INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE

UNIVERSITY OF BRADFORD

Faculty of Engineering & Informatics

School of Engineering

GEOTECHNICAL ENGINEERING: INGENUITY IN SOIL

REINFORCEMENT METHODS AGAINST SCOUR FAILURE

Edison Derrick Mugoya

12006149

BEng (Honours) Civil & Structural Engineering

Project Supervisor: Dr MHA Mohamed

Stage 3 BEng Final Year Project

April 2015

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Edison Derrick Mugoya Final Year Project

INGENUITY IN SOIL REINFORCEMENT METHODS

AGAINST SCOUR FAILURE

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PROJECT AIM

To study the behaviour of various soils to enhance the stability and strength of abutments using soil

reinforcement methodologies.

PROJECT OBJECTIVES

To encourage the use of discrete natural fibres methods to enhance the slope stability on

embankments.

To investigate the use of Geosynthetics materials to increase the cohesion between soil types as an

attempt to increase the overall strength of the soil.

To use soil reinforcement techniques to mitigate scour failure around bridge abutments.

To exploit the flow around an abutment to understand different scour conditions.

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DECLARATION OF INDEPENDENT WORK

Student Name: EDISON DERRICK MUGOYA

UoB Number: 12006149

Course: BEng (Honours) Civil & Structural Engineering

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Table of Contents

INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE ................... i

PROJECT AIM ....................................................................................................................................... ii

PROJECT OBJECTIVES ....................................................................................................................... ii

DECLARATION OF INDEPENDENT WORK ................................................................................... iii

Acknowledgements ................................................................................................................................ ix

Chapter 1 Introduction .......................................................................................................................... 10

Chapter 2 Literature Review ................................................................................................................. 11

Soil Characteristics and Properties ................................................................................................... 11

Soil Classification ......................................................................................................................... 11

Sedimentation ............................................................................................................................... 12

The determination of soil moisture ............................................................................................... 15

A New View of Abutment Scour ...................................................................................................... 17

Stress Conditions for Failure ............................................................................................................ 17

Effective and neutral stresses ........................................................................................................ 21

Scour conditions................................................................................................................................ 25

Shear Strength ................................................................................................................................... 29

Stability of Earth Slopes ................................................................................................................... 35

Clay Slopes ................................................................................................................................... 35

Stability analysis of the infinite slope ........................................................................................... 37

Consolidation Theory ........................................................................................................................ 45

Soil Compaction................................................................................................................................ 46

Factors Affecting the Compaction Process ................................................................................... 47

Different Compaction Methods .................................................................................................... 47

Chapter 3 Case Study ............................................................................................................................ 49

Case study: New Zealand bridge scour experiences ......................................................................... 49

Bulls Road Bridge ......................................................................................................................... 49

Qualitative Analysis of Expected Scour Development ................................................................. 51

Definitions ..................................................................................................................................... 52

Motivation for review ................................................................................................................... 52

Abutment form and construction .................................................................................................. 56

Abutment form .............................................................................................................................. 56

Abutment layout ............................................................................................................................ 57

Abutment construction .................................................................................................................. 57

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Pier proximity ............................................................................................................................... 59

Sediment and soil boundary material ............................................................................................ 60

Flow field ...................................................................................................................................... 61

Design scour depths ...................................................................................................................... 64

Estimation of scour depths ............................................................................................................ 64

An Essential design question ........................................................................................................ 65

Influence of pier proximity ............................................................................................................... 67

Other scour processes.................................................................................................................... 67

Chapter 4 Methodology ........................................................................................................................ 68

Embankments Design and Application ............................................................................................. 68

Soil Reinforcement Techniques ........................................................................................................ 69

Soil Reinforcement by fibre materials .......................................................................................... 69

The main factors that make Natural Geotextiles useful ................................................................ 69

Jute Fibres ..................................................................................................................................... 70

Coir Geotextiles ............................................................................................................................ 71

Bamboo and Timber Fibres........................................................................................................... 71

Combination of Geotextiles .......................................................................................................... 72

Soil Reinforcement by Geosynthetics ........................................................................................... 72

Chapter 5 Implementation of Soil Reinforcement ................................................................................ 73

DIRECT SHEAR TEST ................................................................................................................... 73

What is this test about ................................................................................................................... 73

Purpose:......................................................................................................................................... 74

Standard Reference: ...................................................................................................................... 74

Apparatus: ..................................................................................................................................... 74

Equipment: ........................................................................................................................................ 75

Test Procedure .............................................................................................................................. 77

Chapter 6 Results & Analysis ............................................................................................................... 78

Chapter 7 Conclusions and Recommendations ..................................................................................... 85

References ............................................................................................................................................. 86

Books ................................................................................................................................................ 86

Journals ............................................................................................................................................. 86

Websites ............................................................................................................................................ 87

Appendix ............................................................................................................................................... 89

Appendix A: Project Management and Organisation ....................................................................... 89

Appendix B: Plan of Action .............................................................................................................. 90

Appendix C: Gantts Chart ................................................................................................................. 91

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Appendix D: Mind Map .................................................................................................................... 93

Appendix E: ...................................................................................................................................... 94

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Table 1: Generally accepted soil classification (Online, Reference) .................................................... 12

Table 2: Calculation of the coefficient of uniformity ........................................................................... 14

Table 3: Fellenius’s Construction for centre of Rotation ...................................................................... 36

Table 4: Maximum Dilation angle for all densities, stress and fibre conditions ................................... 83

Table 5: Angle of friction and Cohesion intercept for all series of tests ............................................... 83

Figure 1: Soil Classification Chart (Online, Reference www.nrcs.usda.gov) ....................................... 13

Figure 2: Grain-size curves (Online, Reference www.fao.org) ............................................................ 14

Figure 3: The Neutron Moisture Meter; modern and conceptual images (Online, Reference

www.usyd.edu.au) ................................................................................................................................ 16

Figure 4: Apparatus used to demonstrate difference between effective and neutral stress

(http://www.iitbhu.ac.in/ internet reference) ......................................................................................... 21

Figure 5: Effective stress when there is no water flow (http://www.iitbhu.ac.in/ internet reference) .. 24

Figure 6: A butment-scour conditions: Scour Condition A - hydraulic scour of the main channel bed

causes bank failure, which causes a failure of the face of the abutment embankment (a); Scour

Condition B - hydraulic scour of the floodplain causes failure of the .................................................. 26

Figure 7: Field example of Scour Condition A ..................................................................................... 28

Figure 8: Field example of Scour Condition B. .................................................................................... 28

Figure 9: Field example of Scour Condition C for a wing-wall abutment. ........................................... 29

Figure 10a: Stress conditions at failure. ................................................................................................ 30

Figure 11: Stress conditions including envelope .................................................................................. 30

Figure 12: Particle in contact causing locking ...................................................................................... 31

Figure 13: example of shear failure in soils .......................................................................................... 31

Figure 14: Shear Box apparatus for test on shear strength .................................................................... 32

Figure 15: Triaxial test apparatus for test on shear strength ................................................................. 33

Figure 16: Stability of clay slope .......................................................................................................... 35

Figure 17: Method of Slices for clay Slopes ......................................................................................... 37

Figure 18: Failure condition for an infinite slope of cohesionless soil ................................................. 38

Figure 19: Failure condition of an infinite slope of cohesive soil ......................................................... 40

Figure 20: Slip circle: Cohesive soil ..................................................................................................... 41

Figure 21 ............................................................................................................................................... 42

Figure 22: Friction circle method.......................................................................................................... 44

Figure 23: Optimum Moisture Content graph ....................................................................................... 47

Figure 24: Schematic of long, multi-span bridge over a compound channel. ....................................... 53

Figure 25: Schematic of relatively short bridge over a narrow main channel ...................................... 53

Figure 26: Abutment scour resulting in embankment failure by collapse due to geotechnical

instability. ............................................................................................................................................. 54

Figure 27: Scour at I-70 bridge over Missouri River from 1993 flood. Flow was from left to right.

(Photo from Parola et al. 1998). ............................................................................................................ 55

Figure 28: Plan views of the two common abutment forms: (a) Wing-wall; (b) Spill-through (Ettema

et al. 2010). ........................................................................................................................................... 56

Figure 29: Definitions of embankment length, floodplain width, and main channel width (Ettema et al.

2010) ..................................................................................................................................................... 57

Figure 30: an isometric view of the geometry used for spill-through abutments. ................................ 58

Figure 31: The geometry and dimensions of a standard-stub abutment commonly used for spill-

through abutments (prototype scale indicated); design provided by the Iowa DOT (Ettema et al. 2010)

.............................................................................................................................................................. 58

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Figure 32: The geometry and dimensions of a wing-wall abutment - compacted earth fill embankment

extends back from the abutment structure (prototype scale indicated); design provided by the Iowa

DOT (Ettema et al. 2010). ..................................................................................................................... 59

Figure 33: depicts an example of a bridge with a pier located close to an abutment. ........................... 60

Figure 34: Variation of soil and sediment types at a bridge crossing (Ettema et al. 2010)................... 60

Figure 35: Flow structure including macro-turbulence generated by flow around abutments in a

narrow main channel. (Ettema et al. 2010). .......................................................................................... 61

Figure 36: Flow structure including macro-turbulence generated by floodplain/main channel flow

interaction, flow separation around abutment, and wake region on the floodplain of a compound

channel. (Ettema et al. 2010). ............................................................................................................... 62

Figure 37: For a spill-through abutment well set back on a flood-plain, deepest scour usually occurs

where flow is most contracted through the bridge waterway. .............................................................. 63

Figure 38: Interaction of flow features causing scour and erodibility of boundary (Ettema et al. 2010).

.............................................................................................................................................................. 63

Figure 39:: A common situation of abutment failure; scour has led to failure and partial washout of the

earth fill spill-slope at this abutment. A basic question arises as to how abutment design should take

scour into account. ................................................................................................................................ 66

Figure 40: Failure of abutment fill in September 2009 Georgia flood accompanied by failure of

approach roadway (Hong and Sturm 2010). ......................................................................................... 66

Figure 41: Typical Earth Dike with Drain ............................................................................................ 68

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Acknowledgements

The subject matter of this project appealed to me as I was studying geotechnical engineering

in the University of Bradford 2012-2015 and in response I favoured the idea of reading articles and

journals on the related topic of soil mechanics as an effort to enhance the structures for a sustainability

feasibility study on Whitby’s upper harbour endeavour wharf centre. I thought that I could spot the

house that had sunk slightly due to the geology and not as a cause of seismic activity. Many sections

of this project consists of the accumulations of study’s done in the field of soil reinforcement and

I.C.E and I.Struct have been of major assistance in the up to date goings of the new findings in

methods to improve the technique of strengthening soil mass. The author to whom I acknowledge

most of the study from is Karl Terzachi from his book theoretical soil mechanics tenth print 1943 and

well referenced to Harvard University in appreciation of its liberal encouragement of the pursuit of

knowledge, this book was gratefully dedicated in assembling the acquired understanding of

Geosynthetics toward soil reinforcements.

Journals such as shear behaviour of a geogrids-reinforced coarse-grained soil based on

Reynolds, h.r. and protopapadakis, p heavily assisted me in the make-up of my literature review. Also

the laboratory evaluation of governing mechanism of frictionally resistance was assisted by Craig, r.f

soil mechanics based of the degree of problems in soil mechanics and foundation engineering

problems picked up from Menzies and Simon’s journal. The core sections of experimental and

numerical analysis of large scale pull out tests also found in chapter 4 methodology was

acknowledged mainly by babu sivakumar, g.l. study. The implementation section was an extract from

the journal on soil bioengineering/biotechnical stabilization of slope failure written by engineer, haley

& Aldrich, Inc., Cambridge, Massachusetts and was acknowledged by principal, robbin b. sotir &

associates Inc., Marietta, Georgia.

I would also like to acknowledge my supervisor Dr. M.H.A. Mohamed for his continuous

efforts in guiding my written work evidence of our meetings can be found in the logbook attached at

the appendix shows the proof of the meetings and minutes. Being the head of discipline at the faculty

of engineering & informatics at the university of Bradford he has advised me and guided me through

the structure of the project.

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Chapter 1 Introduction

The concept of soil reinforcement is an ancient technique and is demonstrated abundantly in

nature by animals, birds and tree roots. It is an attempt to improve the stability of the soil such as the use

of plants and their roots to simply hold the ground together to prevent the top soil from sliding.

Constructions using these techniques are known to have existed in the fourth and fifth millennia B.C.

(Jones, 1985). This concept has been favourable for many centuries and has been used for improvement

of certain desired properties of soil such as bearing capacity (𝑞𝑢), shear strength (φ) and permeability to

name a few. The idea in principle was first developed by Vidal (1969), by inducing an element into soils

as a means of increasing its shear resistance to lateral loads. According to Vidal’s idea the interaction

between soils and the reinforcing specimen was to increase its horizontal resistance due to friction

generated by gravity, in this case the unit weight of soil. The first retaining wall built in France 1968 was

based off Vidal’s idea. The technology became wide spread through to USA and Europe but wasn’t

picked upon by Asian countries such as India due to the high cost and availability of the reinforcing

materials. More recently, the soil reinforcement techniques is well established and is used in a number of

applications such as to control drainage, improve soil bearing capacity and prolonging effects of scour

failure.

For a vast majority of reinforced soil structures the main reinforcement force is axial tension.

Earth Structures by themselves are weak, apply a tensile force on the earth structure and it will fail with

minimal effort. The horizontal forces acting on a soil mass can provide enough force to overcome the

resultant forces over the obliquity for the given soil type. However inserting tensile reinforcements in the

horizontal plain will enable vertical faced masses of soil to remain stable. (Jones, 1985) stated that the

mode of action of reinforcement in soils is not through carrying developed tensile stresses, as in

reinforced concrete, but rather of anisotropic reduction of the normal strain rate. A wide range of

materials since (Jones, 1985) have been used and tested as reinforcing materials. Early structures used to

be formed using organic materials such as timber, straw, reed, bamboo and sisal for reinforcements also

known today as fibre materials. Although they are generally considered as less durable they meet the

functionality of its design and are cheap. New materials such as steel, concrete, glass fibre, rubber,

aluminium, and thermo-plastics have been used more successfully due to their material properties.

One of my objectives to this study is to review the present state of knowledge regarding bridge-

abutment scour and evaluate the leading methods currently used for estimating design scour depth. This

study should ease my dissertation in formulating an improved approach to the fight against scour failure.

It focuses on research information obtained since 1990, and that must be considered in updating the

scour-depth estimation methods recommended by AASHTO1, and used generally by engineering

practitioners. Reinforcing soils provides a relatively cheap form of construction for retaining walls,

bridges abutments, marine structures, reinforced slopes and embankments mainly due to the ease and

speed of construction. These methods are also considered to be more aesthetically pleasing to the eye as

they retain the natural beauty of the landscapes whistle using up minimal land.

The main aim is to study the behaviour of various soils such as clay and sands to find new ideas

and methods to enhance the overall slope stability around embankments and to strengthen abutments

using soil reinforcement methodologies. In this study I will also exploit the shear strength, tensile strength

and bearing capacity behaviour of various soils to draw to a conclusion on a more versatile, reliable, cost

effective method to reduce geotechnical related accidents such as landslides that are a result of

overloading or poor engineering techniques.

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Chapter 2 Literature Review

This section covers the background knowledge that is necessary to draw any solution to scour

failure by looking at the fundamentals of soil mechanics. The complexity of bridge abutment scour

necessitates a thorough evaluation of the physical processes involved and their parameterization in scour

depth estimation formulas. As river flow approaches a bridge, the streamlines converge due to the

physical contraction in width and then diverge once through it. Understanding the soil properties and

characteristics prior to construction of bridge abutment could prevent scour failure

Soil Characteristics and Properties

Generally Speaking classifying soil characteristics and its properties in the field of geotechnical

engineering has not been adequately appreciated. Henry R. Reynolds and P. Protopapadakis mentions that

is essential to become familiar with the results of soil tests and to grasp the significance of its properties

and behaviours through experience. Soil moisture content, plastic limit and liquid limit are just some of

the basic fundamentals towards soil reinforcement. Soil reinforcement I believe is a study that will

enhance and change modern living styles. As the push for sustainability is crucial this increases the

demand for structures that are able to be stable and constructed in a manner that is suitable to the

environment. Engineering practices are faced with challenges ; whether a certain soil is suitable for

tipping in an embankment of a specified height, whether the load of the structure is capable of being

supported without undue or even settlement, whether groundwater lowering or artificial cementations for

deep excavation is possible, and other similar foundation problems.

The next few sections is allocated to examples involving soil classification and soil properties

which will in turn provide the basis for the practical engineering problems contained in the later chapters.

Soil Classification

Soil consists of a mixture of mineral practical’s and water, and includes a wide range of materials

from shingle to plastic clay. Soil is any uncemented or weakly cemented accumulation of minerals

particles formed by the weathering of rocks. Most soil can be easily excavated by hand or hand tools. In

the study of soil mechanics it is most important to be able to classify the different soils into defined types

based on their size, shape and nature of the particles. However it also must be recognized that the nature

of the soil particles is largely dependent on its moisture content. The classification tests are two types:-

(1) Mechanical analysis, by means of sieving or sedimentation to determine the size-distribution

of the soil particles.

(2) Index tests, for the soils passing a 36-mesh B.S. sieve, by means of which the type of soil is

deduced from moisture content at standard consistencies.

The generally accepted term and standards are given in the table 1 and table 2 together with the

chart in figure 1.

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Table 1: Generally accepted soil classification (Online, Reference)

Sedimentation

The grains of soil settle in a liquid with a velocity which may be calculated by Stoke’s Law,

which states that the rate at which small sphere sinks in liquid is directly proportion to the square of the

diameter of the sphere (Reynolds & Protopapadakis, 1959). This law applies only when considering grain

diameters between 0.2mm and 0.0002mm. Grains larger than 0.2mm diameter settle with a varying

velocity and particles less than 0.0002mm diameter are in colloidal suspension. Velocity of settling in

cms.per sec.

Equation 1: Stoke’s Law

𝑣 =2(𝜌 − 𝜌𝑤)

9𝑛(

𝑑

2)

2

Where 𝜌 denotes specific gravity of the soil grains, 𝜌𝑤 denotes specific gravity of the liquid, N denotes coefficient of viscosity for the liquid, (0.000103 kg.sec.per.sq.m for water at 20°) D denotes the diameter of the soil grains

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Figure 1: Soil Classification Chart (Online, Reference www.nrcs.usda.gov)

The simplified formula for the spherical particles descending in still water is as follows:-

Equation 2: Stoke's Law for Spherical Particles

𝑣 = 8800𝑑² 𝑜𝑟 𝑑 =√𝑣

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The above equations can used to solve the following problems. Please refer to the appendix for the

solutions to the problems below.

