250



Test 10, Single, X=6m (ST)Test 10, Single, X=6m (LT)

Dep

th (m

)

Tunnel



(Test 10) bending moment Bending Moment (kNm)

-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200 250

Test PG2, Front (ST)Test PG2, Rear (LT)Test 11, Single, X=10m (ST)Test 11, Single, X=10m (LT)

Dep

th (m

)

Tunnel



(Test 11) bending moment


249


-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4 5 6

Dep

th (m

)



Test PG2, Rear (ST)Test PG2, Rear (LT)

Tunnel`

Figure 6.20 Tunnelling-induced pile deflection (Test PG2)


250


-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4 5 6

Dep

th (m

)

Test PG2, Front (ST)Test PG2, Front (LT)Test 10, Single, X=6m(ST)Test 10, Single, X=6m(LT)

`

Tunnel



(Test 10) deflection Lateral deflection (mm)

-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4 5 6

Dep

th (m

)

Test PG2, Rear (ST)

Test PG2, Rear (LT)


Test 11, Single, X=10m(LT)

``Tunnel




251

0

1

2

3

4

5

6

Single pile,Test 10 (X=

6m)

Single pile,Test 11(X=10m)


2-pilegroup,Test

PG2

6-pilegroup,Test

PG4

ST

Pile

hea

d de

flect

ion

(mm

)

(a) Short-term

0

1

2

3

4

5

6

Single pile,Test 10 (X=

6m)



2-pilegroup,Test

PG2

6-pilegroup,Test

PG4

LT

Pile

hea

d de

flect

ion

(mm

)

(b) Long-term

Figure 6.23 Tunnelling-induced pile head deflection in the (a) short-term (b) long-term (Tests PG2, 10 & 11)


252

0

0.5

1

1.5

2

2.5

3

Pile axial force Pile negativebending moment

Pile positivebending moment

Pile headdeflection

ST LT

Sing

le p

ile/ p

ile g

roup

ratio

Positiveeffect ofpile group


0

0.5

1

1.5

2

2.5

3

Pile axial force Pile negativebending moment


Pile headdeflection

ST LT

Sing

le p

ile/ p

ile g

roup

ratio




253

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600 700 800

Test PG3, Front (ST) Test PG3, Front (LT)

Test PG3, Rear (ST) Test PG3, Rear (LT)

Dep

th (m

)

Tunnel



254

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600 700 800

Test PG3, Front (ST)Test PG3, Front (LT)Test 13, Single, X=6m (ST)Test 13, Single, X=6m (LT)

Dep

th (m

)

Tunnel




255


-30

-25

-20

-15

-10

-5

0-200 -150 -100 -50 0 50 100 150 200



Test PG3, Rear (ST)

Test PG3, Rear (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel



256


-30

-25

-20

-15

-10

-5

0-200 -150 -100 -50 0 50 100 150 200

Test PG3, Front(ST)Test PG3, Front(LT)Test 13, Single,X=6m (ST)Test 13, Single,X=6m (LT)

Dep

th (m

)

Tunnel



(Test 13) bending moment Bending Moment (kNm)

-30

-25

-20

-15

-10

-5

0-200 -150 -100 -50 0 50 100 150 200

Test PG3, Rear(ST)Test PG3, Rear(LT)Test 14, X=10m(ST)Test 14, Single,X=10m (LT)

Dep

th (m

)

Tunnel



(Test 14A) bending moment


257

0

0.20.4

0.6

0.81

1.2

1.4

1.61.8

2

Pile axial tensileforce

Pile axialcompression

force

Pile negativebending moment


ST LT

Sing

le p

ile/p

ile g

roup

ratio



0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2

Pile negative bending moment Pile positive bending moment

ST LT


Sing

le p

ile/p

ile g

roup

ratio

Figure 6.32 Single pile over pile group ratio for rear pile (Test 14A/ PG3)


258

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600


Test PG4, Middle (ST) Test PG4, Middle (LT)


Tunnel

Dep

th (m

)



259

Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600 700 800



TunnelD

epth

(m)



Axial Force (kN)

