Geotechnical Evaluation Replacement Namaqua Bridge Over the … · 2018-11-28 · Typical Heel...
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Geotechnical Evaluation
Replacement Namaqua Bridge
Over the Big Thompson River
Loveland, Colorado
Revised 4
Prepared for:
Mr. Mike Oberlander
Interwest Consulting Group
1218 West Ash, Suite A
Windsor, Colorado 80550
Job Number: 16-3048 January 4, 2017
Namaqua Bridge
TABLE OF CONTENTS
Page
Purpose and Scope of Study ...................................................................................... 1
Proposed Construction ................................................................................................ 1
Site Conditions ............................................................................................................ 2
Geologic Setting ........................................................................................................... 3
Subsurface Exploration ............................................................................................... 4
Laboratory Testing ...................................................................................................... 4
Subsurface Conditions ................................................................................................ 5
Seismic Design Parameters ......................................................................................... 6
Geotechnical Consideration for Design .......................................................................... 7
Deep Foundations ........................................................................................................ 8
Shallow Foundations .................................................................................................. 17
Abutment and Wing Walls ............................................................................................ 19
Abutment Approaches .................................................................................................. 21
Lateral Loads ............................................................................................................... 21
Water-Soluble Sulfates ................................................................................................ 25
Soil Corrosivity ............................................................................................................. 27
Project Earthwork ....................................................................................................... 28
Excavation Considerations .......................................................................................... 32
Pavement Sections ....................................................................................................... 34
Closure ........................................................................................................................ 40
Locations of Test Holes ..................................................................................... Figure 1
Logs of the Test Holes ............................................................................... Figures 2 & 3
Legend and Notes ................................................................................................ Figure 4
Gradation Plots .......................................................................................... Figures 5 – 7
Direct Shear Data ................................................................................................ Figure 8
Seismic Response Spectrum ............................................................................. Figure 9
Capacity Reduction Factor Plots ........................................................... Figures 10 & 11
Typical Heel Drain Detail ................................................................................... Figure 12
Summary of Laboratory Test Results ........................................................ Tables 1 & 2
Core Hole Logs ............................................................................................. Appendix A
Pavement Section Calculations .................................................................... Appendix B
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Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 1
PURPOSED AND SCOPE OF STUDY
This report presents the results of a geotechnical evaluation performed by GROUND
Engineering Consultants, Inc. (GROUND), in support of design and construction of a
replacement bridge to carry Namaqua Avenue over the Big Thompson River near
Loveland, Colorado. Our study was conducted in general accordance with GROUND’s
Proposal No. 1511-2098 and associated contract agreement between Interwest
Consulting Group and GROUND, dated August 2, 2016.
A field exploration program was conducted to obtain information on subsurface
conditions. Material samples obtained during the subsurface exploration were tested in
the laboratory to provide data on the classification and engineering characteristics of the
on-site soils. The results of the field and laboratory studies are presented herein.
This report has been prepared to summarize the data obtained and to present our
conclusions based on the proposed construction and the subsurface conditions
encountered. Geotechnical design parameters and a discussion of geotechnical
engineering considerations related to the construction of the bridge also are included.
PROPOSED CONSTRUCTION
We understand that present plans call for either a 155-foot long, single-span bridge
structure, or a longer 3-span bridge structure that will carry Namaqua Avenue over the
Big Thompson River. We anticipate that that the bridge abutments likely will include
abutment and/or wing walls. Additionally 300 to 400 feet of Namaqua Avenue south of
the bridge and about 1,500 feet of Namaqua Avenue north of the bridge will be
reconstructed and widened to include sidewalks and bicycle trails. We also understand
that foundation design for the bridge will be following the current edition of AASHTO of
Load Resistance Factor Design (LRFD) methodology and that pavement design will be
developed using AASHTO (DARWin®) and Larimer County methodologies. If the use of
the current CDOT ME methodology is required, GROUND should be notified so that
appropriate revision can be made.
If the proposed construction differs significantly from that described above, GROUND
should be notified to re-evaluate our conclusions.
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SITE CONDITIONS
The proposed bridge will span the Big Thompson River approximately 0.3 miles north of
Crestone Drive and approximately 0.5 miles south of US Highway 34. The surrounding
land was largely in use as farmland with associated single-family residences and farming
buildings (barns, sheds, etc.) or in use as sand and gravel pits for a concrete and
aggregate company. Namaqua Park was also observed to the southeast of the existing
bridge. Additionally, single-family residences and an elementary school were observed
in the greater project area.