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1. How long would it take for a particle of soil 0.01mm in diameter to settle from the surface to the

bottom of a pond 10ft. if the specific gravity of the water is 1.0 and of the soil is 2.55, and the

coefficient of viscosity of water is 0.1025g.sec.per sq.m? (ans:0.0088cm.per sec using the above

equation)

Figure 2: Grain-size curves (Online, Reference www.fao.org)

Figure 2 above shows the grain-size vs percentage of fineness by weight. The coefficient of uniformity is

the ratio of the particle size for 60% finer by weight to the effective diameter. From inspection of figure 2

the following table can be drawn up:-

Table 2: Calculation of the coefficient of uniformity

Sample Effective diameter for

10% finer

Diameter for 60%

finer

Coefficient of

uniformity

A 0.007mm 0.055mm 0.055

0.007= 7.86

B 0.096mm 0.146mm 0.00.146

0.096= 1.52

C 0.056mm 0.63mm 0.63

0.056= 11.25

Soil Properties; Voids ratio, porosity, moisture content and density.

A Soil is made up of soil particles with voids between the particles filled with either moisture or

air, or both. The Natural moisture content of soil is determined by weighing a sample before and after

drying at 105°C. The loss in weight is expressed as a percentage of the dry weight. When soil is saturated,

the moisture content (𝑤𝑐), voids ratio (e), porosity (n).

To determine the specific gravity of soil particles, a pycnometer bottle is used. This consists of

flask which has a volume of 500c.c. at a certain temperature, usually 20°C, and this volume is marked on

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the neck of the bottle. A soil sample of 25 to 50 grs. Is placed in the bottle, which is then filled up with

distilled water. The liquid is boiled to expel the air the air adhering to the soil particles, and when cool the

bottle is filled up to the mark on the neck and weighed.

The specific gravity of the soil grains,

Equation 3: Specific Gravity of Soil

=𝑤𝑒𝑖𝑔ℎ𝑡 𝑜𝑓 𝑠𝑜𝑖𝑙 𝑖𝑛 𝑡ℎ𝑒 𝑏𝑜𝑡𝑡𝑙𝑒

{𝑤𝑔𝑡. 𝑜𝑓 𝑑𝑟𝑦 𝑠𝑜𝑖𝑙 𝑖𝑛 𝑡ℎ𝑒 𝑏𝑜𝑡𝑡𝑙𝑒} + {𝑤𝑔𝑡. 𝑜𝑓 𝑏𝑜𝑡𝑡𝑙𝑒 + 𝑤𝑎𝑡𝑒𝑟} + {𝑤𝑔𝑡. 𝑜𝑓 𝑏𝑜𝑡𝑡𝑙𝑒 + 𝑠𝑎𝑚𝑝𝑙𝑒 + 𝑤𝑎𝑡𝑒𝑟}

The determination of soil moisture

(Reynolds & Protopapadakis, 1959) had research by Saskatchewan research council and prairie

Road builders section of Canadian construction association based from the determination of soil moisture

and density by nuclear radiation that gave rise to the neutron moisture meter used today to determine soil

moisture on site. Paper was presented to the American society of testing materials in Cleveland, Ohio by

Messrs. D. A. Lane, B.B. Torchinsky and J.W.T. Spinks on the development of the technology.

The moisture meter is based on the fast neutron bombardment of soil, which results in the

reflection of slow neutrons and their action on a detector of indium foil. The apparatus consists of a cone-

shaped head which is lowered into a 2-in. diameter vertical tube sunk into soil. Attached to this head is a

capped aluminium cylinder about 11’’ In length. A neutron source is placed in the position by means of

electro-magnets within the cylinder and holder contain indium foil is lowered into its seating by means of

cord, thus positioning the indium foil around the neutron source. (Figure 3).

The fast neutrons emitted by the neutron source lose their energy by collision with the hydrogen

atom in the water molecules in the soil and therefore, become slow neutrons which are absorbed by the

stable indium foil causing a change to radioactive indium. The slow neutrons are more readily absorbed

than fast neutrons and the activity induced in the foil during a given period depends upon the activity

induced in the foil during a given period now depends upon the amount of moisture in the soil. An

equation has been evolved which relates the amount of activity of the soil to the soil moisture content.

The foil exposed for a definite period, withdrawn in its holder and placed round the Geiger tube of a

portable meter, and a reading is taken at the termination of the further definite period of time. However in

practise a specific procedure of a ten-minute exposure with a further one-minute delay before

measurement has been chosen as a standard routine. An initial reading of the foil should be taken before it

is used and additional five or six identical foils are required for the use, so that a reasonable period of time

can elapse before their re-use. Generally speaking for practical purposes, the zone of influence can be

considered to be a sphere of six inch radius.

The determination of soil density is based on the amount of gamma-ray absorption. The density

meter relies on the amount of gamma-ray absorption of soil which is proportional to its density. The same

neutron source is utilised as before in conjunction with the Geiger tube of the portable meter, but the

cylinder is larger, being about 20ins. Long with a lead shield place between it and the neutron source as

protection to direct gamma radiation. Thus the only rays which can strike the tube are those through the

soil. The tube is connected to the portable rate meter which indicates the average rate at which gamma

rays reach the tube and this value can be calibrated against wet density. It is essential, however that the

relative positions of the source and Geiger tube are constant throughout the test in order to produce

accurate and comparable calibration curves. A number of field experiments have been carried out using

16


aluminium tubes and steel pipes, and it is only necessary to adjust the calibration curves according to the

material used.

These results were checked with oven-drying methods and all determinations for moisture content agreed

within at least 3%, although most of the results were within 2%. Tests from the ground that is flooded

gave similarly effective results, but where soils have organic composition some variation does occur. The

presence of organic material affects the meter in the same way as if water were present and a correction

factor must be introduced dependent upon the soil type and further experimental work is taking place in

this field. Another possible use for neutron meter is to determine the asphaltic content of bituminous

pavement composition. The activation of the meter is similar to that for determining the water content and

compensating calibration curve can be developed to indicate the asphaltic percentage present. The use of

neutron and density meters necessitates certain precautions being taken by the operator. An electromagnet

mounted on a 3-ft. length of 5

16𝑖𝑛𝑐ℎ diameter rod projecting from an ordinary flashlight case facilitates

the safe handling of the 75Ib. When the neutron source is placed in position the operator must not stand

directly over it and must carry a film monitor indicating the exposure to gamma rays, so that he may keep

within the allowable margin.

Figure 3: The Neutron Moisture Meter; modern and conceptual images (Online, Reference www.usyd.edu.au)

17


A New View of Abutment Scour

Based on the foregoing definition of scour and documentation in this report of numerous failures

of bridges due to abutment scour, one of the important initial findings is that many abutment failures

occur due to scour and sliding of the earth fill embankment on the main stream side of the abutment into

the scour hole, or outflanking due to erosion of the earth fill embankment on the floodplain side due to

overtopping or inadequate drainage protection. Even more difficult to evaluate is the vulnerability to

scour caused by lateral shifting of the channel thalweg such that it directs flow adversely towards

abutments and embankments. Whereas much of the laboratory research of recent years has focused on

solid abutments that extend into the soil foundation, such as with sheet piles or other fairly rigid

foundations, more attention should be focused in the future on erodible embankments. Recognition of the

difference between erodible and solid abutments provides a factor for classifying existing scour prediction

formulas and introduces the importance of geotechnical failure caused by hydraulic scour. In addition, it

suggests the need for estimating the strength of the embankment over the range of construction forms

varying from unprotected, compacted soils of various types through rock riprap revetment to the solid

abutment, and incorporating this estimate into a more comprehensive scour prediction formula. These

considerations pose a fundamental design problem in that partial failure of the embankment that occurs as

sliding of earth fill and/or riprap into the scour hole may ultimately reduce the total scour depth while

complete failure of the embankment may be intolerable if it results in failure of the bridge approach slab

or the first bridge span. This leads on to the importance of understanding the stress condition for failure of

various soil types.

Stress Conditions for Failure

Relation between normal stress and shearing resistance makes up the better part of this section.

Stress will be referred to as force per unit of area of the section of soil mass. It has been genuine to

assume that the relation between normal stress (𝜎) on every section through a mass of cohesive soil and

the corresponding shearing resistance (s) per unit area can be represented by an empirical equation.

Equation 4: Coulomb's empirical equation

𝑠 = 𝑐 + tan 𝜙

Provided (𝜎) is compressive stress. The symbol (c) represents the cohesion, which equals to the

bearing shearing resistance per unit area if ( 𝜎) = 0. The equation is known as coulomb’s equation. For

cohesion-less soils (c = 0) the corresponding equation is

Equation 5: Coulomb's equation

𝑠 = 𝜎 tan 𝜙

The Values c and 𝜙 contained in the preceding equations can be determined by means of

laboratory test, by measuring the shearing resistance on plane sections through the soil at different values

of the normal stress 𝜎. In practise we are quite interested in the shearing resistance of saturated or almost

saturated soils. A change of stress in a saturated soil is always associated with some change of water

content. The rate of the change of water content produced by a given change of state of stress depends on

several factors, including the degree of permeability of the soil. If the stresses which ultimately lead to

failure of the test specimen are applied more rapidly than the corresponding change in the water content

of the specimen can occur then parts of the applied normal stress 𝜎 will be carried at the instant of failure,

by excess hydrostatic pressure which is required to maintain the flow of the excess water out of the voids

18


of the soil. At a given value of 𝜎 the part of 𝜎 which is carried by both the value c and φ depend not only

on the nature of soil and its initial state but also on the rate of stress application, on the permeability of the

material, and on the size of the specimen. The value of 𝜙 obtained from such tests is called the angle of

shearing resistance. For clays this angle can have any value up to 20° (exceptionally more) and for loose,

saturated sands any value up to 35°. In other words, no definite value can be assigned to the angle φ for

any soil, because it depends on conditions other than the nature and initial state of soil.

On the other hand if the stresses on the test specimen are applied slowly enough, the normal stress

𝜎 which acts on the surface of the slider at the instant of failure is almost entirely transmitted from grain

to grain. Tests of this kind are known as slow shear tests. The rate at which such tests must be made

depends on the permeability of the soil. If shear tests on sand with a given initial density are made in such

a manner that the stresses are entirely transmitted from grain to grain, we find that the shearing resistance

s = 𝜎 tan 𝜙 is practically independent of the character of the changes of stress which preceded the failure.

For instance, it makes practically no difference whether we increase the unit load on the sample

continuously from 0-1 ton per square foot and then reduce it to 1 ton per square foot. If the load on the

sample at the instant of failure is equal to 1 ton per square foot, the resistance s is the same in both cases.

In other words, the shearing resistance s depends solely on the normal stress on the potential surface of

sliding. A shearing resistance of this type is called a frictional resistance and the corresponding value of

φ represents an angle of internal friction. Within the range of pressure involved in engineering problems

the angle of internal friction of sand can usually be considered constant for practical purposes. Its value

depends on the nature and initial density of the sand. It varies between the extreme limits of about 30° and

50°. The difference between the angle of internal friction of given sand in the densest and in loosest state

may be as high as 15°.

In (Terzaghi, 1962) investigations of soil problems he generally assumed that the angle of internal

friction of sand is identical with the angle of repose described in Article 3. However, as stated above,

laboratory experiments have shown that the angle of internal friction of sand depends on a large extent on

the initial density. In contrast to the angle of internal friction, the angle of repose of dry sand has fairly

constant values. It is always approximately equal to the angle of internal friction of the sand in the loosest

state. BS EN and text books also contain a list of values for the angle of repose of cohesive soils, although

as shown in Article 4, the angle of repose such soils depends on the height of the slope.

When equation 5 is used in connection with stability computations the value 𝜙 always represents

the angle of internal friction of sand. This can only be expressed with the sufficient accuracy by equation

4,

𝑠 = 𝑐 + 𝜎 𝑡𝑎𝑛𝜙

In order to find out whether the term 𝜎tan𝜙 satisfies the requirements for a frictional resistance

i.e. whether the resistance 𝜎tan𝜙 depends solely on the normal stress𝜎, we submit our material with a

given initial water content to two different tests.

In on test we increase 𝜎 from zero to 𝜎1 and determine the corresponding shearing resistance𝑠1.

In the second test, we consolidate our materials under a pressure of 𝜎2 which is greatly larger than 𝜎1 and

finally we determine by means of slow shear test, the corresponding shearing resistance 𝑠′1. The process

of temporarily keeping a sample under pressure which is higher than ultimate pressure is known as

preconsolidation. Experiment show that the shearing resistance 𝑠′1 of preconsolidated material may be

equal to if not greater than𝑠1. If the two values are equal 𝜎 𝑡𝑎𝑛𝜙 in equation 1 represents a frictional

resistance and we are justified considering 𝜙 an angle of internal friction. On the other hand if 𝑠′1 is

19


greater than 𝑠1, we know that the resistance 𝜎 𝑡𝑎𝑛𝜙 represents the sum of a frictional resistance and some

other resistance which is independent𝜎. The most conspicuous permanent change produced by

preconsolidation consists in an increase of the density of the material and a corresponding reduction of

water content. If 𝑠′1 is appreciably greater than 𝑠1 we always find the water content corresponding to 𝑠′1

is lower than that corresponding to 𝑠1. We know from experience that the value c in the equation 1

increases for a given clay with decreasing initial water content. Therefore in most cases we are justified in

drawing the following conclusion. If 𝑠′1 is appreciably greater than 𝑠1, the resistance 𝜎 𝑡𝑎𝑛𝜙 in the first

part is friction produeced by normal stress 𝜎 and the second part is the increase of the cohesion the

reduction of the water content which occurred while the pressure on the specimen was increased from

zero to 𝜎. (Terzaghi,1962) statement can be expressed by the equation below.

Equation 6: Terzaghi 1962

𝑠 = 𝑐 + 𝜎 𝑡𝑎𝑛𝜙 = 𝑐 +𝜎1 + 𝜎3

2𝑁 + 𝜎𝑡𝑎𝑛𝜙𝑓

Whereby 𝜎1 and 𝜎3 represent the extreme principal stresses at failure after the slow test and N is

an empirical factor. The fraction 𝜎𝑡𝑎𝑛𝜙𝑓 of the shearing resistance changes with orientation of a section

through a given point, while the fractions c and 𝜎1+𝜎3

2𝑁 are independent of the orientation. The customary

methods for experimentally investigating the shearing resistance of cohesion soils merely furnish the

values c and 𝜙 on the left-hand side of the equation. The determination of the values 𝜙𝑓 and N requires

elaborate supplementary investigations which belong in the realm of soil physics.

For cemented sand the value 𝑠′1 is usually very close to that of 𝑠1. For such materials the value

𝜎 𝑡𝑎𝑛𝜙 in equation 1 represents only a frictional resistance. On the other hand when experimenting with

clay we find that the shearing resistance 𝑠′1 of the preconsolidated sample is always appreciably greater

than 𝑠1 at the same load. Hence in connection with clay the angle of 𝜙 in equation 1 represents neither an

angle of internal friction nor a constant for the clay, even when its value has been determined by means of

slow shearing tests. If one makes a series of slow tests on clay with a given initial water content after

increasing the pressure on the samples from zero to different values𝜎1, 𝜎2 etc. one gets an equation


If one makes another series of tests on specimens of the same material after preceding

consolidation of the samples under a pressure which is higher than the test pressures one gets another

equation

𝑠 = 𝑐′ + 𝜎 𝑡𝑎𝑛𝜙′

Whereby c’ is considerably higher than c and 𝜙′ is considerably smaller than 𝜙. Hence when

using coulomb’s equation 1 in connection with clays, note that c and 𝜙 contained represent merely two

empirical coefficients in the equation of a straight line. The term cohesion is retained only for historical

reasons. It is used as an abbreviation of the term apparent cohesion. In contrast to the apparent cohesion,

the true cohesion represents the part of shearing resistance of soil which is a function only of the water

content. It includes not only c in coulomb’s equation but also an appreciable part of 𝜎 𝑡𝑎𝑛𝜙. There is no

relation between apparent and true cohesion other than the name.

20


To help you visualize the difference between apparent and real cohesion I’ll need to consider a

material that has increases it’s cohesion with compaction. Clay is a good generic example. By making a

series of shear tests with the material we obtain the following:-


However when we look at the parts of shearing resistance of the material that is due to cohesion

we obtain the equation below also derived from Mohr circle,

𝑠 = 𝑐 + 𝜎1 + 𝜎3

2𝑁 + 𝜎 𝑡𝑎𝑛𝜙𝑓

If you compared the two preceding equations we get that the true cohesion of the material is equal

to c but to:

21


𝑐𝑒 = 𝑐 + 𝜎1 + 𝜎3

2𝑁

If the entire pressure on the clay is transmitted from grain to grain the true cohesion is never

smaller than the apparent cohesion. Therefore if 𝜎 𝑡𝑎𝑛𝜙 from the equation 1 is equal to zero then

𝑠 = 𝑐

For liquids the values c and 𝜙 are zero which means that

𝑠 = 0

Effective and neutral stresses

In real life the voids of every fine-grained soil are partly or wholly filled with water. If we take a

section through a saturated soil, then parts of it will pass through the solid particles and part of it through

the water. In order to assure the mechanical implications of this fact, we need to consider the test

arrangement as shown below in figure 4. This apparatus represents a section through a layer of cohesion

less soil which occupies the bottom of a vessel. At the outset of the test the free water level is supposed to

be located immediately above the surface of the soil and the layer is assumed to be

So thin that we neglect the stress due to the

weight of the soil and water which are loacted

above the horizontal section known as the datum

the (ab). If we raise the water level to an elevation

of ℎ𝑤 above it’s orginal place the normal stress

expereince on the section (ab) increses from

almost zero to 𝜎 = ℎ𝑤𝛾𝑤 where 𝛾𝑤 is the unit

weight of water. Yet this may increse the

compressicve stress from practically zero to 𝜎 on

every horizontal section. In which the soil doesn’t

make a measurable compression of the layer of

soil. On the other hand if we increse the intenstiy

of the pressure on the layer by the same amount

the 𝜎, by loading the surface o fthe layer with

lead shot the resulting compression of the layer is

very appreciable. As seen practised in the field of

the Shanghai lotus river to date. This appartus can

also demonstrate that the postio of water level in

the vessel has no influence on the shearing

resistance s of the soil, whereas an equivalent

soild surchage increases the shearing resistance

a b

ℎ𝑤

Figure 4: Apparatus used to demonstrate

difference between effective and neutral stress

(http://www.iitbhu.ac.in/ internet reference)

22


very much. These and many experiments similar to it lead to the conclusion that the compressive stress in

saturated soil consists of two parts with very different mechanical effects. One of which equals to the

pressure in the water produces neither a measurable compression nor a measurable increase of the

shearing resistance. (Terzachi, 1962) called it neutral stress 𝑢𝑤 . He quotes that the product of the unit

weight of water 𝛾𝑤 and the height ℎ𝑤 to which the water rises in a piezometric tube at the point under

consideration, then the corresponding equation is

𝑢𝑤 = ℎ𝑤𝛾𝑤

The height as shown in figure4 represents the piezometric head at the point of observation. It can

be positive or negative. Hence how the neutral stress can be a positive or negative value. If it is positive it

is usually called the pore-water pressure.