-30

-25

-20

-15

-10

-5

00 100 200 300 400 500 600 700 800

Test PG2,Rear (ST) Test PG2, Rear (LT)

Test P4, Middle (ST) Test PG4, Middle (LT)

Tunnel

Dep

th (m

)


Figure 6.35 Tunnelling-induced rear pile in 2-pile group (Test PG2) and corresponding

middle pile in 6-pile group (Test PG4) axial force


260


-30

-25

-20

-15

-10

-5

0-60 -40 -20 0 20 40 60 80 100


Test PG4, Middle (ST)

Test PG4, Rear (ST)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel

(a) Short-term


-30

-25

-20

-15

-10

-5

0-60 -40 -20 0 20 40 60 80 100 120


Test PG4, Middle (LT)

Test PG4, Rear (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel

(b) Long-term

Figure 6.36 Tunnelling-induced pile bending moment (a) in the short-term (b) in the long-term (Test PG4)


261


-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200

Test PG4, Front (ST)Test PG4, Front (LT)


Dep

th (m

)

Tunnel




-30

-25

-20

-15

-10

-5

0-100 -50 0 50 100 150 200

Test PG4, Middle (ST)Test PG4, Middle (LT)Test PG2, Rear (ST)Test PG2, Rear (LT)

Dep

th (m

)

Tunnel



middle pile in 6-pile group (Test PG4) bending moment


262


-30

-25

-20

-15

-10

-5

0-1.5 -1 -0.5 0 0.5 1 1.5 2 2.5 3

Dep

th (m

)



Test PG4, rear (ST)

Tunnel`

(a) Short-term


-30

-25

-20

-15

-10

-5

0-1.5 -1 -0.5 0 0.5 1 1.5 2 2.5 3

Dep

th (m

)



Test PG4, rear (LT)

Tunnel`

(b) Long-term

Figure 6.39 Tunnelling-induced pile deflection in the (a) short-term (b) long-term (Test PG4)


263


-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4

Dep

th (m

)





Tunnel

`


Figure 6.40 Tunnelling-induced pile deflection (Tests PG2 and PG4)


-30

-25

-20

-15

-10

-5

0-2 -1 0 1 2 3 4

Dep

th (m

)

Test PG2, Rear (ST)

Test PG2, Rear (LT)



Tunnel`


Figure 6.41 Tunnelling-induced pile bending moment (Tests PG2 and PG4)


264

00.20.40.60.8

11.21.41.61.8

22.22.4

Pile axial force Pile negative bendingmoment

Pile positive bendingmoment

ST LT

2-Pi

le G

roup

/ 6-P

ile G

roup

ratio Positive

effect ofpile group

c


00.20.40.60.8

11.21.41.61.8

22.22.4

Pile axial force Pile negative bendingmoment


ST LT


2-Pi

le G

roup

/ 6-P

ile G

roup

ratio



265

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600




Dep

th (m

)

Tunnel



266

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600



TunnelD

epth

(m)


Figure 6.45 Tunnelling-induced front pile in 2-pile group (Test PG3) and

corresponding front pile in 6-pile group (Test PG5) axial force Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600



Dep

th (m

)

Tunnel



middle pile in 6-pile group (Test PG5) axial force


267


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160



Test PG5, Rear (ST)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel

(a) Short-term Bending Moment (kNm)

-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160



Test PG5, Rear (LT)

Dep

th (m

)

Tunnel

Dep

th (m

)

Tunnel

(b) Long-term

Figure 6.47 Tunnelling-induced pile bending moment in the (a) short-term (b) long-term (Test PG5)


268


-30

-25

-20

-15

-10

-5

0-120 -80 -40 0 40 80 120 160 200



Dep

th (m

)

Tunnel




-30

-25

-20

-15

-10

-5

0-120 -80 -40 0 40 80 120 160 200

Test PG5, Middle (ST)Test PG5, Middle (LT)Test PG3, Rear (ST)Test PG3, Rear (LT)

Dep

th (m

)