The existing Namaqua Avenue was a 2-lane, asphalt-paved road and was aligned,
generally, north to south. Minor to moderate distress was observed in the existing
asphalt pavements (limited alligator cracking, local rutting, and minor transverse
cracking). The road grade elevation relative to the surrounding land near the crossing
varied locally and the road grade was raised between approximately 2 to 10 feet. A
retaining structure was observed along the northwest embankment of the road near the
existing bridge.
The Big Thompson River flowed to the east. The stream banks were relatively gentle on
the west (upstream) side of the bridge and moderately to relatively steep slopes on the
east (downstream) side of the bridge. These slopes displayed between approximately 3
feet (upstream side) and approximately 10 feet (downstream side) of relief. The
surrounding topography was generally rolling and sloped gently to the east near the
project site. Grasses and other relatively small native vegetation along with large,
mature trees were observed growing on the road shoulder and in the greater project
area.
Rip rap up to at least 2½ feet in dimension were noted along the existing bridge
abutments and bent, and below the flow-control structure on the downstream side of the
bridge. Rocks up to about 8 inches in dimension were noted among the native deposits.
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GEOLOGIC SETTING
Published geologic maps, such as Colton (1978)1 depict the site as underlain by the
upper Holocene Post-Piney Creek Alluvium (Qpp) and the Pleistocene Broadway
Alluvium. In the project area, alluvium typically consists of sands, gravels, and boulders,
with silt and clay deposits locally. These surficial deposits are mapped as underlain by
the upper Cretaceous Lower Shale Member of the Pierre Shale Formation (Kpl). In the
project area, the Lower Shale Member of the Pierre Shale consists of the Mitten Black
Shale member, Sharon Springs Member, and Gammon Ferruginous which are largely
made up of dark olive-gray bentonitic shale. The shales typically are moderately to
highly expansive and are often very hard, well-cemented, and difficult to excavate.
A portion of the Colton (1978)1 geologic map is reproduced below.
1 Colton, R.B., 1978, Geologic map of the Boulder-Fort Collins-Greeley area, Front Range Urban Corridor,
Colorado: U.S. Geological Survey, Miscellaneous Investigations Series Map I-855-G, scale 1:100,000.
Approximate Project Alignment
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SUBSURFACE EXPLORATION
Subsurface exploration for the project was conducted in August 2016. Two (2) test
holes were drilled at locations on either side of the bridge with a conventional, truck-
mounted, drilling rig to evaluate the subsurface conditions as well as to retrieve samples
for laboratory testing and analysis. The test holes were advanced to depths of about 41
to 50 feet below existing grades. A GROUND engineer directed the subsurface
exploration, logged the test holes in the field, and prepared the samples for transport to
our laboratory.
Samples of the subsurface materials were retrieved with a 2-inch I.D. California liner
sampler and a 1⅜-inch I.D. Standard Penetration Test sampler. The samplers were
driven into the substrata with blows from a 140-pound hammer falling 30 inches, in
general accordance with (in the case of the 1⅜-inch sampler) the Standard Penetration
Test described by ASTM Method D1586. Penetration resistance values, when properly
evaluated, indicate the relative density or consistency of soils. Depths at which the
samples were obtained and associated penetration resistance values are shown on the
test hole logs.
The approximate locations of the test holes relative to the existing Namaqua Avenue
bridge are shown in Figure 1. Logs of the test holes are presented in Figures 2 and 3,
and explanatory notes and a legend for the logs are provided in Figure 4. Logs of the
core runs are presented in Appendix A.
LABORATORY TESTING
Samples retrieved from our test holes were examined and visually classified in the
laboratory by the project engineer. Laboratory testing of soil samples included standard
property tests, such as natural moisture contents, dry unit weights, grain size analyses,
and Atterberg limits. Unconfined compressive strength, direct shear, water-soluble
sulfate content, swell-consolidation and a suite of corrosivity tests were completed on
selected samples, as well. Resilient modulus testing also was performed on a
composite sample obtained from the pavement test holes. Laboratory tests were
performed in general accordance with applicable ASTM and AASHTO protocols.
Results of the laboratory testing program are summarized in Tables 1 and 2 and on the
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test hole logs. Gradation plots are provided in Figures 5 through 7 and may be used for
scour analysis. Direct Shear data is presented in Figure 8.