Subsurface water is divided into zones of positive and negative pore pressures. The dividing line

is the groundwater table (also known as phreatic surface) where the pressure is equal to atmospheric

pressure. Below the groundwater table, the soil is fully saturated, and the pore pressure is above

atmospheric pressure and positive in value. Above the groundwater table where the soil is unsaturated, the

pore pressure is below atmospheric pressure and hence is negative in value. In this zone, the pore water is

continuous or semi continuous and the pore water pressure is below atmospheric pressure. The magnitude

of the negative pore pressure (sometimes called soil suction) is controlled by surface tension at the air-

water boundaries within the pores and is governed by grain size. In general, the finer the soil particles, the

larger the saturation capillary head, and hence the higher the negative pore pressure. Rainfall infiltration

from the ground surface may rapidly reduce the magnitude of negative pore pressure. Any change in these

pore pressures alters alter the shear strength of soil and therefore has a tremendous effect on the slope

stability.

The water level measured in a piezometer within the saturation zone coincides with the water

table. However, the pore pressures are no longer hydrostatic if there is a flow. In this instance, the pore

pressure from any point within the soil mass is computed by means of a flow net, from the difference in

head between the point and the free water surface.

By lowering effective stress, positive pore pressure reduces the available shear strength within the

soil mass thereby decreasing the slope stability. Increase in positive pore pressure can be rapid after a

period of heavy rainfall. That is a major reason why many slope failures occur after heavy rainfall. The

rate of increase, however, depends on many factors such as the rate of rainfall, the nature of the ground

surface, the catchment area, and the soil permeability. Pore pressure below the groundwater table can be

assessed using analytical, numerical, and graphical methods. Various analytical methods are available for

determining flow nets and pore pressure distributions in slopes. Numerical techniques using finite

difference or finite element method provide powerful tool for obtaining pore water distributions in slopes.

They are the only means by which transient flow situations can be fully modelled.

Negative pore pressures increase the effective stresses within a soil mass and improve the

stability of a slope. (Ho and Fredlund, 1982) suggested increase in shear strength due to negative pore

pressure as

23


Equation 7: Ho and Fredlund, 1982

𝑐 = 𝑐′ + (𝑢𝑎 − 𝑢𝑤 )𝑡𝑎𝑛𝜙𝑏

Where c = total cohesion of the soil

c’= effective cohesion

(𝑢𝑎 − 𝑢𝑤 ) = matrix suction

𝜙𝑏 = the slope of the plot of matrix suction when - is held constant

Here, a matrix suction (𝑢𝑎 − 𝑢𝑤 ) increases the shear strength by (𝑢𝑎 − 𝑢𝑤 )𝑡𝑎𝑛𝜙𝑏. The

increase in soil strength can be represented by a three-dimensional failure surface using stress

variable (𝜎 − 𝑢𝑎 𝑎𝑛𝑑 𝑢𝑎 − 𝑢𝑤 ), as shown in Figure. These negative pore pressures reduce in magnitude

when the degree of saturation increases and become zero when the soils are fully saturated; the major

problem in evaluation of stability in unsaturated soils is associated with the assessment of reduction in

negative pore pressure and possible increase in positive pore pressure as a function of rainfall history.

Let us consider full saturation of the rock, including the joint, where no drainage of water is

allowed. If we assume that water is incompressible and that no flow of water into or out of the joint is

allowed, the volume of the test specimen including the joint must remain constant. Under this condition,

the water must sustain stresses sufficient to prevent volume change of the specimen. The total applied

stress across the joint will be transmitted by the rock asperities and by the water. If the water carries some

of the normal stress, then the rock asperities carry less normal load and therefore has less shear strength

than it would be if drained. The normal stress transmitted by the water is equal to the joint water pressure.

The stress transmitted through the rock asperities is, therefore, equal to the applied stress minus the joint

water pressure. The joint shear strength will now be reduced proportionally. The reduced normal stress

acting through the rock contacts is termed as the effective normal stress and is given by

𝜎′𝑛 = 𝜎𝑛 − 𝑢

Where by

𝜎′𝑛 =effective normal stress 𝜎𝑛 = normal stress 𝑢 =water pressure

The total stress imposed on such a soil will be sustained by the soil; the effective stress, and the

pore pressure, u. A reservoir can be used to create an upward seepage through the soil sample. For this

purpose, we assume that the valve leading to the upper reservoir is closed. Thus, there is no water flowing

through the soil sample (figure 4). This is the case of no seepage the Effective stress is

𝜎′𝑛 = 𝐻2(𝛾𝑠𝑎𝑡 − 𝛾𝑤)

Upward seepage conditions can be induced in the laboratory using constant-head permeability

test apparatus, in figure 5. The upper reservoir causes the water to flow upward through the soil sample. If

the hydraulic gradient is large, the upward-seepage force will cause the effective stress within the soil to

24


become zero, thus causing a sudden loss of soil strength in accordance with the effective-stress principle.

However, if downward seepage is allowed, effective stress sigma' is

𝜎′𝑛 = 𝐻2(𝛾𝑠𝑎𝑡 − 𝛾𝑤)

Figure 5: Effective stress when there is no water flow (http://www.iitbhu.ac.in/ internet reference)

In the analysis of stability of slopes in terms of effective stresses, the pore water pressure

distribution is of fundamental importance and its evaluation is one of the prime objectives in the early

stages of any stability study.

25


Scour conditions

One method for classifying abutment scour depends on abutment location in a channel, the

relative erodibilities of sediments forming the main-channel bed and soils forming the floodplain (see

Figure 45), as well as to the shear strength of the compacted earth fill forming the approach embankment.

In addition, other conditions such as stream morphologic changes and lack of control of highway runoff

can lead to abutment scour under unexpected and less well-defined circumstances.

Three common conditions of abutment scour

Figure 49-c illustrates the three scour conditions for spill-through abutments:

1. Scour Condition A. Scour of the main-channel bed, when the channel bed is far more erodible than the

floodplain. Figure 49a illustrates how scour of the main-channel bed causes the main-channel bank to

become geotechnically unstable and collapse. The collapsing bank undercuts the abutment and

embankment, which in turn collapses locally. Soil, and possibly riprap, from the collapsed bank and

embankment slide into the scour hole;

2. Scour Condition B. Scour of the floodplain around the abutment. This condition also is equivalent to

scour at an abutment placed in a rectangular channel, if the abutment is set back from the main channel.

As the amount of bed-sediment transport on a floodplain usually is quite low, this scour condition usually

occurs as clear-water scour. Figure 49b shows that the floodplain scours around the abutment, and

especially slightly downstream of it. The scour hole locally destabilizes the embankment side slope,

causing embankment soil, and possibly riprap, to slide into the scour hole; and,

3. Scour Condition C. Scour Conditions A and B may eventually cause the approach embankment to

breach near the abutment, thereby fully exposing the abutment column. For this condition, scour at the

exposed stub column essentially progresses as if the abutment column were a pier, as illustrated in Figure

49c. For the same reasons as given for Condition B, this scour condition usually occurs as clear-water

scour.

26


Figure 6: A butment-scour conditions: Scour Condition A - hydraulic scour of the main channel bed causes bank failure, which

causes a failure of the face of the abutment embankment (a); Scour Condition B - hydraulic scour of the floodplain causes failure

of the

27


The three scour conditions may occur also for wing-wall abutments. However, a couple of additional

erosion processes can result in failure of the main-channel bank and the approach embankment:

1. The local flow field generated at the corners of the abutment can cause local scour at those locations;

and,

2. Exposure of the piles beneath the abutment pile cap can cause river-bank and embankment soil to be

eroded out from beneath the pile cap.

Provided no substantial geotechnical failure of the abutment occurs for scour Conditions A and B,

scour deepens to an equilibrium level commensurate with the abutment flow field’s capacity to attain a

balance with the rate of sediment inflow to the scour region (live-bed scour) or the channel boundary’s

resistance to erosion (clear-water scour).

A scour event (or series of events) at an abutment, may involve a sequence of all three scour

conditions, resulting in several local maxima for scour depth for a wing-wall abutment. When an

abutment is close to the main channel, Condition A may develop relatively quickly, with Condition B

occurring at a slower rate. Either, or together, Scour Conditions A and B may eventually cause the

approach embankment to undergo a slope-stability failure. If the embankment extensively washes out, so

as to expose the abutment structure, scour may then develop at the abutment structure as if the abutment

were a form of pier (Condition C). Accordingly, an important design consideration is that the stub or

wing-wall abutment should not fail when exposed; i.e. foundations of wing-walls should be deep enough

that the wing-walls do not fail when exposed to a pier-like scour condition.

For design estimation of scour depth, it is useful to consider the likely rates or sequences in which

the three scour conditions developed, and to ask -- What is the greatest scour depth that reasonably could

occur near the abutment? Will that scour depth pose a slope-stability problem for the earth fill

embankment adjoining an abutment foundation or for the floodplain bank of the main channel? What is

the deepest scour that could occur at the abutment column foundation itself, and does that scour occur

when the embankment is breached so as to fully expose the abutment column? The set of photographs in

Figures 50 through 51 depict situations where Scour Conditions A, B, and C occurred at bridge

abutments.

28


Figure 7: Field example of Scour Condition A

Figure 8: Field example of Scour Condition B.

29


Figure 9: Field example of Scour Condition C for a wing-wall abutment.

Shear Strength

Soil failure may occur as a result of exceeding the maximum shear stress that the soil can

handle/sustain. This is a major factor in understanding soil behaviour in evaluating its influence on soil

reinforcements. When it comes to the analysis of the stability of soil masses, whether at a point or on any

plane within a soil mass the shear stress becomes equal to the shear strength of the soil then failure will

occur at this point. Prior to the principles of effective stress, the shear strength 𝜏𝑓of soil at a point on a

particular plane was expressed by coulomb as a linear function of the normal stress at failure 𝜎𝑓 on the

plane at the same point.

Equation 8: Shear Strength

𝜏𝑓 = 𝑐 + 𝜎𝑓𝑡𝑎𝑛𝜙

Where c and 𝜙 are shear strength parameters referred to as the cohesion intercept and the angle of

shearing resistance, respectively. However, in accordance with the principle that governs shear stress with

in soil it can be resisted only by the skeleton of solid particles, shear strength should be expressed as a

function of effective normal stress a failure 𝜎′𝑓 the shear strength parameters being giving as c’ and 𝜙′.

𝜏𝑓 = 𝑐′ + 𝜎′𝑓𝑡𝑎𝑛𝜙′

Failure will thus occur at any point in the soil where a critical combination of shear stress and

effective normal stress develops. It should be expected that c’ and 𝜙′ are simply mathematical constants

defining a linear relationship between shear strength and effective normal stress. It easier to think of these

stress as inter-particle forces; therefore, if effective normal stress is zero then shearing resistance must be

zero (unless there is cementation between the particles) and the value of c’ would be zero. This brings me

to this point of crucial to the implementation of shear strength parameters.

In most 2 dimensional representations we plot shear stress 𝜏 against effective normal stress 𝜎′.

This is a state that can be represented either by a point with coordinates𝜏 𝑎𝑛𝑑 𝜎′, or by a Mohr circle

defined by the effective principles stresses with coordinates𝜎′1 𝑎𝑛𝑑 𝜎′3. As shown below the stress

points and Mohr circles represents stress states at failure (fig6a & fig7).

30


Figure 10a: Stress conditions at failure.

The line through the stress points that are touching the Mohr circle may be straight or slightly

curved and is referred to as the failure envelop. A state that means any stress points that are plotted above

this line is impossible as the soil would be in motion; either a twist or turning one. There two methods to

quantify the shearing strength parameters. (1) The envelope is represented by the straight line as defined

in the equation above. From which the parameters c’ and 𝜙′ can be obtained. These are referred to as

tangent parameters and are only valid over a limited range. It has been used for since Coulombs

discovered it. If the failure envelope is slightly curved the parameters are obtained from a straight line

Sh

ear

Str

ess

(MP

a)

Normal Stress (MPa)

Figure 11: Stress conditions including envelope

31


approximation to the curve over the stress range. It should be noted that the use of tangent parameters

does not reflect the real life scenario.

From this point I will be covering the shear strength of sand and the shear strength of clay as

found from reconstituted specimens in the laboratory. When sand particle arrangement is confined

laterally, strains can only in the vertical direction, such as in an oedometer test. As the vertical stress is

increased small groups of particles in the loose sand will collapse to the surrounding voids producing a

volumetric (or vertical) stain on the soil. This produces a more tightly packed arrangement with which a

larger number of particles are now in contact with each other. This makes it more fixed in place as it is

given less freedom to move about. This is a phenomenon known as locking. Thus the vertical stress-strain

curves are concaved. The term stiffness represents the gradient of this curve so it can be seen that the soil

is becoming stiffer. As we increase the vertical stress that is in contact between the particles, then the

particles will begin to facture and crush producing yield and allowing vertical strains to increase. An

increase in the number of particles (fig 8) due to crushing will produce a further particle contact which in

turn reduce the average contact stress between particles and causes the stiffness to continue increasing

due to further locking.

Figure 12: Particle in contact causing locking

The level at which structural collapse, locking and yielding occur depends on the initial overall

density and inherent strength or for a lack of a better word crushability of particles. Under confined

compression where horizontal strains are restricted, as the vertical stress increases the horizontal stress

increases. The horizontal stress does not need to be as large as the vertical stress this is so; because parts

of the latter will be supported by the shearing resistance of the sand. (Jaky, 1944) claimed that the

horizontal stress to vertical ratio denoted as 𝑘0 at rest, or lateral strain condition is as follows: 𝑘0 = 1 −

𝑠𝑖𝑛𝜙. Whereby 𝑘0 is the coefficient of earth’s pressure at rest.

Figure 13: example of shear failure in soils

32


The failure as shown in figure 9 above is due to inadequate strength at shear interfaces. Soil

derives most of its strengths from cohesion and frictional resistance. Cohesion is a measure of the forces

that cement the particles together. The above images could well be dry sand with no cementation, dry

sand with some cementation, soft clay or stiff clay. As mentioned above Coulombs law based on the

Mohr’s Failure criteria it suggests that the figure 9 suffered from a lack of good adhesives beneath the

road that would give an increased strength and protection against slides. Vertical drains are not visible

here so I think the water couldn’t escape easily loosening the soil by introducing a pressure difference

from the side to the other making an active pressure scenario 𝜎ℎ < 𝜎𝑣 . Such case means that the soil is

pushing outwards downslope, causing a rotational slip as seen in figure 9.

Shear box test and triaxial test are good measures of the shear strengths of soils. When a direct

measure of the shear strength of a granular soil is desired a shear box test is often used. Although the

results are from reconstituted data with densities and particle arrangements that are different from those

found in situ an allowance should be made for this. Nowadays the shear box test tends to be used for

investigation of shear strength properties of the more unusual granular materials where correlation

between 𝜙 and in situ tests such as SPT or CPT are not available or unreliable. These include crushable

sands like calcareous, vesicular sands, granular fills (fragmented rock particles both soft and hard, waste

materials (colliery spoil) and the shear strength of interfaces between two construction materials (steel

and sand, steel piles) or plastic and clay (geomembrane and clay liner). Other applications which have

utilised this test are for quick undrained strength of clay and cut-plane or reversal test (returning the split

specimen to its starting point) for the determination of the drained residual strength.

Figure 14: Shear Box apparatus for test on shear strength

33


Figure 15: Triaxial test apparatus for test on shear strength

The triaxial test apparatus was first developed in the 1930s and has larger been replaced by the

direct shear test in commercial laboratories. It consists of applying a shearing stress with a cylindrical

sample of soil by changing the principles stresses𝜎1 𝑎𝑛𝑑 𝜎3. The commonest procedure is to keep the

triaxial cell pressure 𝜎3 constant (𝜎2 = 𝜎3 with axial symmetry) and increasing the axial or vertical stress

𝜎1 until failure is achieved. The essential feature as shown in figure 11 below is standardized at 38mm

and 100mm diameter specimens. It has a height: diameter ratio of 2:1 to ensure that the middle section of

the specimen is free to shear. If this ratio is less than 2:1 then shear stresses at the end of the sample in

contact with the platens will affect the results by constraining the failure planes. The soil specimen is

surrounded by rubber membrane to stop the cell fluid from entering the soil and altering its moisture

content. For weaker soil specimens a correction to account for the restraint provided by the membrane

should be applied. The axial stress is applied by a motorised drive which raises the specimen and the cell

against the piston reacting on a load frame. A proving ring or load ring or load cell between the piston and

load frame measures the axial force F from which the principle stress difference or deviator stress,

𝜎1 − 𝜎3, is calculated using the equation below.

𝐹

𝐴= 𝜎1 − 𝜎3

The strength of the soil is obtained from Mohr circle plot. AS the stress is applied the specimen often

becomes barrel-shaped so the vertical stress in the middle of the specimen must be determined from the

force measured and this increased area by applying an area correction to each reading (Barnes, 2010).

The corrected area A in the middle of the specimen is obtained for a drained test from:

𝐴 = 𝐴0(1−𝜀𝑣)

(1−𝜀𝑎)

Where 𝐴0 is the initial cross-sectional area, 휀𝑎 is the vertical or axial strain and 휀𝑣 is the volumetric strain.

For an undrained test where the volumetric strain is zero (휀𝑣=0) the corrected area is given by

34


𝐴 = 𝐴0

(1−𝜀𝑎)

Now tying in to what I’ve mentioned above I conclude this section with the residual strength in

soil. Although the critical state strength is often referred to as the ultimate strength this condition is

achieved with homogenous shearing, i.e. all the samples are undergoing the same shear strain and these

strains are not excessively large. It has been seen particularly from the studies of the old landslips done by

(Skempton, 1964) where significant straining has occurred on thin shear surfaces that the operative shear

strength on these surfaces was much lower than the critical state strength. What I mean is the residual 𝜙𝑟′

value for London Clay for example can be as low as 10° whereas at the critical state 𝜙𝑐𝑣′ is greater than

20°. (Barnes, 2010) it’s essential therefore to identify the presence or otherwise, pre-existing slip surfaces

in clay soil on a sloping site. Small changes have been noted in the surface topography and also in the

pore pressure conditions from earlier sections of this project. Residual strength is then attained when large

shear strains have occurred on the thin zone or plane of sliding in clay soil. This formation is where the

clay particles have been arranged to produce a firm preferred orientation in the direction of the slip

surface. (Lupini, 1981) recognised three modes of residual shear behaviour;

Turbulent

This occurs where behaviour is dominated by rotund particles. For soils dominated by platy

particles with high inter-particle friction this mode may also occur. In this mode energy is

dissipated by particle rolling and translation. No preferred particle orientation occurs and residual

strength still remains high so that 𝜙𝑟′ can be taken as𝜙𝑐𝑣′.

Sliding

When behaviour is dominated by platy, low friction particles, sliding occurs on a shear surface

with strongly oriented particles and the strength is low. 𝜙𝑟′ depends mainly on the mineralogy

coeffeicent of inter-particle friction 𝜇 and pore water chemistry.

Transitional

This involves turbulent and sliding behaviour in different parts of a shear zone. The residual shear

strength can be obtained using a ring shear apparatus as (Bishop, 1971 and Bromhead, 1978)

found. A ring-shaped thin sample of remoulded soil is sheared in a direct shear manner by

rotating the upper half of the sample above the lower half with sufficient strain until a slip surface

is formed on which the lower strain is measured from the torsion applied. As illustrated in figure

6 the residual strength 𝜏𝑟 is related to the normal stress 𝜎𝑁′ applied on the slip surface by:

𝜏𝑟 = 𝜎𝑁′ 𝑡𝑎𝑛𝜙𝑟′

Although for many soils the plot of 𝜏𝑟 vs 𝜎𝑁′ shows a small cohesion intercept 𝑐𝑟′ or a curvature

of the plot will show me that the stress range is applicable to the site conditions and therefore must be

determined by 𝜙𝑟′. In majority of the cases, if the clayey content is 40-50% or more or the plasticity index

is 30-40% or more then the 𝜙𝑟′ value can be expected to be lower than 15° (Lupini et al, 1981).