Tunnel



middle pile in 6-pile group (Test PG5) bending moment


269

0

0.20.4

0.6

0.81

1.2

1.4

1.61.8

2



force



ST LT

2-Pi

le G

roup

/ 6-P

ile G

roup

ratio



0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2



force



ST LT


2-Pi

le G

roup

/ 6-P

ile G

roup

ratio



270

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600 700 800



Tunnel

Dep

th (m

)




271

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600 700 800



Tunnel

Dep

th (m

)


Figure 6.53 Tunnelling-induced rear pile in capped-head 2-pile group (Test PG2) and

corresponding rear pile in fixed-head 2-pile group (Test PG3) axial force


272


-30

-25

-20

-15

-10

-5

0-150 -100 -50 0 50 100 150 200



Dep

th (m

)

Tunnel


Figure 6.54 Tunnelling-induced front pile in capped-head 2-pile group (Test PG2) and corresponding front pile in fixed-head 2-pile group (Test PG3)bending moment


-30

-25

-20

-15

-10

-5

0-150 -100 -50 0 50 100 150 200

Test PG3, Rear (ST)Test PG3, Rear (LT)Test PG2, Rear (ST)Test PG2, Rear (LT)

Dep

th (m

)

Tunnel



corresponding rear pile in fixed-head 2-pile group (Test PG3) bending moment


273

0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2

Pile compressionforce

Pile negative bendingmoment


ST LT

Cap

ped-

head

pile

/ Fix

ed-h

ead

pile

Positiveeffect offixed-headpile


0

0.2

0.4

0.6

0.8

1

1.2

1.4

1.6

1.8

2

Pile compressionforce

Pile negative bendingmoment


ST LT

Positiveeffect offixed-headpile

Cap

ped-

head

pile

/ Fix

ed-h

ead

pile



274

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600



TunnelD

epth

(m)




275

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600



TunnelD

epth

(m)


Figure 6.59 Tunnelling-induced middle pile in capped-head 6-pile group (Test PG4)

and corresponding middle pile in fixed-head 6-pile group (Test PG5) axial force

Axial Force (kN)

-30

-25

-20

-15

-10

-5

0-200 -100 0 100 200 300 400 500 600



Tunnel

Dep

th (m

)



corresponding rear pile in fixed-head 6-pile group (Test PG5) axial force


276


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160



Dep

th (m

)

Tunnel


Figure 6.61 Tunnelling-induced front pile in capped-head 6-pile group (Test PG4) and corresponding front pile in fixed-head 6-pile group (Test PG5) bending moment


277


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160

Test PG4, Middle (ST)Test PG4, Middle (LT)Test PG5, Middle (ST)Test PG5, Middle (LT)

Dep

th (m

)

Tunnel


Figure 6.62 Tunnelling-induced middle pile in capped-head 6-pile group (Test PG4)

and corresponding middle pile in fixed-head 6-pile group (Test PG5) bending moment


-30

-25

-20

-15

-10

-5

0-80 -40 0 40 80 120 160

Test PG4, Rear (ST)Test PG4, Rear (LT)Test PG5, Rear (ST)Test PG5, Rear (LT)

Dep

th (m

)

Tunnel



corresponding rear pile in fixed-head 6-pile group (Test PG5) bending moment


278

-40

-30

-20

-10

0

10

20

30

40

Capped-head (PG4), ST Fixed-head (PG5), ST

Capped-head (PG4), ST Fixed-head (PG5), ST

Front Middle Rear

Figure 6.64 Variation of maximum bending moment for front, middle and rear pile in the short-term (Tests PG4 and PG5)

-60

-50

-40

-30

-20

-10

0

10

20

30

40

50

60

70

80

Capped-head (PG4), LT Fixed-head (PG5), LT

Capped-head (PG4), LT Fixed-head (PG5), LT

Front Middle Rear

Figure 6.65 Variation of maximum bending moment for front, middle and rear pile in the long-term (Tests PG4 and PG5)


279

0

0.5

1

1.5

2

2.5

3



force



ST LT

Positive effect offixed-head

Cap

ped-

head

pile

/ Fix

ed-h

ead

pile

ratio


0

0.5

1

1.5

2

2.5

3



force



ST LT


Cap

ped-

head

pile

/ Fix

ed-h

ead

pile

rati

Figure 6.67 Capped-head pile over fixed-head pile ratio (middle pile, Test PG4/PG5)