35


Stability of Earth Slopes

In the design scope for sandy soils, the angle I made by the slope with the horizontal should be

smaller than the angle of internal friction of sand, 𝜙. Normally in loose sands the angle of friction is about

32°, but this angle increase to 40° with very dense sands. It is important to remember that the angle of

slope for stability of cohesionless soil is independent of the height, which may be indefinite. Furthermore,

the weight of the material doesn’t affect the stability of the slope; therefore the safe angle for a submerged

sand slope is the same as that for one composed of dry sand, with the exception of the special case of

damp sand, which has a high angle of repose due to capillary attraction. Special conditions exist with

partially submerged sand slopes affected by tidal conditions which may cause the stability of the fine sand

slope to be considerably less than that of dry sand. Assuming the angle of safe slope with the horizontal is

I then for submerged cohesionless soil slope

𝑇𝑎𝑛 𝑖 = 𝜌 − 1

𝜌 + 𝑒𝑡𝑎𝑛𝜙

Conditions which exist with submerged slopes subject to sudden draw-downs; may be caused in a similar

way with embankments of fine sand exposed to rainstorms sufficiently heavy to result in saturation of the

sand fill.

Clay Slopes

A slip which has taken place in clay slope has three definite characteristics for example a crack

appears at the top of the bank, a portion of the material in the bank slips downward and there is a heave at

the toe as shown below.

Figure 16: Stability of clay slope

In a bank of homogeneous clay materials the slip line failure in the slope closely follows the arc of circle,

and for stability

𝑊 × 𝑑 = 𝐿 × 𝑠 × 𝑅

Where W = weight of the segment of soil of unit thickness, L = length of arc segment, R = radius of the

cylindrical surface of shear, d = distance of the centre of gravity of the segment from a vertical through

36


the centre of curvature and s = average intensity of shear resistance per unit area of the cylindrical

surface. From the above equation it can be observed that in order to investigate the stability of clay slope

it is necessary to ascertain the weight of the soil, the apparent cohesion and the angle of internal friction

of soil concerned. These data may be obtained from shear tests as shown in earlier sections under the

heading shear strength.

It must be noted and emphasised that whistle the height of sand slopes are entirely different and

safe slope is a function of height. Sands possess an angle of repose; whistle clays do not have such a

characteristic, although their behaviour is measured by their shear strength. If for intense the factor of a

clay slope is F, then

𝐹 =𝐿 × 𝑠 × 𝑅

𝑊 × 𝑑

The factor of safety cannot be considered as something absolute, as slope with factors of safety

less than unity have proved to be stable, but in the design of new works for cuttings and embankments it

is advisable to maintain a safety factor between 1.25 and 1.5. The centre of critical circle is found by trial

and error for the minimum value of the factor of safety, and the following notes will assist in the

determination of centre for this circle. So if the shear strength increases with depth then the slopes are

steeper than 45°, then we can use the following table3 for values of angle 𝛼 𝑎𝑛𝑑𝛽 to find the centre of the

arc in figure 12.

Table 3: Fellenius’s Construction for centre of Rotation

Slope Angle of Slope (°) Angle 𝜶 (°) Angle 𝜷 (°)

1-0.58 60 29 40

1-1 45 28 37

1-1.5 33-47 26 35

1-2 26-34 25 35

1-3 18-26 25 35

1-5 11-19 25 37

37


If slopes are flatter than 45°, or if the clay is homogenous, then the centre of the critical circle lies on

vertical through the mid-point of the slope. The circle tends to be deep and would tangent at an

underlying layer of harder clay is such a stratum existed. However when making adjustments for the

centre of the critical circle, horizontal movements are more likely sensitive than vertical one.

Figure 17: Method of Slices for clay Slopes

Figure 13 shows a problem involving a railway cutting that was made in clay which tended to

increase in shear strength with depth. The soil has a weight 120 lb.per sq. cu.ft and an average shear

resistance of 600 lb.per.sq.ft. It was assumed that the cutting was 6m, and it is proposed to adopt slope of

1 to 1.5. Using the values from table 3 we can ascertain the centre of the critical circle and calculate the

safety factor for the proposed slopes. O the centre of circle, to the arc from the toe of the slope to the top

of the vertical section 8. The weight of each slice is proportional to the centre ordinate and the average

height of each slice as plotted vertically below the curvature.

Stability analysis of the infinite slope

The limit equilibrium method is used for the analysis of finite slopes. Slopes extending to infinity

do not exist in nature. For all practical purposes any slope of great extent with soil conditions essentially

same for all identical depth below the surface are known as infinite slopes. Infinite slopes in dry

sand: The figure 14 shows the failure conditions for an infinite slope of cohesionless soil.

The factor of safety of an infinite slope is defined as the ratio of soil strength in the required shear stress

for equilibrium. The factor of safety against sliding is given by

38


𝐹 =𝜏𝑓

𝜏=

tan (𝜙)

tan (𝑖)

Where 𝜏𝑓=shear strength

𝜏 = mobilized shear strength due to gravity

𝜙 = angle of internal friction

i = inclined angle of slope.

Figure 18: Failure condition for an infinite slope of cohesionless soil

Infinite slope in 𝑐 − 𝜙 soils in 𝑐 − 𝜙 soil, the slope is stable as long as the slope angle i is equal

to or less than the angle of internal friction 𝜙. If the slope angle i, is greater than 𝜙, the slope can be stable

only upto limited height known as critical height is given by

𝐻𝑐 =𝐶

𝛾(tan(𝑖) − tan(𝜙))𝑐𝑜𝑠²𝑖

Where C= cohesion

𝛾= unit weight

39


𝜙= angle of internal friction

i = slope angle

𝐻𝑐= critical height

If the factor of safety Fc is applied in cohesion, the mobilized cohesion at depth H, given by

Then the depth H calculated by using mobilized cohesion Cm will not be critical. The factor of safety

against height also represents the factor of safety with respect to cohesion Fc.

Fc is given by

A dimensionless parameter called a stability number is often useful for the analysis of slopes of 𝑐 −

𝜙 soils and can be defined by the following equation

Where 𝑠𝑛 = stability number (a dimensionless quantity). The reciprocal of stability number is known as

stability factor.

40


Figure 19: Failure condition of an infinite slope of cohesive soil

Stability analysis of finite slopes; Failure of finite slopes occurs along a curved surface. In

stability analysis of finite slopes, the real surface of rupture is replaced by an arc of a circle. As to the

mode of failure, the slope may fail basically in the following two ways; 1 the failure surface passing

through the toe of the slope or above the toe of slope is known as slope failure. 2 the rupture is deep

seated and passes through the embankment supporting soil below the toe of the slope is known as base

failure.

1. Slope failure above toe (or face failure)

2. Slope failure through toe (or toe failure)

41


3. Base failure

The base failure generally occurs particularly when the soil beneath the embankment is softer and

more plastic than the slope forming soil itself. There are several methods available for the stability

analysis but the following methods are simple and widely practiced.

1. Slip circle method (Swedish circle method)

2. Friction circle method

Swedish circle method or method of slices: The method assumes the surface of sliding is an arc of

a circle. This was established by studying the failure or embankments in Sweden. (fig.16)

Analysis of the purely cohesive soils ( = 0 analysis)

Consider a likely circular slip surface AD (fig.16) with centre at O.

Figure 20: Slip circle: Cohesive soil

42


The disturbing moment of the cylinder of the soil about O= Wx. Where ‘x’ is the distance of the

line of action of W from the vertical line passing through the centre of rotation. If Cu is the unit cohesion,

L= length of the slip arc,

𝐴𝐷 =2𝜋𝛾𝛿

360

The shear resistance developed along the slip surface will be equal to Cu.L. The resisting moment

preventing the soil from moving is all due to friction along arc length AD which has a lever arm equal to

radius r about O. Resisting moment = cohesion x arc length AD x r = CuL.r Factor of safety against

sliding = (Resisting moment / Disturbing moment)

= 𝑐𝑢𝐿𝑟

𝑊𝑥

A series of slip circles are checked, and the lowest factor of safety is the likely failure plane.

𝑐 − 𝜙 Soil (𝑐 − 𝜙 Analysis). In order to study stability of the slope of a 𝑐 − 𝜙 soil, a possible slip circle is

chosen and divided into strips of equal width as shown in fig 17 (a) and (b).

a) Slip circle: Friction soil

Figure 21

b) One strip

43


Consider one strip as shown in figure 17(b). The forces between the slices are neglected. Vertical

weight W can be considered in two components (1) at right angles to arc of circle (normal component)

and (2) tangential to arc of circle (tangential component).

Disturbing moment about centre O = T x r

Total driving moment = 𝑟Σ𝑇

Where Σ𝑇 = algebraic sum of all tangential components

The resisting force on one strip is made up of cohesion and friction and is given by:

Resisting force= 𝑐Δ𝐿 + 𝑁𝑡𝑎𝑛 (𝜙)

Resisting moment on one strip= 𝑐Δ𝐿 + 𝑁𝑡𝑎𝑛 (𝜙)

Total resisting moment = [𝑐ΣΔ𝐿 + 𝑡𝑎𝑛 (𝜙)Σ𝑁]𝑟

[𝑐ΣΔ𝐿 + 𝑡𝑎𝑛 (𝜙)Σ𝑁]𝑟

Where Σ𝑁 = sum of all normal components, L= arc length

Factor of safety 𝐶𝐼 + tan(𝜙)Σ𝑁

Σ𝑇

Friction circle method (Fig 17. (a) and (b)): The friction circle method of stability analysis of

slope is applicable to 𝑐 − 𝜙 soils. The friction circle method also assumes the failure surface as the arc of

a circle.

44


Figure 22: Friction circle method

Fig.18 shows a failure arc of radius r with O as the centre. In the friction circle method of analysis

of 𝑐 − 𝜙 soil, the resultant reaction vector R at an obliquity of 𝜙 to an element of the failure arc will be

tangential to the small circle of radius kr.sin 𝜙. The small circle of radius kr.sin 𝜙 is therefore called the

friction circle.

In the 𝜙 circle system with a known 𝜙, the following quantities are known.

i. The magnitude and direction of weight of sliding wedge (W)

ii. Direction of resultant reaction (R)

iii. The direction of total cohesion CI (parallel to the chord, I = AC)

To determine the magnitude of R and cohesion Cm, force triangle is constructed in which

magnitude of W is known. The factor of safety with respect to cohesion based on the assumption that

frictional strength has been fully mobilized, is given by

𝐹𝑐=

𝐶𝑐𝑚

A number of slip circles are analysed and the lowest factor of safety is the likely failure plane.

Stability of the slopes of earth dam: Earth dams must be safe against slope and foundation failure

for all operating conditions. There are three generally recognised critical stages based on pore pressure fir

which the stability of the embankment should be ascertained. These three situations are (i) end of

construction, (ii) steady-state seepage and (iii) rapid drawdown.

Usually construction pore pressure reaches their maximum values when the embankment reaches

maximum height. After the reservoir has been filled for a long time, pore pressure is determined by steady

state seepage conditions and may be estimated by the construction of flow net. Rapid lowering of the

reservoir produces the third critical situation, particularly for low permeable soils. The upstream slope

stability can be critical for the construction of rapid drawdown condition. The downstream slope should

be checked for the construction and steady state seepage condition.

45


Consolidation Theory

Previously the effect of seepage was assumed that the volume occupied by the water per unit of

volume of the soil was independent of the state of stress in the soil. If this condition was satisfied the

quantity of water which flows out of an element of soil such as a, flow represented in hydrostatic pressure

conditions on four sides of a prismatic element of sand, is equal to the quantity of water which enters the

element, regardless of whether or not the state of stress in the soil changes. This condition, known as

continuity condition, is expressed in a mathematical terms by the differential equation as shown below.

Equation 9: Differential equation

𝜕𝑣𝑥

𝜕𝑥𝜕𝑥𝜕𝑧𝜕𝑦 +

𝜕𝑣𝑧

𝜕𝑧𝜕𝑧𝜕𝑥𝜕𝑦 = 0

There is no real soil which strictly satisfies the continuity condition, because every change in the

state of stress produces a certain change in the volume of voids,Δ𝑛, per unit volume of soil. Yet if the soil

is very permeable and not very compressible, the change of the porosity due to a change in the state of

stress in the soil can usually be disregarded.

(Terzachi, 1943) stated that a change effective stresses in a highly compressible soil, such as clay

or sand-mica mixture, is likely to produce an important change Δ𝑛 in the volume of voids 𝑛. Hence if the

voids of such a soil are completely filled with water and remain in that state a change in effective stresses

involves a change in water content of soil. Every process involving a decrease of the water content of a

saturated soil without replacement of water by air is called a process of consolidation. The opposite

process is called a process of swelling, which involves an increase of water content due to an increase of

the volume of voids. A further complication arises if the soil combines high compressibility with low

permeability. Both of these properties are exhibited to a high degree by fat clays. In soils with such

characteristics, changes to the water content due to a change in the state of stress takes place very slowly,

because the low permeability of soil does not permit a rapid transfer of the water from one part of the

mass of soil to another or to an adjoining highly permeable, compressible stratum. This phenomenon

produces a time lag between a change of the external forces which acts on a feebly permeable,

compressible stratum and the corresponding change of the water content of soil. It is the principal cause

of the progressive settlement of foundation on clay and of many other processes of outstanding practical

importance.

Some assumptions where made in the theory of consolidation with a few exceptions to all

existing theories of consolidation prior to 1943. The assumptions being that the voids of the soil are

completely filled with water; both the water and solid constituents of the soil are perfectly

incompressible; Darcy’s law is strictly laid; the coefficient of permeability k is a constant; and the time

lag of consolidation is due entirely too low permeability of soil. The theories contained in the following

articles are used based on the following supplementary assumptions, unless a departure from these

assumptions is specifically mentioned. The clay is laterally confined; both the total and the effective

normal stresses are the same for every point of any horizontal section through the clay and for every stage

of the process of consolidation; an increase in the effective pressure from an initial value �̅�𝑜 to a final

value �̅� reduces the void ratio of the clay from an initial value 𝑒𝑜 to a final value e; the ratio

𝑎𝑣𝑐 =𝑒𝑜 − 𝑒

�̅� − �̅�𝑜𝑔𝑚−1𝑐𝑚²

46


Is assumed to be a constant for the range of pressure �̅�𝑜 to �̅�. It is called the coefficient of

compressibility. If the effective pressure is reduced from an initial value �̅� to a final �̅�’ the void ratio

increases from an initial e to a final e’. The ratio

𝑎𝑣𝑐 =𝑒′ − 𝑒

�̅� − 𝑝′̅𝑔𝑚−1𝑐𝑚²

Is also assumed to be a constant for the range of pressure �̅�𝑜 to �̅�’. It is called the coefficient of

elastic recovery. From the first equation we obtain the following

𝑎𝑣𝑐(�̅� − �̅�𝑜) = 𝑒𝑜 − 𝑒

The quantity 𝑒𝑜 − 𝑒 represents the decrease of volume of the voids in a block of soil with the

initial volume 1+𝑒𝑜. The initial voids ratio 𝑒𝑜 corresponds to a volume of voids per unit volume of soil

𝑛𝑜 =𝑒𝑜

1+𝑒𝑜 and the final voids e to 𝑛01 =

𝑒

1+𝑒𝑜 therefore the decrease ∆𝑛 of the volume of voids per unit of

the initial volume of the soil is denoted as

∆𝑛 = 𝑛0 − 𝑛01 =𝑎𝑣𝑐

1 + 𝑒0

(�̅� − �̅�0) = 𝑚𝑣𝑐(�̅� − �̅�0) = 𝑚𝑣𝑐∆�̅�

Wherein ∆�̅� is the increase of the effective unit pressure. The value

𝑚𝑣𝑐 =𝑎𝑣𝑐

1 + 𝑒0𝑔𝑚−1𝑐𝑚²

This is called the coefficient of volume decrease. The corresponding value for a process of

swelling due to a reduction of the effective pressure is

𝑚𝑣𝑠 =𝑎𝑣𝑠

1 + 𝑒0

This is called the coefficient of volume expansion. If there is no possibility of misunderstanding

the second subscript on the symbol 𝑎𝑣 and 𝑚𝑣 will be omitted. The preceding assumptions determine the

physical properties ascribed to the ideal clay which constitutes the subject of the subsequent theoretical

investigations. The equation listed above represents a crude approximation of the relation between the

effective pressure on real clay in a state of complete lateral confinement and the corresponding void ratio.

Soil Compaction

Soil compaction means increasing soil density that makes working with soil easy, helps in

erecting stable structures, and reduces maintenance costs. Compaction of soil brings stability and strength

with it. Foundations fail most commonly because of improper compaction methods or poorly compacted

soil that allows water to seep through the foundation and cause structural damage. Implementing

mechanical methods to compact soil means densifying the soil, filling the pore spaces, improving the

shear resistance of soil, and providing better water movement through the soil particles. The compaction

process largely depends upon the type of soil you are dealing with because different soils have different

physical properties and accordingly different compaction methods should be adopted. Compaction also

prevents frost damage of soil and increases its durability.

47


Factors Affecting the Compaction Process

Compaction of soil depends upon various factors. Among them, grain size distribution of soil,

optimum moisture content, maximum dry density, layer thickness, and environmental factors are some of

the important things to consider. Optimum moisture content (OMC) is the percentage of water present in

soil mass at which a specific compaction force can dry the soil mass to its maximum dry weight. The

adjacent graph shows that the void ratio at OMC is approximately zero and soil is densely compacted. For

different types of soils, OMC and maximum dry density curves are different.

In the figure 19, W stands for water content and ρ (d) stands for Dry Density of soil mass.

Standard Proctor and Modified Proctor tests are conducted to determine OMC and the dry density

of soil masses. The basic difference between these two tests is the size and weight of hammer used to

compact the soil mass. The number of blows remains the same, but the falling height is changed from 12

inches to 18 inches in the Modified Proctor test. Other popular methods of determining OMC and

maximum dry density are mentioned below.

Sand Cone Test - Suitable for a large sample, delivers accurate results but requires huge area and

more time to perform.

Shelby Tube Test - Suitable for deep and under pipe haunches, not suitable for gravels and only

works for a small sample.

Nuclear Gauge Test - Statistically reliable, easy to redo and fast method.

Different Compaction Methods

Compaction of soil, in simple words, means applying external pressure to the soil mass so that its

characteristic properties improve with regards to construction purposes. Technically speaking, static and

vibratory forces bring soil particles together by exerting pressure on them. Static forces apply load on the

surface of soil particles, exerting dead weight of the machine in a downward direction. These forces do

not go skin deep and work only for the upper surface of the soil mass.

Vibratory forces, on the other hand, work for the whole soil mass and are not limited to the upper

surface only. Along with the dead-weight of machine, compactors and vibrators are connected that not

only exert pressure on the soil mass, but also shuffle the entire soil mass so that the overall soil mass is

Figure 23: Optimum Moisture Content graph

48


compacted uniformly. Both the top and deeper layers get blows from the vibrator and compactor resulting

in denser and tightly packed soil.

For mechanical soil compaction, four main compaction techniques are mentioned here.

1. Kneading Compaction

2. Pressure Compaction

3. Vibration Compaction

4. Impact Compaction

Compaction equipment is selected based on the type of soil. For clayey soils, kneading techniques

and equipment have to be used because clay soils exist in the form of clods and kneading is the best way

to break the clods and densify clay soils. On the other hand, for granular soils, a vibratory or shaking

motion of the compacting device is required so that uniform compaction is achieved. Popular compaction

equipment types are mentioned below.

Smooth Wheel Rollers - Single axle, equipped with a steel cylinder, sand or water are used to

increase its self-weight. Pushes the soil in the direction of movement and results in soil compaction.

Sheep Foot Roller - Different style of sheep rollers, can be used for different types of soil, and best

suited for cohesive soils. Covers less surface area but pressure per unit area is very high resulting in

healthy compaction.

Vibratory Drum Roller - Suitable for compaction of sand, gravel, asphalt, and other heavy aggregates.

Very powerful compaction devices provide uniformly dense soil because of vibrator attached to them.

Vibratory Pad foot Compactors - Compactors work mainly in landfill regions and pad foot

compactors have pads attached to their drums making them work fast and deliver efficient results in

confined and tight areas.

Tamping Foot Rollers - Basically, these devices are compactors but are popularly known as tamping

foot rollers. Kneading, impact compaction and pressure compaction happen simultaneously with

these devices.

Different classification systems divide soils according to their characteristic properties and accordingly

compaction method is selected.

49


Chapter 3 Case Study

Case study: New Zealand bridge scour experiences

By Stephen E. Coleman1 and Bruce W. Melville2

Abstract: Details of three case studies of scour damage for New Zealand bridges are presented.

These cases cover ranges of bed materials, flood magnitudes, and river morphologies, and illustrate a

range of scour processes occurring at bridge foundations. The resulting scour predictions highlight the

value of a judicious use of the proposed methodology, the effect of sediment supply and transport balance

considerations for engineering projects, aspects of river morphology to be considered in bridge foundation

design, and that the combination of various components of scour needs to be considered when assessing

bridge scour. Soil reinforcement has many scenarios here is the solution to prolonging the bridge

abutment life.

Scour at a bridge crossing a river can be classified as general scour, contraction scour, or local

scour. General scour occurs irrespective of the existence of the bridge and can occur as either long-term

or short-term scour. Short-term general scour develops during a single or several closely spaced floods.

This of scour includes scour at channel confluences, scour arising from a shift in the channel thalweg or

braids within the channel, scour at bends, and bed-form migration. Long-term general scour has a

considerably longer timescale, normally of the order of several years or longer and includes progressive

degradation and (lateral) bank erosion. Degradation is the general lowering of the riverbed that occurs, for

example, downstream from a dam. Bank erosion may result from channel widening, meander migration, a

change in river controls, or a sudden change in the river course (e.g., with the formation of a meander-

loop cut-off).

Bulls Road Bridge

The Bulls Road Bridge crossing the Rangitikei River was opened in 1949 and is located along

State Highway 1 between Bulls and Sanson on the North Island of New Zealand. Pier Q of the 2-lane, 19-

span structure was undermined by scour on June 15, 1973 (Figs. 2, 3). Each pier of the bridge was a

reinforced concrete slab type founded on two rows of six vertical 0.4 m octagonal reinforced concrete

piles. In 1941, a test bore log at Pier Q (Fig. 3) indicated a 4.7-m thick gravel surface stratum underlain by

a thin mudstone layer and 3.8 m of fine black sand. The sand contained particles composed of the mineral

magnetite of specific gravity Ss = 5.12. A bore log taken at Pier Q subsequent to the 1973 failure event

indicated similar stratigraphy. The piles at Piers P and Q extended to about the level of the hard gravels

underlying the black sand (Fig. 3).

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The largest floods at a Rangitikei River gauging site approximately 80 km upstream of the bridge

site occurred in 1897 and 1926, with peak flows of 3,800 and 1,798 m3/s respectively. The 1897 flood

severed all six bridges over the Rangitikei River at the time. The mean flow recorded at this gauging site

is 63 m3/s. The average annual maximum flow over the 42-year continuous record (1955–1996) at this

site is 733 m3/s. Metal was extracted from the river from 1949 onward. With six extraction sites within

2.7 km upstream and 2.2 km downstream of the bridge, the metal extraction resulted in the mean level of

the bridge cross section falling steadily by 0.5 m over the period 1945–1972. The minimum bed level at

the bridge site fell relatively steadily by 3.0 m from 1945–1970, then rapidly by 1.7–2 m from 1970 until

the 1973 failure. A high terrace that is relatively erosion resistant defines the southern bank of the wider

channel. The main flow channel deflected off this terrace just upstream of the bridge and passed

underneath the southern end of the bridge in the early 1970s (Fig. 3). Anecdotal evidence indicates the

existence of about 6 m depth of scour for the main channel at Pier Q about 1 month prior to the failure.

The deepening of the main flow channel under the bridge because of local and general scour resulted in

an old timber pier being exposed immediately downstream of Pier Q in 1971–1972 (Fig. 3).

The peak flow during the June 1973 flood was 675m3/s, and was estimated to approximate an

annual flood event. During this flood, the southern river terrace and significant debris built up against the

timber pier downstream of Pier Q concentrated flow at Pier Q at an angle of 557 to the bridge centreline.

The reduced fixity at the base of the piles of Pier Q as the scour developed combined with water flows at

an oblique angle exerting lateral pressures on the exposed piles resulted in hinging occurring both at the

base of the piles and also at the underside of the pile cap and the collapse of Pier Q on June 15. The

collapse of the bridge deck connecting Piers P and Q also caused the rotation of Pier P, with the

suspended spans adjacent to Piers P and Q falling into the river (Fig. 3). An empty school bus being

driven over the bridge at the time of failure struck Pier R during its descent into the river. The driver of

the bus was injured but survived the incident.

The Bulls Road Bridge failure can be attributed to a combination of general scour arising from

gravel mining and local pier scour. The local scour was exacerbated by the obliqueness of the flow to the

pier, the flow constriction caused by the debris-enlarged timber pier immediately downstream of the

51


bridge pier, and the depth of local scour for the flow oblique to Pier Q having eroded through the

overlying protective gravel and mudstone layers to the easily erodible fine sand stratum (Ettema 1980).

The maximum depth of scour measured below the armoured bed level was about 12.2 m. Ettema (1980)

comments that the scour could have developed deeper had if not been arrested by the lower strata of hard

gravel and mudstone and had the pier not collapsed into the scour zone.

Qualitative Analysis of Expected Scour Development

An analysis of local scour for the layered sediments is presented by Ettema (1980). General scour

is discussed below. In 1947 the river was significantly braided, with numerous channels upstream of the

bridge site (Fig. 4). Flow was principally in the vicinity of the middle of the bridge, although it was

distributed across the site. The bed was relatively uniform over the width of the cross section. As

indicated above, from 1949 onwards, sediment was removed from the river upstream and downstream

from the bridge site. Qualitative predictions of stream response to watershed changes can be made on the

basis of the balance relation by Lane (1955) and shown in Appendix I. For a decrease in sediment

transport rate Qs owing to instream mining, this relation predicts a decrease in stream slope Se, with

accompanying degradation at the bridge site owing to the reduced sediment supply.

Indicative relations between channel slope and other channel parameters (Shen et al. 1981;

Richardson et al. 1990; Schumm 1971) further suggests a possible transition from an existing braided

channel form to a meandering channel for the reduction in sediment supply. A possible associated

reduction in channel width, increase in flow depth, reduction in channel aspect ratio, and increase in

channel sinuosity are also predicted. With the removal of sediment, the number of braided channel

branches for the Rangitikei River was reduced to 1–2 main channels meandering within the wider channel

(Fig. 4). The new flow channels were deeper and narrower than previous channels. Channel width

decreased sharply by 37% over the period 1963–1967, with a significant left-hand bend/right hand bend

combination developing immediately upstream of the bridge during 1966–1970.

With the increasing sinuosity of the river

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channels, river control works were developed over the period 1949–1973 to protect the outsides of the

meander bends upstream of the bridge. The river meanders, reinforced by the bank protection works,

provided a controlled direction of flows toward the erosion-resistant southern riverbank, these flows being

de-flected to pass underneath the southern end of the bridge at an angle of about 557 to the bridge

centreline in the early 1970s (Figs. 3, 4). Bank protection works then acted to reinforce the concentration

of flows resulting from the sediment mining operations; the combination significantly intensifying scour

conditions for Pier Q, which subsequently failed for essentially an annual flood event. The Bulls Road

Bridge case provides a classic example of the effect of sediment supply and transport balance conditions

on engineering projects, the effects of intervention into the river in the vicinity of the bridge site clearly

impacting on the bridge stability. The implications of such considerations are readily apparent for many

regions in which sediment mining has been and is presently taking place. Variability in river course

clearly needs to be addressed in bridge scour design. This is reflected in the main river channel for the

Bulls Road Bridge, moving to pass principally between Piers J to M toward the centre of the bridge from

1992 onward (Fig. 4). Any underpinning of the piers around Pier Q that was undertaken in 1973 would

then also need to be considered for the piers located in the present main channel. The 2 other cases on

Blackmount road Bridge and Mahitahi river road bridge MRRB is available:

http://ascelibrary.org.ezproxy.brad.ac.uk/doi/pdf/10.1061/%28ASCE%290733-

9429%282001%29127%3A7%28535%29. Last accessed on 19/02/15

Definitions

Abutments comprise several structural parts, notably an abutment column supporting one end of a

bridge deck, and the column which is set amongst, or backed by, a compacted earth fill approach

embankment. This chapter may use the term “embankment/abutment” to describe the full structure –

approach embankment and abutment column structure, but where necessary the separate terms will be

applied for clarity. The “embankment” is considered to be the earth fill that extends from the abutment

column into the floodplain away from the stream, while the term “abutment” refers to the column and the

support structure facing the stream. Chapter 2 Literature review covers/describes the main structural

consideration when it comes to design.

Abutment scour herein is taken to be scour at the bridge-opening end of an abutment, and directly

attributable to the flow field developed by flow passing around an abutment. It includes the effects of

flow acceleration due to channel flow constriction as well as local large-scale turbulence effects due to

flow separation which are present in varying relative proportions depending on the upstream approach

flow distribution and flow distribution at the bridge section; abutment column type, foundation type and

location; flow curvature; and near-field river morphology.

The Section that follows explains the current understanding of abutment scour. Abutment scour

may cause embankment failure, abutment column failure, or both. Observations of abutment scour

indicate that scour frequently may initiate a geotechnical-type failure of the earth fill embankment.

Failure of the abutment column itself is less commonly observed. Although failure of the embankment

may occur with the abutment column (and bridge structure) remaining intact, it is a most undesirable

condition that renders the bridge approach dangerous for road vehicles.

Motivation for review

The need to evaluate present knowledge about abutment scour processes and failure conditions,

and determine the extent to which existing scour-estimation methods reflect this knowledge, is expressed

in several publications prepared by national agencies and societies in the US: e.g., NCHRP Reports p24-

08 (“Scour at Bridge Foundations”) and 2007(p178) (Parola et al. 1996, Lagasse and Zevenbergen 2004),

as well as NCHRP Report (p417) (Parola et al. 1998), USGS (2003), and Kattell and Eriksson (1998).

53


However, few situations of water flow and boundary erosion are more complex and challenging to

understand than those associated with scour of bridge abutments. The sketches in Figures 32 and 33

convey a sense of the complexities faced during estimation of scour depths at bridge waterways.

Abutment sites may vary widely in their specific details. Figure 32 illustrates a wide, multi-span bridge

whose abutments are considerably set back from the bank on broad flood plains. As depicted in Figure 33,

the abutments for shorter bridges are in close proximity to each other; in such cases the abutments often

may be set close to the bank of a channel whose morphology is quite irregular and varies markedly with

flow stage. Both Figures 32 and 33 indicate how flow approaching a bridge waterway converges then

diverges once through it. As it does so, it passes around bluff bodies, generating, transporting, and

eventually dissipating large-scale turbulence structures (large eddies shed in a recognizable pattern due to

flow separation albeit intermittently with time). The flow is bounded by erodible boundaries of complex

and changing form that have widely varying compositions and characteristics. Even the classification of

abutment scour as an independent bridge scour component is problematic, because contraction scour and

abutment scour are linked processes that usually occur together during flood events.

Figure 24: Schematic of long, multi-span bridge over a compound channel.

Furthermore, the sketch in Figure 34 shows the effect of hydraulic erosion of bed and banks on the

integrity of certain boundary components (banks and embankments) after a geotechnical slope-stability

failure. Such failures add additional complexity to waterway flow and scour, and thereby to scour-depth

estimation. It can be readily appreciated from Figures 32 through 34 that scour indeed is a long-standing and

vexing problem in hydraulic-engineering research, not to mention bridge foundation design.

Figure 25: Schematic of relatively short bridge over a narrow main channel

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Figure 26: Abutment scour resulting in embankment failure by collapse due to geotechnical instability.

The development of practical design methods for predicting scour depths at bridges has been

hampered by inadequate knowledge about, and formulation of, important component processes and their

interaction during scour. The scour-estimation methods presently available do not adequately take all these

considerations into account. As would be expected, early work on abutment scour focused on the simpler and

idealized situations of scour; notably, abutment scour simulated as scour at a rigid structure extending at depth

into a bed of uniform sand. Commensurately, the existing relationships and guidelines apply to simplified

abutment situations, such as an abutment placed in a straight rectangular channel, and are roughly based on

empirical or regression equations fitted to a collection of data from laboratory tests with model abutments

(whose construction does not always resemble that of actual abutments). Such design relationships can only be

extrapolated with considerable uncertainty to actual field conditions. That extrapolation often results in overly

conservative estimates of scour depth. Conservatism is understandable and indeed useful, but can be

expensive for large abutment structures. Moreover, when existing design methods inadequately embody

certain scour processes, there is a risk that the manner or location of actual scour failure will differ from that

assumed for the estimation relationship or guideline. Additionally, an overlooked process may trigger or

exacerbate scour at a site where a scour problem had not been anticipated. There are several prominent

knowledge gaps about processes whereby scour could occur in ways and places not accounted for by existing

prediction methods or programs of bridge monitoring (e.g., geomorphic change in channel alignment,

inadequate estimation of peaks and periods of design flows, proximity of old or new bridges, the role of large-

scale turbulence, inadequate control of storm-water drainage at the bridge site). These gaps are not only

limited to flow and geomorphic processes but also relate to sediment type in terms of fine-grained sediments,

which experience interparticle physico-chemical forces, and coarse grained sediments whose movement is

resisted by gravity forces alone.

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The threat posed by scour was realized early in the struggle to construct and maintain bridges.

Over the ages it has been dealt with in several ways, but the threat has not yet been adequately addressed.

In antiquity, for instance, Roman engineers recognized the threat. Whenever they built a new bridge they

usually would place on the bridge an appeasing inscription to Janus, the Latin god of bridges (and portals

generally), or to the local deity of the river or stream being crossed. Engineers in Japan and Korea reduce

the threat for bridge abutments at major, levee-flanked rivers flowing through heavily urbanized regions.

They do so by not locating bridge abutments on the floodplain, but instead locating them outside the

levee; in this manner, flow contraction through a bridge waterway is minimized or practically eliminated,

and the abutments are not exposed to scour. On the whole, though, bridge scour continues to be a threat.

The case depicted in Figure 35 is an example of scour failure that occurred fairly recently (1993 flood in

Midwestern U.S.) for an unusually large flow that exceeded the design flow for the bridge waterway. The

maximum scour depth measured two months after the flood was 17 m in the floodplain on the upstream

side of the bridge (Parola et al. 1998).

Figure 27: Scour at I-70 bridge over Missouri River from 1993 flood. Flow was from left to right. (Photo

from Parola et al. 1998).

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Abutment form and construction

The main design characteristics of an abutment can be described in terms of abutment form, the

overall layout of an abutment’s approach embankment, and the abutment’s construction configuration.

These characteristics, together with the waterway’s channel morphology, boundary sediments and soils,

as well as flow-resistance features (e.g., vegetation state of the floodplain), influence the flow field

around the abutment, and therefore, scour. A striking, and somewhat complicating, characteristic of

bridge abutments is that few abutment situations are alike, as Figures 32 and 33 exemplify. Accordingly,

the development of a method for estimating scour depth at abutments requires that the abutment forms,

layouts, and construction configurations of common practical importance be identified.

Abutment form

Two general forms of abutment exist as illustrated in Figure 36: 1. Wing-wall abutments,

including vertical-wall abutments; and, 2. Spill-through abutments Spill-through abutments have sloped

sides, whereas wing-wall abutments have a vertical face and wing-walls that retain an earth fill approach

embankment. The wing-walls can be oriented at various angles to the abutment’s central panel, although a

45o angle is representative. A wing-wall abutment with wing-walls angled at 90o to its central panel is

sometimes called a vertical-wall abutment, and it is fairly common for small abutments. Sheet-pile

caissons extending into channels also may be viewed as a type of vertical-wall abutment. Various

alternative names exist for these two general abutment forms.

Figure 28: Plan views of the two common abutment forms: (a) Wing-wall; (b) Spill-through (Ettema et al. 2010).

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Abutment layout

In a somewhat simplified manner, it is useful to discuss abutment layout in terms of the length,

L, of approach embankment, floodplain width, Bf, main channel width, Bm, overall width of the main

channel and floodplain at a bridge crossing of a waterway, B, and embankment top width, W. These

variables are indicated in Figure 37 except for W which is shown in Fig. 36.

Figure 29: Definitions of embankment length, floodplain width, and main channel width (Ettema et al. 2010)

Bridge abutments can be characterized as conforming to the following layout arrangements,

which can be represented in terms of the variables L, Bf, and B: 1. the abutment is located on the

floodplain of a compound channel (L ≤ Bf). This layout is typical for spill-through abutments. It is usual

for the abutment to be set back from the main-channel bank so that a vehicle (and wildlife) can pass

between the abutment and the bank. A minimum setback distance of about 10 ft (3.05 m) is common

practice, if site layout allows, but the setback distance on large rivers with wide floodplains may be

considerably more ; 2. The abutment extends up to the bank of the main channel (L ≈ Bf). This layout is

typical for wing-wall abutments, especially for channels having a narrow, or no, floodplain. Wing-wall

abutments are common for bridges over small streams; and, 3. The abutment is located in a rectangular

channel, and no floodplain is present. This layout is not common, although it is essentially similar to a

relatively short abutment on a wide floodplain and is representative of wide-braided channels. Also, it is

similar to channel-control structures (e.g., spur-dikes, groins, barbs, hard-points), coffer-dams, and

construction caissons. The nature of an abutment inevitably requires that the layout of an abutment be

tailored to fit the local topography of a bridge site. Therefore, to varying extents each abutment inevitably

differs in layout. Other variations in abutment layout can be found; e.g., many small bridges in Maine

have wing-wall abutments that extend into the main channel (Lombard and Hodgkins 2008).