0

0.5

1

1.5

2

2.5

3



force



ST LT

Cap

ped-

head

pile

/ Fix

ed-h

ead

pile

ratio




280

0

0.5

1

1.5

2

2.5



force



Pile headdeflection

PG1, front PG1, rear PG2, front PG2, rear PG3, front PG3, rear

PG4, front PG4, middle PG4, rear PG5, front PG5, middle PG5, rear

LT/S

T ra

tio

Long-termeffect

Figure 6.69 Long-term over short-term ratio (Tests PG1, 2, 3, 4 & 5)

Chapter 7 Conclusions

281

CHAPTER SEVEN

CONCLUSIONS


The overall purpose of this research study was to investigate tunnel-soil-pile

interaction in soft clay. The objectives were achieved through (a) the development of a

centrifuge tunnel excavation technique to simulate the inward tunnel deformation

pattern commonly observed in practice (b) a centrifuge model study of tunnelling-

induced soil movements in free-field analysed using Particle Image Velocimetry (PIV)

technique (c) a centrifuge model study of tunnelling-induced single pile responses, and

(d) a centrifuge model study of tunnelling-induced pile group responses.

In this study, a series of centrifuge model tests were conducted to investigate

the effects of tunnelling on soft clay, single piles and pile groups in clay. A total of

twenty one centrifuge model tests - Tests 1 and 2 (tunnel-soil interaction), Tests 3 to

16 (tunnel-single pile interaction) and Tests PG1 to PG5 (tunnel-pile group interaction)

- were performed. The effects of factors such as volume loss, pile tip and head

condition, pile length, pile-to-tunnel distance, floating and end-bearing pile groups,

size of pile group and pile groups with capped-head and fixed-head conditions of pile

groups on pile due to tunneling were examined. In addition, the observed pile

behaviours are evaluated against the measured free-field soil movements due to

tunnelling.


282

7.1.1 Technique for Simulation of Tunnelling

In the present centrifuge model study, an innovative tunnelling simulation technique

was developed to simulate the inward tunnel deformation due to over-excavation

commonly observed in practice. An oval-shape ground deformation pattern is imposed

as the boundary condition and the gap parameter (GAP) proposed by Lee et al. (1992)

is used to quantify the amount of tunnel over-cut. The experimental results from

tunnel-soil interaction presented in Chapter 4 provided clear evidence that the present

model tunnel was able to simulate the precise volume loss during the tunnelling

process. Moreover, this model could provide a very uniform oval-shape of the GAP

throughout the entire length of the model tunnel. This ensured that the volume loss was

constant along the model tunnel. Such simulation technique provides flexibility and

reproducibility to study the tunnel-soil-pile interaction in a centrifuge model that could

not be measured in field tests, particularly the time effects of soft clay.

However, the simplified two-dimensional model tunnel was unable to simulate

the complex three-dimensional effects of tunnelling before and after the passing of the

tunnel boring machine (TBM). Nevertheless, in a situation where the tunnel excavation

has passed a particular section, the vectors of the ground movement developed will be

more or less in the plane perpendicular to the tunnel axis. Consequently it is reasonable

to assume that a plane strain model of long tunnel section would be a reasonable

representation of tunnelling-induced soil movements; this is usually referred to as a

two-dimensional simulation (Taylor, 1998). Moreover, the results from Taylor (1998)

and the present study showed that the two-dimensional model tunnel was able to

measure the transverse responses of soils and piles.


283

7.1.2 Tunnel-Soil Interaction

The centrifuge model tests with the application of PIV had provided a considerable

body of data to examine the patterns of soil movement induced by tunnelling in soft

clay. The surface settlement trough in clay generally follows the Gaussian distribution

curve in the short term. The magnitude of maximum ground surface settlement

increases with time and tunnel volume loss. The settlement magnitude is larger in the

long-term and the settlement trough is wider as compared to that in the short-term. The

observed data confirmed that the empirical equation proposed by Mair et al (1993) is

applicable in the prediction of the subsurface settlement troughs in clay in the short-

term. Empirical equations in the short-term and long-term were proposed for the

distribution of inflection point in soft clay.