Abutment construction

It is usual for the top width of the earth fill embankment to accommodate minimally a road width

of 24 ft (7.22 m) plus two shoulders of width 8 ft (2.41 m), giving an overall top width of 40 ft (12.04 m).

The side-slopes of earth fill approach embankments commonly are set at 2H:1V, though slopes range

from about 2H:1V to 3H:1V. Figure 38 is an isometric view of the geometry used for spill-through

abutments. The embankment geometry for wing-wall abutments is essentially similar to that shown in

Figure 38, except that the vertical face of a wing-wall abutment retains the end of the embankment.

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Abutments usually comprise a concrete support wall founded on a pile cap supported by piles or

on a spread footing, and adjoin an earth fill approach embankment. Pile supports are more common than

are footing supports, unless the abutment is founded directly on rock. Spill-through abutments are formed

around a “standard-stub abutment,” which comprises a concrete stub supported by a pile cap on two rows

of circular piles. The design and dimensions of a common standard-stub abutment column are shown in

Figure 39. Wing-wall abutments usually have similar foundation layouts as the standard-stub abutments,

except that they include wing-walls extending from the central stub. Figure 40 shows the design and

dimensions of a common wing-wall abutment.

Figure 31: The geometry and dimensions of a standard-stub abutment commonly used for spill-through abutments (prototype

scale indicated); design provided by the Iowa DOT (Ettema et al. 2010)

Figure 30: an isometric view of the geometry used for spill-through abutments.

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Figure 32: The geometry and dimensions of a wing-wall abutment - compacted earth fill embankment extends back from the

abutment structure (prototype scale indicated); design provided by the Iowa DOT (Ettema et al. 2010).

The elevation of the pile cap and the detailed arrangement of piles may vary from bridge site to

bridge site. At some sites, the pile cap is located at, or near, the top elevation of the floodplain, whereas at

other sites the piles extend upward through the embankment earth fill. In this latter case, the piles directly

support a cross beam, which in turn supports the beams of the bridge deck. Also, for some sites, wing-

wall abutments may be supported by sheet piles driven in approximately the same plan layout as the

abutment. The foregoing descriptions of common abutment forms and construction arrangements are not

reflected in the leading design guides and bridge-monitoring guides addressing scour at bridge abutments.

For example, FHWA’s (2009) guide for bridge inspectors does not fully portray the complexity of an

abutment structure and its flow field, or possible failure mechanisms due to scour, as elaborated in this

report. Chrisohoides et al. (2003) and Ettema et al. (2010), for example, provide useful visualizations of

abutment flow, as currently understood.

Pier proximity

Many bridges over rivers are constructed with a comparatively short first deck span, such that a

pier is located very close to an abutment. There are construction-economy advantages in having the pier

close to the abutment and riverbank, and the arrangement often facilitates a clear span over the river. This

construction advantage, however, raises a question as to whether pier proximity could adversely influence

abutment scour (and vice-versa). Figure 41 depicts an example of a bridge with a pier located close to an

abutment.

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Sediment and soil boundary material

Figure 34: Variation of soil and sediment types at a bridge crossing (Ettema et al. 2010).

The boundary material of the main-channel, floodplain, and embankment components of a

bridge-waterway boundary usually comprise different zones of alluvial sediments and soil, as indicated in

Figure 42. Abutment scour usually occurs within several zones of sediment and soil, leading to different

erosion processes and varying rates of erosion.

Figure 33: depicts an example of a bridge with a pier located close to an abutment.

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Alluvial non-cohesive sediment (sands and gravels) most frequently forms the bed of the main

channel, whereas the channel’s floodplain may be formed from considerably finer sediments (silts and

clays), typically causing the floodplain soil to be more cohesive in character than the bed sediment of the

main channel. The banks of the main channel usually are formed of the floodplain soils, and thus also

may behave cohesively so as to stand at a fairly steep slope. Most abutments have an earth fill approach

embankment formed of compacted soils. The soils may have been excavated from the floodplain or have

been brought to the bridge site from elsewhere. The earth fill embankment is placed and compacted to a

specific value of shear strength so as to support the traffic load. Direct, dynamic simulation of the

strength behaviour of an earth fill embankment or a floodplain soil poses a practical difficulty for

laboratory experiments on scour at bridge abutments. The difficulty is to replicate, at a reduced scale, the

shear strength of a representative earth fill embankment. To date, no study appears to have attempted

experiments that closely replicated the strength behaviour of an embankment with mixed soil types.

Flow field

Flow through a bridge waterway narrowed by a bridge abutment and its embankment is

essentially flow around a short stream wise contraction3 . Figure 43 schematically illustrates the

characteristic flow features and the connection between the contraction and the formation of a complex

flow field around the abutments. The flow width narrows and the flow accelerates through the

contraction, generating macro-turbulence structures (eddies and various vortices spun from the

contraction boundary) that shed and disperse within the flow. Flow contraction and turbulence at many

bridge waterways, though, is complicated by the shape of the channel. It is common for waterways to

traverse a compound channel formed of a deeper main channel flanked by floodplain channels, as shown

in Figure 44. To varying extents, all flow boundaries are erodible. As this figure indicates, the major flow

features of a short contraction prevail at a bridge waterway comprising a two-lane road. The contraction

lengthens for dual-carriageway highways like freeways or expressways.

Figure 35: Flow structure including macro-turbulence generated by flow around abutments in a narrow main channel. (Ettema

et al. 2010).

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Figure 36: Flow structure including macro-turbulence generated by floodplain/main channel flow interaction, flow separation

around abutment, and wake region on the floodplain of a compound channel. (Ettema et al. 2010).

Though the short-contraction analogy is somewhat simplistic, an important point to be made is

that the flow field around an abutment, like the flow field through an orifice, is not readily delineated as a

contraction flow field separate from a local flow field established near the abutment. The two flow

features (flow contraction and large-scale turbulence) are related and difficult to separate. Either of the

flow features may dominate, depending on the extent of flow contraction and the characteristics of the

abutment and its foundation. When an abutment barely constricts flow through the waterway, scour at the

abutment may develop largely due to the local flow field generated by the abutment. This flow field is

characterized by a local contraction of flow and by generation of large-scale turbulence. For a severely

contracted bridge waterway, flow contraction dominates the flow field and a substantial backwater occurs

upstream of the bridge. In this situation, the approach flow slows as it approaches the upstream side of the

bridge, and then accelerates to a higher velocity as it passes through the bridge waterway. When the

foundation of the end of an abutment comprises a solid contiguous form extending into the bed (flood

plain or main channel), scour development may become similar to that at a wide pier where the flow

becomes contracted and large-scale turbulence is produced. Such abutments include situations where a

sheet-pile skirt is placed around the toe of the spill-slope of a spill-through abutment (to protect against

spill-slope instability and failure), or when a wing-wall column is founded on sheet-piles. Embankment

and abutment structures create potentially erodible short contractions. Higher flow velocities and large-

scale turbulence around an abutment may erode the abutment boundary. Commonly, the bed of the main

channel is more erodible than the floodplain, because the bed is formed of loose sediment, while the

floodplain is formed of more cohesive soil often protected by a cover of vegetation. Accordingly, two

prime scour regions typically develop, as borne out by field observations of scour, as indicated in Figure

46:

• One region is where the boundary is least resistant to hydraulic erosion. This could be the main

bed if flow velocities (and unit discharges) are sufficiently large; and,

• The other region is where the flow velocities (and unit discharges) and turbulence are greatest.

This usually is near the abutment.

For the simpler situation of an abutment well set back on a flood-plain, laboratory experiments

indicate that deepest scour usually coincides with the region where flow contraction is greatest (Ettema et

al. 2010, Melville et al. 2006). Figure 45 illustrates this for a spill-through abutment. For spill-through

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abutments comprising erodible embankments flow contraction dominates the abutment flow field. Once

scour begins, the geometry of the bridge waterway (as a short contraction) changes. The deepened flow at

the scour region draws more flow, because flow contraction is locally eased there.

The extent and maximum depth of scour at abutments can be complicated by the mix of materials

forming the compound channel and the abutment’s embankment, and other considerations such as the

proximity of a pier.

Figure 38: Interaction of flow features causing scour and erodibility of boundary (Ettema et al. 2010).

Figure 37: For a spill-through abutment well set back on a flood-plain, deepest scour usually occurs where flow is most

contracted through the bridge waterway.

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Abutment scours as a design concern

The principal design concerns can be expressed in terms of set of questions:

1. What is the greatest scour depth that reasonably could occur near the abutment?

2. Will that scour depth pose a slope-stability problem for the embankment?

3. What scour depth should be used in estimating the required length of pile support?

4. What is the deepest scour that potentially could occur at the abutment column itself?

5. Does that scour occur when the embankment is breached so as to fully expose the abutment

foundation?

Design scour depths

When considering the possibility of embankment failure, two scour depths must be estimated, in

accordance with the design concerns:

1. One scour depth is needed for stable embankment design; and,

2. The second scour depth is required for determining the length of piles underpinning the abutment

column, or elevation of column footing (if a footing foundation is to be used). For design estimation of

scour depths, it is necessary to consider the absolute elevations and locations attained by scour. The

location of deepest scour relative to the concern of embankment stability differs from that associated with

column stability. Additionally, the likely rates or sequences in which the scour develops are important, as

explained in Ettema et al. (2010)

Estimation of scour depths

There are several approaches to estimate the two scour depths mentioned abutment layout. Scour depth

associated with embankment stability subject to scour can be addressed in two ways, as described below.

1. Hydraulic then geotechnical calculations. Estimate the potential maximum depth of scour that may

develop without immediately considering the geotechnical failure of the embankment on the floodplain

near the abutment. Once this scour depth is estimated, its effect on the geotechnical stability of the main

channel bank and embankment can be estimated. If the bank and embankment were found unstable, they

would collapse. Failure of the head slope, or spill-slope, is an undesirable condition, which may have

most serious consequences if road traffic is not immediately prevented from accessing the bridge

approach. The integrity of the abutment column also may be affected by embankment failure, but this

may not be the worst case for the column, as discussed below. Embankment failure acts to relieve flow

contraction, diminish macro-turbulence generation, and consequently reduce the maximum scour depth

attained. The geotechnical strengths of the embankment and floodplain soils, therefore, may significantly

influence abutment scour depth, as well as contribute uncertainty to scour-depth estimation; and,

2. Geotechnical calculation. For given (or measured) geotechnical strength properties of the embankment

earth fill near the waterway, estimate the maximum limiting steepness for embankment stability. The

maximum scour depth attainable then is determined in the context of the limiting maximum steepness of

the embankment. No hydraulics calculation is needed, but the position of deepest scour must be

estimated. An important point here is that the location of maximum scour depth has substantial bearing on

embankment stability and thus the prospect of abutment failure. Once the embankment fails, flow

contraction is relieved, flow area increases, maximum velocity near the abutment diminishes, and scour

65


will not deepen. To be kept in mind, though, is the relative timing of scour development and embankment

failure, and the undesirable consequences of full embankment failure. Scour depth associated with

abutment-column stability should be considered in two ways. First, the abutment-column may be

rendered unstable due to embankment failure as described above. Secondly, following embankment

failure the abutment column may be exposed to the flow in the manner of a pier. This case must rely on a

semi-empirical relationship such as used for estimating scour depth at a bridge pier, because an exposed

abutment essentially is a pier. The complexity of flow field and sediment movement at a column is

practically the same as at a bridge pier. These design concerns are drawn together in more detail in the

NCHRP p24-20 report by Ettema et al. (2010) as a sequence of design steps that take into account

abutment location, geotechnical properties of embankment and floodplain, and the erodibilities of main-

channel bed and floodplain.

An Essential design question

An essential design question to be addressed by agencies designing bridge abutments – and not

addressed during this evaluation study – concerns how abutment design should best take abutment scour

into account. Many experiments and field observations of abutment failure indicate that failure typically

occurs as the geotechnical collapse and washout of the abutment’s earth fill embankment. Under severe

situations, the abutment column also may fail in a manner similar to scour failure of a bridge pier.

Embankment failure may limit the development of abutment scour to a potential maximum depth,

because the exposed embankment soil erodes laterally, increasing the flow area and easing flow velocities

in the area of deepest scour. The essential question leads to the following more specific questions:

1. What scour depth(s) should be considered for abutment design (the potential deepest scour,

scour leading to embankment failure, or scour at an exposed abutment column)?

2. Is embankment failure (with bridge super-structure remaining intact) acceptable?

3. As the embankment near an abutment column often is a relatively weak or vulnerable location

of bridge waterway, what design considerations should be contemplated in order to strengthen

embankments in the vicinity of an abutment column?

Then, how would such strengthening affect abutment scour or scour at a nearby pier? It is

noteworthy that all the illustrations of abutment scour in this report show failure of an abutment’s earth

fill embankment. The example shown in Figure 47 is representative of many abutment failures. Another

example of embankment failure is shown in Figure 48 for a flood in the Atlanta metro area in 2009. Flow

coming from the left floodplain as well as overtopping of the bridge severely eroded the left embankment,

exposed the abutment and resulted in the approach span to the bridge deck falling into the stream.

66


Figure 39:: A common situation of abutment failure; scour has led to failure and partial washout of the earth fill spill-slope at

this abutment. A basic question arises as to how abutment design should take scour into account.

Figure 40: Failure of abutment fill in September 2009 Georgia flood accompanied by failure of approach roadway (Hong and

Sturm 2010).

67


Influence of pier proximity

The influence of pier proximity on the three scour conditions is slight, at least for the pier form

and construction depicted previously in Section sediment and soil boundary. Flume experiments

(NCHRP p24-20) show that abutment scour is dominated by the flow field established by an abutment.

Once scour initiates, and deepens below the pier’s pile cap, pier presence does not substantially increase

flow contraction or the strength of large-scale turbulence structures.

For Scour Condition A at spill-through abutments, pier presence may increase maximum scour

depth by approximately 10% when Lp/W < 2; where, W = embankment top width and Lp = distance from

abutment to pier. The increase results because pier presence close to an abutment slightly increases flow

contraction, as flow is deflected around the pier (as if the abutment were lengthened). For Scour

Condition B, pier presence acts to increase flow contraction but it also acts to partially block the dispersal

of riprap stone. The net influence for Scour Condition B is a lessening of scour depth.

Other scour processes

Abutment scour may develop consequent to several processes of flow and bed-sediment

movement:

1. Localized scour attributable to change in main channel alignment and morphology, which

adversely affects abutment location and orientation relative to flow in the main channel. Lateral shift of a

channel may direct flow adversely towards abutments not designed for a lateral shift in the channel. The

deeper scour commonly resulting from this possibility must be considered in the scour design of

abutments;

2. Scour of the approach embankment flank on the floodplain. This condition may occur when

the floodplain flow converging towards the bridge waterway undercuts the flank of the approach

embankment. This scour mechanism is less common.

3. Erosion along the flanks of an abutment, which may develop because of inadequate control of

road drainage along an abutment. Such erosion exposes the earth fill at the end of the abutment, making

the abutment more prone to erosion by flow in the main channel; and

4. Degradation of the main channel bed. This process occurs in response to an overall propensity

of the main-channel flow to degrade associated with the reduction in the bed-sediment load along the

channel. It also could result from the upstream advance of head-cutting of the channel bed, because the

channel has steepened hydraulically. Bank erosion with channel widening may accompany degradation

and lead to erosion attack of the embankment.

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Chapter 4 Methodology

Embankments Design and Application

What I’ve noticed about the general trend in the methods that have been used prior to my final

year project based on Soil reinforcement are all described in chapter 2 & 3 for designing and investigating

the stability of slopes using soil reinforcement techniques, these may be applied when considering

earthworks such as bridge abutments and embankments. However certain additional features require

consideration and embankments may be divided into two types, depending on whether they retain water

or used for landscaping. The design of earth dams is considered to be outside the scope of the present

volume but the subject may be pursued by the study of the soil properties as quoted above in my

Literature review.

An embankment may fail either by slip or settlement as shown in the chapter 3 Case Study Bulls

road bridge. It is therefore necessary to investigate the soil properties of the ground on which an

embankment is to be built as well as the soil which is used in construction. The settlement of the

embankment, apart from soil compaction of the bank itself, may be due to the consolidation of the

underlying material due to the load of the fill, and this aspect will be referred to in a later section dealing

with the consolidation analysis. If water were to be retained then the permeability of the soil requires

careful investigation and more crucially the type of soil fill. In most cases the shear strength, moisture

content and liquid limit tests are necessity. In addition to the study on soil compaction an essential part of

the embankment design and the principles of compaction this will be briefly covered in this Chapter.

Seeing as Embankments are proving to be a cost-effective and natural meanings of shaping the

landscapes it has more of an interest to modern developments in many fields such as energy whistle

maintaining the aesthetic landscaping of most locations, it provides reservoir’s that are a means of water

storage that can be sustainable managed by towns and cities and provide above average strength for

development above it say a road or baggie jumping activities to mention a few. Owing to the speed at

which embankments are now being constructed this enables the use of dense fills to be formed at the

outset as would be obtained after years of settlement and use. Figure 31 is an example of a typical

embankment.

Figure 41: Typical Earth Dike with Drain

69


Soil Reinforcement Techniques

Soil Reinforcement by fibre materials

Natural fibres can be grouped into three classes.

Blast fibres: These are extracted from the stems of plants such as Jute, Flax, Hemp and Ramie. Flax,

Hemp and Ramie are usually used in twines, canvases and fishnets whistle Jute is used in sacking,

carpets, sandbags and concrete curing.

Leaf fibres: Sisal is a great example of a leaf fibre, mainly used as cordage or rope making material and

mats in many LEDC counties such as Kenya.

Seed/fruit fibres: Cotton and coir used in a similar manner to leaf fibres.

Natural fibres are considered to be an ancient cheap technique solution to soil reinforcements.

They are suitable for geotextiles as long as they give good mechanical properties and a reasonably

resistance to microbial attacks. Coir and Jutes fibres have been exploited as a reasonable suitable

geotextile in many Asian countries like India for example. (Babu, 2006) explains that materials such as

bamboo and timber can also be used as a natural reinforcement in many applications due to their high

tensile strength that typically ranges from 600-800kN/m² and gives a compressive strength ranging from

200-300kN/m². Straw was used by China and many African countries as random reinforcing materials,

mainly seen on mud huts and stone fencing without cementation. Organisations such as (IJMA) Indian

Jute Mills Association and (IJIRA) Indian Jute Industries’’ Research Association have good practice and

modern uses of these natural fibres even to date. Kolkata an Indian town have been using coir as a civil

engineering material, Rao and Balan (2000), prepared an excellent compilation that I will refer to for

more details. In the following sections a brief overview of the above materials and their properties will be

covered.

The main factors that make Natural Geotextiles useful

As far as reinforcement is concerned the tensile stress-strain behaviour of the geotextiles is a

major governing factor in making good estimation on the performance of the natural fibres. During the

installation process, the fibres are subjected to dynamic impact tests to enhance its properties such as

puncture resistance, tensile strength and shock resistance to sudden impacts. The geotextiles should also

be able to sustain a reasonable amount of erosion so as to make the reinforcement durable, adaptable and

cost effective to design and construct. The surface should be well covered with scopes for sufficient

passage of light for growth of vegetation. This property will help the material develop intimate contact

with the soil particles so that run-off doesn’t pass under it, hence further resisting the movement of soil,

seed and nutrients. Geotextile are water absorbents, this property helps to reduce the run-off velocity that

tends to wash away the top soil.

In filtration applications, the geotextile are placed between soft clays of very low permeability,

and free draining media like granular non-cohesive materials such as sands to act as a separator and filter.

The permeability of geotextiles must be significantly higher than the clayey soil, almost as high as a free

draining medium. This is to give it good horizontal and vertical drains, the geotextiles must be thick

enough to have a suitable in-plane permeability or transmissibility characteristics. (Babu, 2006) points out

some agricultural application of fibre pores, they must be sufficient for aeration and strength to withstand

handling stresses whistle maintaining a non-toxic level for the plants. They must conserve enough

moisture and heat for germination and protect seedlings from the scorching sun and lashing of rain during

70


early stages. All the above is to make any engineering use of the geotextile environmental friendly and

aesthetically pleasing to the eye.

Jute Fibres

Jute is one of the most valuable natural fibres produced mainly from India and Bangladesh. Under

the product name soil saver, it has been an export commodity for many years. More times than often it’s

referred to as the golden fibre of Bangladesh. The chemical compositions of Jute are 61% Alpha

cellulose, 24% Hemi cellulose (including uronic acid derivatives), 11.5% of Lignin and 3.5% of fatty

acids and waxy substances with some nitrogenous and mineral matter.

Jute-based geotextiles are evolved using a process of mixing a blend of yarns. Jute usually is a

woven product although in some cases a non-woven product is preferable. They give us a large variety of

range in use from good engineering properties that suit different requirements such as roads and paving to

the humble rubber in pencil cases. The roll –width of jute fabric is normally up to 3m long and length of

up to 100m. Geojute fabrics, jute meshes and jute Matts, nets, drains etc. are but an example to of the

types of products that can be produced from jute. It properties as shown in (table 1 and 2) below have an

engineering quality with many benefits.

Properties Range of Values

Fibre length (mm) 180-800

Fibre diameter (mm) 0.10-0.20

Specific gravity 1.02-1.04

Bulk density (kg/m3) 120-140

Ultimate tensile strength (N/mm2) 250-350

Modulus of elasticity (kN/mm2) 26-32

Elongation at break % 2-3

Water absorption 25-40

Properties Range of Values

Thread diameter (mm) 1.75-1.85

Mesh size (cm2) 3x3 (double)

Weight (g/m2) 680-750

Elongation at break (wet) % 15-20

Trapezoidal tear strength (N) 300-350

Permeability (cm/sec.)

-under unstressed conditions

-under all round pressure of 500kN/m2

>10−2

>10−3 − 10−4

Grab tensile strength (wet) N 800-900

Tensile strength (kN/m) 15-25

Puncture resistance (N) 350

CBR Pushing through resistance (N) 1300-3000

Cone drop penetration (mm) 8-10

(Aziz and Ramaswamy, 1991) brought out the advantages of using jute fabrics from a study of its

compaction characteristics, load-settlement relationships, unconfined strength tests, CBR tests, plate tests

and some giving some of the results in table 1.

Table 1: Typical properties of Jute (Aziz and Ramaswamy, 1991)

Table 2: Typical properties of Jute Fabrics

71


Coir Geotextiles

Coir is a natural product, which is available in most parts of the world more particularly India at

very low cost. Coir is attained from the husk of coconuts. Such Products have many geotechnical interests

and are manufactured from coir fibres. An excellent treatise on coir geotextiles can be found online

written by Rao and Balan (2000) for their Kerala State Coir Corporation Limited. It consists of the

methods of making coir geotextiles, its properties and application with a couple of case studies. For more

details the table below shows a chemical composition of coir fibre.

Table 1: Chemical composition of coir fibre

Lignin 41-45%

Cellulose 36-43%

Hemi-cellulose 0.15-0.25%

Pectin 4%

Water solubles 5.25%

Ash 2.22%

Advantages and disadvantages of coir

Advantages

The reinforcement product made of coir are comparatively stronger and stable than jute and other

materials

The initial strength and stiffness and hydraulic properties of coir reinforcement are almost

comparable to those of similar products made from polymer materials

When the requirement of reinforcement is for short duration, coir products can be a good

By chemical treatment and polymer coating the life of coir products ca be improved

Wide variety of geotextiles is economically viable and is environment-friendly for protecting

slopes in erosion prone areas

It takes 15 times longer than cotton and 7 times longer than jute to degrade

Very low raw materials price

No chemical, lubricants or emulsions are added at the time of spinning. Hence there are no toxic

elements on the yarns after it is spun and is safe for use.

Disadvantages

The coir products are vulnerable to different degradation action in the filed such as those induced

by installation damage, temperature fluctuations, chemical actions, hydrolysis, ultra-violet

degradation and other degradations

The reinforcement products of natural coir fibres develop their ultimate strength at higher strains

as the modulus is low. Hence their capacity is not fully utilized if there is a restriction on the

maximum deformations that structure can undergo. This can be overcome by coating them with

thermosetting polyester to increase their strength and stiffness.

The areas of application have been limited till date and are mainly in field of soil erosion control.

Coir fibres, extracted from dry coconuts husk exhibit a wide range of dimensions which makes

conventional spinning process difficult.

Bamboo and Timber Fibres

Strength and deformation characteristics of several natural materials such as bamboo, timber and fibres

are favourable to synthetics. It is a fact with regards to the deformation modulus and creep natural

72


materials such as bamboo/timber are superior to synthetics. Datye (1988) gives a good read on the use of

bamboo and timber as a more cost advantage over Indian Coir’s. Typical Properties of these materials are

as follows.

Table 2: Typical properties (Datye, 1988)

Timber (40mm diameter) Bamboo dowels

Weight (gm/m2) 1430 Allowable tensile

strength

80,000kPa

Strength at 2% strain

(kN/m)

15 Allowable compressive

strength

30,000kPa

Modulus at 2% strain

(kN/m)

1250

Advantages and disadvantages of Bamboo/Timber

Bamboo dowels can be efficiently used as reinforcement in preventing downhill movement of

slopes.

Cost advantages are significant

Materials characteristics are superior with respects to deformation modulus.

The small timber cribs are cost effective in comparison to concrete cribs.

By judicious design, a combination of small timber ties and crib facings could replace gabions.

The small-sized timber reinforcements are suitable for anchored construction.

The extensions are smaller and form of joints used impart adequate flexibility.

Timber reinforcement compare very effectively with some synthetic grids.

They can replace steel rods in short pile reinforcements of bases of embankments on soft soils.

The disadvantage is that they are prone to attack from microorganisms.

Combination of Geotextiles

Over the decades there have been some attempts to combine the relative advantages of coir, jute or

synthetic materials by mixing these materials in adjustable proportions.

Soil Reinforcement by Geosynthetics

Geosynthetics reinforced soil walls (RSWs) have been widely used throughout the world in road

embankment and retaining structures. RSWs have many advantages including aesthetics, short

construction period, good wall stability, cost effectiveness, good seismic response, strong adaptability on

soft highly compressible foundation soils and the ability to tolerate large differential settlement (Tatsuoka

et al., 1997; Bloomfield et al., 2001; Yoo and Jung,2004).

73


Chapter 5 Implementation of Soil Reinforcement

DIRECT SHEAR TEST

What is this test about

This test focuses on how effective randomly oriented discrete fibres on the mechanical response of fine

sand. Compaction and direct shear tests were performed on the sand specimens of different densities from

unreinforced and reinforced with fibres in different proportions. The presence of reinforcement gives rise

to an extra resistance to compaction, causing a less dense packing as the quantity of fibres increase the

peak shear strength and the strain required to reach the peak. The post-peak strength at large amounts of

strain was also higher when fibres were included. The observation of fibres leads to a more dilative

behaviour. Below you will notice a range of effective normal stresses employed, a linear failure envelope

was recorded for all densities and fibre concentrations. The increase of the peak shear strength was almost

linear for all densities at low effective normal stress and approaching the limiting value for the highest

normal stresses. For the loosest specimens reinforced with the highest amounts of fibres could be

employed in the lab using a moist tamping fabrication method, the relative increase of the peak shear can

be observed to be more than 50%.

It has been recorded that for the same confining stress, the strength of the reinforced sand increases by

reducing the average grain size 𝐷50 (Maher and Gray 1990, Gray and Al-Refeai 1986). Also a better

graduation- increase in co-efficient of uniformity, 𝑐𝑢 and a smaller 𝐷50 result in higher contribution to

strength. The mixing process between the soil, water and fibres was stopped by visual examination; it was

considered that the fibres are relatively well distributed throughout the soil mass. 3 layers of mixed soil of

different heights were used for the shear box specimens. Each layer was delicately deposited into the box

to ensure a zero drop height and minimal disturbance and each layer was compacted to the predetermined

height using a small drop light rectangular hammer. The presumably shear plane locates in the central part

of the middle layer. The specimen dimension were carefully measured and recorded in order to estimate

the initial voids ratio. This represented the intergranular voids ratio; the voids seemed to be occupied by

air, water and fibres.

Before starting the experiment it was considered necessary to estimate the maximum dosage of fibres that

can be mixed with a given amount of sand, placed into a given volume and compacted using moist light

tamping. This estimate was said to be no more than 10% of the volume of sand in the shear box. Four

different densities of sand have been chosen for this test. For each sand density, the moist sand was

divided in 3 equal layers into square moulds of 60x60mm. Each layer was compacted using the same

compaction device used for direct shear specimen fabrication but with a circular hammer. For a given

density the specimen was first fabricated with a fixed amount of bamboo fibres. Then the fibres dosage

was each time steadily increased until the limit beyond which was practically impossible to compact the

mixed soil any further.

74


Purpose: This test is performed to determine the consolidated-drained shear strength of a sandy to silty soil. The

shear strength is one of the most important engineering properties of a soil, because it is required

whenever a structure is dependent on the soil’s shearing resistance. Scour Failure is no different, using

soil reinforcement techniques to mitigate this can be seen in a shear test. Basically the adhesive properties

of the soil that is directly in contact with the water is the most vulnerable however a shear test will give us

the maximum horizontal displacement to which the most erosion is observed. Adding an additive like

fibres will see a huge increase in the strength of the soil properties gives it this ability to withstand high

amount of horizontal forces; which in-turn increases the soils ability to resist erosion. The shear strength

is needed for engineering situations such as determining the stability of slopes or cuts, finding the bearing

capacity for foundations, and calculating the pressure exerted by a soil on a retaining wall.

Standard Reference: ASTM D 3080 - Standard Test Method for Direct Shear Test of Soils Under Consolidated Drained

Conditions.

Apparatus: A Pair of swan-neck type yokes attached to the mid points of the upper frames side walls brings the point

of shear load transmission to the specimen centre. The shear load is applied in equal measure to each

wing through an arm. The transmission of the shear load from each arm to each wing is done through a

ball race. The load pad is clamped to the upper frame prior to the application of any shear displacement.

The apparatus test is 60x60x60mm specimens. Sheets of flexible neoprene membranes 1mm thick are

attached to the internal walls of the shear box (2sheet spanning the divide) to prevent excessive specimen

extrusion via the opening. Silicone grease was used to attach the shielding to the internal wall of the shear

box. The initial opening between the upper and lower frames of the shear box was installed prior to

specimen deposition.

Prior to the experiment a few series of direct shear tests were performed in order to access the ability of

the apparatus to reproduce test results for specimens sort of a control test on unreinforced sands.

75


Equipment: Direct shear device, Load and deformation dial gauges, Balance.

76


77


Test Procedure Weigh the initial mass of soil in the pan.

1. Measure the diameter and height of the shear box. Compute 15% of the diameter in millimetres.

2. Carefully assemble the shear box and place it in the direct shear device. Then place a porous

stone and a filter paper in the shear box.

3. Place the sand into the shear box and level off the top. Place a filter paper, a porous stone, and a

top plate (with ball) on top of the sand

4. Remove the large alignment screws from the shear box! Open the gap between the shear box

halves to approximately 0.025 in. using the gap screws, and then back out the gap screws.

5. Weigh the pan of soil again and compute the mass of soil used.

6. Complete the assembly of the direct shear device and initialize the three gauges (Horizontal

displacement gage, vertical displacement gage and shear load gage) to zero.

7. Set the vertical load (or pressure) to a predetermined value, and then close bleeder valve and

apply the load to the soil specimen by raising the toggle switch.

8. Start the motor with selected speed so that the rate of shearing is at a selected constant

rate, and take the horizontal displacement gauge, vertical.

9. Displacement gage and shear load gage readings. Record the readings on the data sheet. (Note:

Record the vertical displacement gage readings, if needed).

10. Continue taking readings until the horizontal shear load peaks and then falls, or the horizontal

displacement reaches 15% of the diameter.

78


Chapter 6 Results & Analysis

Date Tested: March 30, 2015 Tested By: Edison Derrick Mugoya Project Name: Soil Reinforcement Sample Number: Fibre 0.8, 0.9, 1.0 Visual Classification: Brown uniform sand

Shear Box Inside Diameter: 60mm

Area (A): 3600mm2

Shear Box Height: 60mm

Soil Volume: 216000mm3

Initial mass of soil and pan: 1000. g

Final mass of soil and pan: 720.82 g

Mass of soil: 279.18 g

Density of soil (?): 1.65 g/cm3

79


Direct Shear Test Data

Displacement rate: _______ Normal stress: 55.3kPa

Horizontal Dial Horizontal Load Dial Horizontal Shear

Reading Displacement Reading Shear Force Stress

(0.001 in) (in) (lb) (psi)

0 0 0 0 0

10 0.01 4 5.142 1.064

19 0.019 4.3 5.231 1.082

29 0.029 4.8 5.379 1.113

36 0.036 5 5.439 1.126

44 0.044 7 6.033 1.248

51 0.051 8 6.33 1.31

57 0.057 13.5 7.963 1.648

63 0.063 15 8.409 1.740

70 0.07 17 9.002 1.863

76 0.076 19 9.597 1.986

84 0.084 20 9.893 2.047

91 0.091 22 10.488 2.170

100 0.1 22.5 10.636 2.201

107 0.107 23 10.785 2.232

114 0.144 23.5 10.933 2.262

121.5 0.1215 25 11.379 2.355

129 0.129 25.5 11.527 2.385

137 0.137 26 11.675 2.416

145 0.145 27 11.973 2.478

152 0.152 27.5 12.121 2.508

160 0.16 28 12.270 2.539

179 0.179 25 11.379 2.355

80



Normal stress: 106.4kPa

Horizontal Horizontal Load Dial Horizontal Shear

Dial Reading Displacement Reading Shear Force Stress

(0.001 in) (in) (lb) (psi)

0 0 0 0 0

4.5 0.0045 8 6.330 1.31

11 0.011 12 7.517 1.556

17 0.017 13.5 7.963 1.648

23 0.023 15.5 8.557 1.77

30 0.030 16.5 8.854 1.832

37 0.037 18.5 9.448 1.955

44 0.044 20 9.894 2.047

50 0.05 23 10.785 2.232

56 0.056 25.5 11.527 2.385

62 0.062 29 12.567 2.60

70 0.07 31.5 13.309 2.754

77 0.077 33 13.755 2.846

82 0.082 36 14.646 3.031

88 0.088 39 15.537 3.215

94 0.094 42 16.428 3.4

101 0.101 44 17.022 3.522

108 0.108 48 18.210 3.768

115 0.115 49 18.507 3.83

121 0.121 54 19.991 4.13

127 0.127 56.5 20.734 4.291

136 0.136 57.5 21.031 4.352

141 0.141 60 21.774 4.506

148 0.148 61.5 22.219 4.599

155 0.155 62 22.368 4.62

81



Normal stress: 208.5kPa

Horizontal Horizontal Load dial Horizontal Shear

Dial Reading Displacement Reading Shear Force Stress

(0.001 in) (in) (lb) (psi)

0 0 0 0 0

1 0.001 16 8.706 1.801

5 0.005 22 10.488 2.170

10 0.01 27 11.972 2.478

15 0.015 31 13.16 2.723

21 0.021 34 14.052 2.908

28 0.028 36 14.646 3.031

34 0.034 41 16.131 3.338

39 0.039 41.5 16.279 3.37

42 0.042 43 16.725 3.461

51 0.051 45 17.319 3.584

61 0.061 47 17.913 3.707

68 0.068 50 18.804 3.891

74 0.074 54 19.99 4.13

82 0.082 56 20.586 4.26

88 0.088 58 21.18 4.383

94 0.094 61 22.071 4.568

101.5 0.1015 63 22.665 4.690

109 0.109 67 23.85 4.937

115 0.115 72 25.337 5.244

122 0.122 75 26.228 5.428

128 0.128 78 27.119 5.612

133 0.133 82 28.307 5.858

138 0.138 83 28.605 5.92

142 0.142 83 28.60 5.92

82


7

(ps

i)

6 5.9 psi

Str

es

s

5 4.7 psi

4

Sh

ea

r

3 2.6 psi

Ho

rizo

nta

l

2

Normal Stress =55.3kPa

1 Normal Stress =106.4kPa

Normal Stress =208.5kPa

0

0 0.02 0.04 0.06 0.08 0.1 0.12 0.14 0.16 0.18 0.2

Horizontal Displacement (in)

The table below shows the value of 𝜓𝑚𝑎𝑥 for all unreinforced specimens. For a constant effective

normal stress, the dilation increases with the density, whereas for a constant voids ratio, the tendency

for dilation is inhibited by a higher effective normal stress. The shear stress-horizontal displacement

curves above for the specimens at the same void ratio show that the reinforced sand exhibits a greater

peak shear stress than the unreinforced sand. The stress-strain relationship seems to be similar at low

horizontal displacement, but the reinforced soil is able to absorb more energy and therefore producing a

higher peak shear stresses. Increasing the amount of fibres leads to a larger horizontal displacement

being required to reach the peak shear failure. The residual strength is also increased with the fibre

concentration and reductions in the post peak shearing resistance systematically observed. AS Expected

a higher confining pressure gives higher shear strength and deformation values at failure; the bond

which develops between the fibres and the sand grains is enhanced.

For all the specimens densities considered in this experiment the shape of the volumetric curves of

reinforced sand taught us that an initial contraction must have taken place followed by dilation.

However the amount of vertical deformation was higher and the volumetric response became more

dilative with the presence of fibres and the fibre content. The addition of fibres to sand does not inhibit

the dilatancy, this behaviour can result in an increase in the stress increment required to initiate pore-

pressure build-up within the sand particles.

83


For similar densities and fibres concentrations the higher the effective normal stresses the small the

dilatancy. The maximum values of the dilation angle 𝜓𝑚𝑎𝑥(°) are shown in the table below. A

nonlinear variation of 𝜓𝑚𝑎𝑥 with the fibre concentration was observed.