In the short-term, an “Immediate Shear Zone” with large soil movement above

the tunnel can be identified, while the zone outside the immediate shear zone may be

identified as the ‘Support Zone’. In the long term, the significant soil movement zone

extends much wider. In addition, soil settlement was noted to be more dominant than

lateral soil movement in the long term.

7.1.3 Tunnel-Single Piles Interaction

During tunnelling, the ground around the tunnel often moves towards the tunnel

opening. The resulting ground movements induce additional axial (settlement and axial

force) and lateral (deflection and bending moment) responses on adjacent pile

foundations.


284

Test Series 1 studied the effects of volume loss on pile performances. It was

found that the induced pile bending moment triples and the pile settlement and

deflection increase by almost 2.5 times when volume loss increases from 3% to 6.5%,

in this particular case.

For Test Series 2, three different pile tip conditions, namely “floating” pile,

“socketed” pile and “end-bearing” pile were investigated to study the effects of pile tip

condition. It is noted that a floating pile is mainly governed by pile settlement when

tunnelling is carried out adjacent to it whereas socketed piles are likely to be governed

by the material stress of the pile. On the other hand, some opposite trends were

observed in the fixed-head when compared to free-head. It is noted that tensile force

and relatively large negative bending moments were induced at the pile head due to

total fixity. Nevertheless, these responses had led to the reduction in drag load and

positive bending moment at the pile waist.

In a short pile, with pile base at or above the tunnel crown, especially those

located in the immediate shear zone, the pile structural responses (axial force and

bending moment) were of less significant if compared to those of long pile. However,

there will be excessive pile movements (settlement and deflection) because of lack of

anchorage of pile into the stable support zone. In this respect, a longer pile with pile

length in the stabilised support zone tends to provide more resistance to the pile

movements but will attract more bending moment and axial force.

Test Series 4, 5 and 6 examined the effects of pile-to-tunnel distance for

different pile head and tip conditions. Generally, it was observed that the pile


285

responses decrease with increase in pile-to-tunnel distance. The induced pile axial

forces were observed to decrease fairly linearly with increase in pile-to-tunnel distance

for Test Series 4 (floating piles) and 5 (end-bearing pile) and the axial force in Test

Series 5 is always much higher than that of Test Series 4 for all pile to tunnel distance.

In addition, the most significant difference in Test Series 5 (free-head) and Test Series

6 (fixed-head) is that tension force is induced in the fixed-head pile due to total fixed

condition at the pile head. Generally, the maximum induced bending moments

decrease fairly linearly with increasing pile-to-tunnel distance when the magnitude is

relatively small as in the case of pile bending moment in the short-term and positive

bending moment observed in Series 6. However, the bending moments decrease

exponentially when the magnitude is relatively large. Generally, the induced bending

moments in end-bearing piles (Test Series 5) are larger than the floating piles (Test

Series 4) as the restraint at the pile toe would restrict the pile lateral movement and

induce a larger bending moment. On the other hand, the results revealed that negative

bending moments were induced at pile head due to total fixity condition for fixed-head

piles in Test Series 6 as compared to the free-head piles in Test Series 5. As the

bending moment profile was offset toward the negative bending moment for fixed-

head pile, the positive bending moment in fixed-head pile (Test Series 6) are

consistently lower than that of free-head pile (Test Series 5). It can be established from

the test results that induced bending moments are generally small beyond a horizontal

offset of 2D from the tunnel centre. Generally, it was observed that the pile deflection

dropped rapidly from 1D to 1.5D, with a much smaller decrease from 1.5D to 2D for

both Series 4 and 5. This is because the lateral soil movements decrease with

increasing distance of pile location to the tunnel. The pile head deflection for end-

bearing piles (Test Series 5) was smaller that of floating piles (Test Series 4),


286

regardless of the pile-tunnel distance, as the lower portion of the pile was restrained

and would not moves.