Table 4: Maximum Dilation angle for all densities, stress and fibre conditions

Maximum Dilation

angle, 𝜓𝑚𝑎𝑥(°)

Voids Ratio

0.8 0.9 1.0

Fibre Content (%)

Normal Stress (kPa) 0 0.3 0.4 0 0.3 0.5 0.8 0 0.3 0.5 0.8

55.3 12.5 15.6 15.6 8.6 9.9 12.3 13 1.9 5.5 8.7 12.7

106.4 11.5 13.5 14.4 4.7 7.5 9.0 11.0 1.7 4.2 5.8 10.7

208.5 9.2 11.7 13.4 3.2 6.5 7.5 8.5 2.1 4.0 5.0 7.4

310.6 - 10.2 11.0 - 4.7 6.6 - - - - -

However for the limited range of effective normal stresses performed in this experiment an easier

quantification of the effects of fibres reinforcement can be noted to be a linear envelop found by linear

regression would be a reasonable approximation of the state of stress at failure. The extrapolation of

this line towards the zero effective normal stress gives a cohesion intercept. As a system of uncemted

particles, the sand should not have any true cohesion in terms of effective stresses. However, the inter-

particle interactions (interlocking) induced by the tamping fabrication technique seems to be important

and this can explain the existence of an apparent cohesion that increases slightly with the addition of

fibres. Although further analysis is required we can draw to a summarized conclusion, that the failure

envelopes are defined by the terms angle of friction φ° and cohesion intercept c. The shear strength

parameters obtained from direct shear tests on unreinforced and reinforced sand are also summed in the

table below.

Table 5: Angle of friction and Cohesion intercept for all series of tests

Voids

Ratio

Fibre Content (%)

0 0.3 0.4 0.5 0.8 1.0

C

(kPa) φ° C

(kPa) φ° C

(kPa) φ° C

(kPa) φ° C

(kPa) φ° C

(kPa) φ°

0.8 8.5 34.4 8.6 37.4 11.5 37.6 - - - - - -

0.9 4.5 31.4 8.9 33.6 - - 9.1 35.7 10.18 37.1 - -

1.0 0.8 30.9 5.1 32.8 - - 3.9 35.3 - - 7.0 39.4

84


The Peak shear strength variations with the amount of fibres are presented above. The values of the

peak shear stress can be calculated using the equations of failure lines derived in chapter 2 of this study.

The evolution of the peak shear strength is almost linear for all specimens’ densities at low effective

normal stresses. The slope of these lines increase with the effective normal stress. For all normal

stresses, much closer trends could be seen for specimen with 0.9 and 1.0 voids ratio. For higher

densities and normal stresses greater than 200kPa the evolution of the peak shear stress approaches a

limiting level. IT can be clearly observed that a fibre concentration of 0.8% gives practically the same

peak shear stress for specimens with 0.9 and 1.0 voids ratio. A similar case was seen in Gray and Al-

Refeai (1986), Rajan et al. (1996) and Murray et la. (2000) but for lower confirming stress levels.

85


Chapter 7 Conclusions and Recommendations

An experiment test was undertaken to investigate the individual effect of randomly placed bamboo

fibres on the mechanical behaviour of the sand. Compaction and direct shear test were conducted on

unreinforced and reinforced sand.

The results of the compaction tests indicated that the maximum dry density of reinforced sand decreases

with increasing fibres content. An optimum moisture content of 10%, independently of the amount of

fibres was noted.

The results of the direct shear box test of the randomly oriented fibres increased the failure peak shear

strength as well as the corresponding horizontal displacement. The specimens become more dilative

with the increasing amount of fibres, the dilation is not inhibited by the presence of fibres. This seems

to suggest that the use of mixed materials can improve the static liquefaction behaviour of loose sands.

The residual strength is also affected by the fibre content. The trend in change of the apparent cohesion

and angle of the shearing resistance seems to be consistent with the fibre inclusions. A Linear failure

line for reinforced specimens has been obtained for all densities and fibre concentration and this seems

to be in agreement with the post-test observations in our control. The failure seems to be due to the

slippage or pull out rather than the stretching or breaking of fibres.

The positive effect of the fibre reinforcement was clearly established during the analysis of the data

obtained from the testing. However the raw data that was obtained from the experimental test is not

accurate and extensive enough to establish the specific effect of the fibre on the soil. The time

constrains of the project and the lack of means of cleaning the flume; that would blocked, due to the

sand particles clogging the pump system sort of pushed my study to a simple testing method that is

adequate for obtaining a positive relationship between the shear strengthen and the it’s effect to the

addition of natural fibres. During the handling of the fibres we concluded the fibres are a very flexible

and strong material. I recommend testing the initial content of the fibres in order to increase the

precision in the measurements of the moisture content of the reinforced soils. It would also be

beneficial to observe the absorption capacity of the fibres and test any change in its stiffness and

strength.

Given more time further analysis including the an experimental test using the triaxial test would provide

a variation of horizontal and vertical compression stresses that would allow establishing the cohesion

and friction angles of each soil type. This test would also allow performing testing on un-drained and

drained soil condition and comparisons between the two conditions may draw a different conclusion.

86


References

Books

Purushothama, P.R. (2013) Soil Mechanics and Foundation Engineering. 2nd

edn. Pearson

Sivakumar Babu, G.L. (2006) An Introduction to Soil Reinforcement and Geosynthetics. University

Press (India) Private Limited

Craig, R.F. (2004) Craig’s Soil Mechanics. 7thedn. Spon Press

Reynolds, H.R. & Protopapadakis, P (1959) Practical Problems in soil mechanics. Crosby Lockwood

& Son, Limited

Barnes. G (2010) Soil Mechanics Principles and Practice. 3rd

edn Palgrave Macmillan

Journals

Chia-Nan, Liu, Kuo-Hsin Yang, Minh Duc Nguyen. (2014) ‘Behavior of geogrided reinforced sand

and effect of reinforcement anchorage in large-scale plane strain compression’, Geotextiles and

Geomembranes, 42(1), pp. 479-493

Jinchun Chai, Suksun Horpibulsuk, Shuilong Shen, John P. Carter. (2014) ‘Consolidation analysis of

clayey deposits under vacuum pressure with horizontal drains’, Geotextiles and Geomembranes,

42(1), pp. 437-444

M.R. Abdi, A.R. Zandieh. (2014) ‘Experimental and numerical analysis of large scale pull out tests

conducted on clays reinforced with geogrids encapsulated with coarse material’, Geotextiles and

Geomembranes, 42(1), pp. 494-504

R. Kerry Rowe, Lauren.E. Ashe, W. Andy Take, R.W.I. Brachman. (2014) ‘Factors affecting the

down-slope erosion of bentonite in a GCL’ Geotextiles and Geomembranes, 42(1), pp. 445-456

Michael Heibaum. (2014) ‘Geosynthetics for waterways and flood protection structures controlling

the interaction of water and soil’ Geotextiles and Geomembranes, 42(1), pp. 374-393

A. Bouazza, R.M. Singh, R.K. Rowe, F. Gassner. (2014) ‘Heat and moisture migration in a

geomembrane GCL composite liner subjected to high temperatures and low vertical stresses’

Geotextiles and Geomembranes, 42(1), pp. 555-563

Jian-Feng Xue, Jian-Feng Chen, Jun-Xiu Liu, Zhen-Ming Shi. (2014) ‘Instability of a geogrid

reinforced soil wall on thick soft Shanghai clay with prefabricated vertical drains: A case study’

Geotextiles and Geomembranes, 42(1), pp. 302-311

Moustafa I. Awad, Burak F. Tanyu. (2014) ‘Laboratory evaluation of governing mechanism of

frictionally connected MSEW face and implications on design’ Geotextiles and Geomembranes,

42(1), pp. 468-478

87


Hong-Hu Zhu, Cheng-Cheng Zhang, Chao-Sheng Tang, Bin Shi, Bao-Jun Wang. (2014) ‘Modeling

the pullout behavior of short fiber in reinforced soil’ Geotextiles and Geomembranes, 42(1), pp. 329-

388

S.G. Chung, H.J. Kweon, W.Y. Jang. (2014) ‘Observational method for field performance of

prefabricated vertical drains’ Geotextiles and Geomembranes, 42(1), pp. 405-416

Xiaobin Chen, Jiasheng Zhang, Zhiyong Li. (2014) ‘Shear behaviour of a geogrid-reinforced coarse-

grained soil based on’ Geotextiles and Geomembranes, 42(1), pp. 312-328

Briaud, J-L., Chen, H.-C., Li, Y., Nurtjahyo, P., and Wang, J., “SRICOS-EFA Method for Contraction

Scour in Fine-Grained Soils.” Journal of Geotechnical and Geoenvironmental Engineering, ASCE,

Vol. 131, No. 10 (2005) pp. 1283-1294.

Cardoso, A.H., and Bettess, R., “Effects of Time and Channel Geometry on Scour at Bridge

Abutments.” Journal of Hydraulic Engineering, ASCE, Vol. 125, No. 4 (1999) pp. 388-99.

Websites

Sturm, Terry W.; Ettema, Robert; Melville, Bruce W. (1990). Evaluation of Bridge-Scour Research -

Abutment and Contraction Scour Processes and Prediction: (NCHRP Document 181). Available:

https://app.knovel.com/web/view/swf/show.v/rcid:kpEBSRACS1/cid:kt00ABNWPA/viewerType:pdf

/root_slug:evaluation-bridge-scour?cid=kt00ABNWPA&page=5&b-toc-cid=kpEBSRACS1&b-toc-

root-slug=evaluation-bridge-. Last accessed 20/01/2015.



https://app.knovel.com/web/view/swf/show.v/rcid:kpEBSRACS1/cid:kt00ABNWY4/viewerType:pdf/

root_slug:evaluation-bridge-scour?cid=kt00ABNWY4&page=10&b-toc-cid=kpEBSRACS1&b-toc-

root-slug=evaluation-bridge-scour&b-toc-url-slug=abutment-form-construction&b-toc-

title=Evaluation%20of%20Bridge-Scour%20Research%20-

%20Abutment%20and%20Contraction%20Scour%20Processes%20and%20Prediction%3A%20(NC

HRP%20Document%20181). Last accessed 20/01/2015.



https://app.knovel.com/web/view/swf/show.v/rcid:kpEBSRACS1/cid:kt00ABNX21/viewerType:pdf/r

oot_slug:evaluation-bridge-scour?cid=kt00ABNX21&page=3&b-toc-cid=kpEBSRACS1&b-toc-root-

slug=evaluation-bridge-scour&b-toc-url-slug=abutment-scour-design&b-toc-






https://app.knovel.com/web/view/swf/show.v/rcid:kpEBSRACS1/cid:kt00ABNX6I/viewerType:pdf/r

oot_slug:evaluation-bridge-scour?cid=kt00ABNX6I&page=5&b-toc-cid=kpEBSRACS1&b-toc-root-

slug=evaluation-bridge-scour&b-toc-url-slug=scour-conditions&b-toc-


88




Stephen E. Coleman and Bruce W. Melville. (2001). Journal of Hydraulic Engineering. Available:

http://ascelibrary.org.ezproxy.brad.ac.uk/doi/pdf/10.1061/%28ASCE%290733-

9429%282001%29127%3A7%28535%29. Last accessed 19/02/2015.

Krishna Reddy. (2002). Engineering properties of soil Based on Lab testing. Available:

http://www.uic.edu/classes/cemm/cemmlab/Experiment%2012-Direct%20Shear.pdf.

Last accessed 17/03/15.

89


Appendix

Appendix A: Project Management and Organisation

This Section represents a progress check based on the plan of action. At the end of each week

I should have accomplished the named tasks; agreed upon with my supervisor Dr MHA Mohamed. A

Log Book is kept as a measure to the meetings attended with Mohamed and his contributions are

noted. This is based on my actual progress therefore it’s flexible to time constraints. It also gives me

ahead of schedule what to expect in the form of a tick box.

What needs to be done Date of

completion

Cover Page 3/10/14

Title Page

Project Aim & Objectives

Abstract

Table of Content 3/10/14

List of Figures

List of Tables

List of Abbreviations

Acknowledgements

Chapter 1 Introduction 3/10/14



10/11/14


Chapter 5 Analysis and Results



Stress Conditions for Failure

20/10/14

Shear Strength

27/10/14

Tensile Strength

Permeability and seepage

Bearing capacity

Lateral earth pressure

Consolidation Theory

Stability of Slopes

Soil Compaction

Mechanics of drainage

Shanghai’s June 27 2009 ‘Lotus riverside’ building collapse. Bridge project on RC

piers

After the Literature review in depth and research, what has been done about it??

Application approach/apparatus/ actually implemented idea that has been done in the

past or is the current method to solving the above issues!!!

90


Soil reinforcement techniques; materials/mechanical properties, geometry and

dimensions, assumptions/justifications/simplifications etc.

Labs, Prep, investigation, hypothesis, apparatus, results also found in appendix,

discussion of results, conclusion and evaluation of result.

Discussion of the whole project as a whole project as a whole, proposal/new

originality/unique/incorporated ideas of my own ideas.

Conclusion of Project, Recommendation for future work; given more time I

would…, with extent of my work I could…, I’d look forward to other methodologies

with further analysis and research...

References, Appendices; referred to for more details

Appendix B: Plan of Action

This Section represents the segments of work that need to be completed so as to make up the bulk of

the whole project. It’s a Gantts Chart on what need to be done and by when. This is part of phase 1 of

project strategy, allowing myself to monitor and keep a track of the goings of the project with direct

relation to the design plan.

Lit

era

ture

Rev

iew

What needs to be done When it should be done

by

Soil Characteristics and Properties 20/10/14

Stress Conditions for Failure 25/10/14

Shear Strength 27/10/14

Tensile Strength 27/10/14

Permeability and seepage 10/11/14

Bearing capacity 11/11/14

Lateral earth pressure 13/11/14

Consolidation Theory 15/11/14

Stability of Slopes 17/11/14

Soil Compaction 19/11/14

Mechanics of drainage 21/11/14

Case Study Shanghai’s June 27 2009 ‘Lotus riverside’ building

collapse. Bridge project on RC piers

10/11/14

Methodology After the Literature review in depth and research, what

has been done about it?? Application

approach/apparatus/ actually implemented idea that has

been done in the past or is the current method to

solving the above issues!!!

Soil reinforcement techniques; materials/mechanical

properties, geometry and dimensions,

assumptions/justifications/simplifications etc.

28/01/15

Analysis and

Results

Labs, Prep, investigation, hypothesis, apparatus, results

also found in appendix, discussion of results,

conclusion and evaluation of result.

12/02/15

Implementation of

Soil

Reinforcement

Discussion of the whole project as a whole project as a

whole, proposal/new originality/unique/incorporated

ideas of my own ideas.

12/03/15

Conclusions and

Recommendations

Conclusion of Project, Recommendation for future

work; given more time I would…, with extent of my

work I could…, I’d look forward to other

methodologies with further analysis and research...

References, Appendices; referred to for more details

02/04/15

91


Appendix C: Gantts Chart

This section is used to point out and identify the main objectives with respects to time intervals on the general progress of the final year project. These

figures are used as approximations to encourage decision based from the plan of action. It describes the activities that should be undertaken. On the left of the

chart is a list of the activities and along the top is a suitable time scale. Each activity is represented by a bar; the position and length of the bar reflects the start

date, duration and end date of the activity. This allows me to

see at a glance:

What the various activities are

When each activity begins and ends

How long each activity is scheduled to last

Where activities overlap with other activities, and by

how much

The start and end date of the whole project

Table 1 as shown gives a layout of the projects main tasks

and an approximation of the duration required to finish the task

in good time before the handover date 23th April 2015. This

has helped me in my scheduling and organisation of the project

and it welfare.

Final Year Project Tasks Start Date Duration

(days)

End date

Planning Scope and research 29/09/2014 4 03/10/2014

Project Aims and Objectives 03/10/2014 31 03/11/2014

Layout design 08/10/2014 5 13/10/2014

Chapter 1 Introduction 13/10/2014 2 15/10/2014

Chapter 2 Literature Review 13/10/2014 39 21/11/2014

Chapter 3 Case Study 15/10/2014 5 20/10/2014

Chapter 4 Methodology 27/10/2014 93 28/01/2015

Chapter 5 Analysis and Results 24/11/2014 80 12/02/2015

Chapter 6 Implementation of Soil Reinforcement 08/12/2014 94 12/03/2015

Chapter 7 Conclusions and Recommendations 29/12/2014 94 02/04/2015

Type collected data and proof reading 02/02/2015 73 16/04/2015

Print all documents 18/04/2015 2 20/04/2015

Hand in Project 20/04/2015 3 23/04/2015

Table 1: List of activities with their start and finish time intervals

92


29 S

epte

mb

er 2

014

06 O

ctob

er 2

014

13 O

ctob

er 2

014

20 O

ctob

er 2

014

27 O

ctob

er 2

014

03 N

ov

ember

20

14

10 N

ov

ember

20

14

17 N

ov

ember

20

14

24 N

ov

ember

20

14

01 D

ecem

ber

20

14

08 D

ecem

ber

20

14

15 D

ecem

ber

20

14

22 D

ecem

ber

20

14

29 D

ecem

ber

20

14

05 J

anu

ary 2

015

12 J

anu

ary 2

015

19 J

anu

ary 2

015

26 J

anu

ary 2

015

02 F

ebru

ary

201

5

09 F

ebru

ary

201

5

16 F

ebru

ary

201

5

23 F

ebru

ary

201

5

02 M

arch

20

15

09 M

arch

20

15

16 M

arch

20

15

23 M

arch

20

15

30 M

arch

20

15

06 A

pri

l 2

01

5

13 A

pri

l 2

01

5

20 A

pri

l 2

01

5

27 A

pri

l 2

01

5

04 M

ay 2

01

5

Planning Scope and research

Project Aims and Objectives

Layout design

Chapter 1 Introduction




Chapter 5 Analysis and Results



Type collected data and proof reading

Print all documents

Hand in Project

Figure 1: Gantts Chart on the Final year projects

Tasks and Activites

93


Appendix D: Mind Map

Engineering Techniques

Use of fibres mixes both

natural and artificial

Materials required for

maxima strength in design

Principles and influencing

factors affecting design

Design

Geotextiles

Design of grid shapes and

roughness

Design factors governing the

usage of natural geotextiles

Durable, reliant and

economical

Cost of production

Are these methodologies

readily available

Bearing capacity of soil

Soils Stability under lateral

earth pressure

Seismic activity and its effect

on reinforced soil masses.

Coastal erosion to reinforced

soils

Properties

Soil properties

Shear strength tests

Triaxial test

Consolidation analysis on

various soils such as cohesive

and cohesionless

Improvements on slopes

Experiment tests and analysis

on various soil to study it’s

behaviour

Stress conditions and failure

Soil types

Land Fills

Slope stability

Retaining walls

Embankments

Earth Dams & estuaries

How well construction of

grids is managed!

Drainage

Geosynthetics, Materials

available, properties of mix in

formation of grid

Soil Nailing and Anchoring

Application

Relationship of design and

real life

Integrated methods of fibres

and grids

Methods currently being used

in industry

Enhanced adjustment to

improve efficiency

Filtrations capabilities and

speed effects on reinforced

soils

Understanding the natural

Soil

Reinforcements

94


Appendix E:

INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE

Engineering

Transcript of INGENUITY IN SOIL REINFORCEMENT METHODS AGAINST SCOUR FAILURE