Some similarities and differences were drawn in the comparisons of soil and

single pile behaviour in the cases of both inward (present study) and outward (Ran,

2004) tunnel deformations. It is noted that the measured short-term surface settlement

trough follows the Gaussian distribution curve fairly well with the inflection point (i)

at approximately 7.5 m for both tunnel deformation cases. The most distinct difference

in the soil behaviour was that the soil moved significantly away from the tunnel in the

case of outward tunnel deformation, whereas in the case of inward tunnel deformation,

the soil moved towards the tunnel. It was revealed that the pile axial force and pile

settlement behaviour and profile were fairly similar regardless of the deformation

pattern but the outward tunnel formation would induce larger pile responses as

compared to the inward tunnel deformation under the same volume loss. On the other

hand, the pile lateral responses (bending moment and deflection) were opposite in

direction for both inward and outward tunnel deformations, respectively, in terms of

profiles and magnitude.

7.1.4 Tunnel-Pile Groups Interaction

In the case of a capped-head floating pile group (Test PG1), the pile group was

generally beneficial as the average pile group responses (bending moments, axial,

settlement and lateral deflection) are smaller than the average of those of single piles at

the same locations. This is because the rigidity of a pile group provides more resistance

to the tunnelling-induced soil movements.


287

The scenario becomes more complicated when different pile cap conditions

were modeled, as the head conditions played a vital role in dictating the pile responses.

Generally, capped-head piles (Test PG2 (2-pile group) and Test PG4 (6-pile group))

demonstrate significant pile-cap-pile interaction among the piles. On the other hand,

the fixed-head piles (Test PG3 (2-pile group) and Test PG5 (6-pile group)), behaved

like single piles standing side by side without direct pile-cap-pile interaction, except

that the magnitude was affected by the total number of piles because the behaviors was

largely governed by the pile-soil-pile interaction.

When the pile group size increased from 2-pile to 6-pile, the position of the pile

within a group demonstrated a totally different transfer mechanism in lateral pile

responses, as compared to the single pile where the responses reduced consistently

when the distance of pile-tunnel increases. The pile-cap-pile interaction in capped-head

6-pile group (Test PG4) would moderate the induced pile bending moments among the

piles within a pile group. As a result, the induced pile bending moments in the middle

row was smaller than that of rear row. This is contrary to the induced lateral soil

movements, in which the corresponding lateral soil movement at the middle row of

piles was larger than the corresponding lateral soil movement at the rear row of pile.

This suggests that part of the bending moments of the pile in the middle row was

transferred to the rear piles due to the interaction through the pile cap. It is worth

noting that the axial forces reduce when the position of the pile-tunnel increases. Thus

the pile-cap-pile interaction in capped-head 6-pile group (Test PG4) has less influence

on the axial force as compared to bending moment and the induced pile axial forces

were mainly influenced by the soil settlement and distance effects.


288

On the other hand, the piles in fixed-head 6-pile group (Test PG5) behaved like

single piles in term of axial force and bending moment, except that the magnitude is

affected by the total number of piles. In contrast, there was an increment in the pile

bending moment due to the totally fixed-head condition, which restrained the pile

movement that resulted in high bending moment.

7.2 RECOMMENDATIONS FOR FUTURE STUDIES

The findings from the centrifuge experiments in the present study provided the basis

for the understanding of tunnel-soil-pile interaction. Some possible areas that could be

explored further are discussed here:

• In the present study, the smallest volume loss that was modeled in the

centrifuge test was 3%. However, with recent advancement in tunnelling

technology, the volume loss can be controlled to less than 1%. Hence, an

improvement to the current model tunnel to a smaller volume loss is

recommended.

• Future work is needed to study three-dimensional tunnel excavation in order to

study the longitudinal effects of tunnelling. To achieve this, modification of the

present two-dimensional model tunnel is needed. It is also recommended that

mechanical model tunnel with several small segments be developed to simulate

three-dimensional tunnel excavation.


289

• It is proposed that numerical analysis could be performed to validate the

centrifuge experimental results and parametric studies could be carried out to

improve the understanding of various effects of tunnelling.

• The effects of soil strength on tunnelling-induced soil movement and pile

responses could be further explored. In the present study, normally

consolidated clay used was relatively soft, and it would be interesting to study

the responses of stiffer clays.

References

290

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