SUBSURFACE CONDITIONS
The test holes penetrated approximately 6 to 9 inches of asphalt before penetrating
sand, silt, and clay fill materials to depths between 4 and 11 feet below existing grade.
Additionally, a base course like material, approximately 12 inches in thickness was
encountered underlying the asphalt in Test Holes 1, 2, and P-1. Native sands and
gravels where encountered underling the fill materials that extended to depths of about
22 to 26 feet. These soils were underlain by clay shale bedrock which was encountered
at depths between 31 and 33 feet.
We interpret the native sands and gravels to be alluvial (stream-laid) deposits and the
bedrock clay shales to be Pierre Shale materials.
Fill soils were recognized in the test holes, and likely are present elsewhere near the
bridge site. Delineation of the complete lateral and vertical extents of fills at the site, or
their compositions, was beyond our present scope of services. If fill soil volumes and
compositions at the site are of significance, they should be evaluated using test pits.
Additionally, given the coarse nature of the materials encountered, it is difficult to readily
differentiate between coarse fill materials and base course materials. Section of base
course may be present beneath pavements where they were not indicated on the drill
logs.
Coarse gravel and boulders, as well as similarly sized fragments of debris are not
represented well in samples obtained from small diameter test holes. Therefore, such
coarse materials may be present, even where not included in the descriptions herein.
Given the general proximity of the site to the mountain front, the presence of boulders in
the subsurface in the deposits along river must be considered likely.
Fill fine to coarse sands and silts with local clays. They were dry to moist, non- to
slightly plastic, soft to very stiff or loose to dense, and brown dark brown to red brown in
color.
Sands and Gravels ranged from clean to clayey or silty, fine to coarse sands to that
included local boulders, and clay and silt beds. They were slightly moist to wet, non- to
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slightly plastic, medium dense to very dense, and light brown to brown to gray brown in
color.
Clay Shale consisted of interbedded, non-durable, silt and clay shales. They were
slightly moist to moist, slightly to moderately plastic, very hard and dark gray to black in
color.
Groundwater was encountered in the test holes at depths of about 12 to 20 feet below
existing grade at the time of drilling. The test holes were immediately backfilled after
drilling operations due to safety concerns. Groundwater levels can be expected to
fluctuate, however, in response to annual and longer-term cycles of precipitation,
irrigation, surface drainage, land use, and the development of transient, perched water
conditions. We anticipate that water levels at the bridge abutments will tend to correlate
to water levels in the Big Thomson River, however.
SEISMIC DESIGN PARAMETERS
Based on extrapolation of available data to depth and our experience in the project area,
we consider the site likely to meet the criteria for a Seismic Site Classification of D
according to the 2014 AASHTO classification (Table 3.10.3.1-1). To evaluate the
Seismic Site Classification quantitatively, seismic shear wave velocity testing and/or
exploration to depths of at least 100 feet should be performed. We consider the
likelihood of achieving a Site Classification of C at the subject location to be moderate to
high. Seismic parameters that are applicable to this bridge site based on a Site
Classification of D are listed below:
Peak Ground Acceleration (PGA): 0.060 g
Short Period Spectral Acceleration (SS): 0.125 g
Long Period Spectral Acceleration (Sl): 0.032 g
FPGA: 1.6
Fa: 1.6
FV: 2.4
SDS: 0.200 g
SDl: 0.077 g
AASHTO Seismic Category: 1
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A seismic response spectrum for the site, based on these parameters is provided in
Figure 9.
GEOTECHNICAL CONSIDERATIONS FOR DESIGN
Typically in the project area, heavy structures such as bridges are supported on deep
foundations bearing in the bedrock underlying the site. Pierre Shale Formation clay and
shales were encountered in foundation test holes at depths of about 22 to 26 feet
beneath the overlying alluvial and fill soils. Therefore, site appears suitable for
supporting the new bridge on driven pile foundations driven to refusal in the relatively un-
weathered bedrock. Anticipated pile tip depths are discussed below in the Deep
Foundations section of this report. Drilled piers also are feasible, but likely will be more
difficult to install.
We estimate that likely, post-construction movements of properly installed deep
foundations at the site will be about ½ inch.
Additionally, a Geosynthetic Reinforced Soil – Integrated Bridge System (GRS-IBS) may
be a cost effective alternative to driven piles. A GRS system is created by combining
geosynthetic reinforcement and a high quality granular fill to form a composite material
that is capable of carrying high bridge loads. This composite material is used in
conjunction with a GRS abutment and integrated bridge approaches to create the GRS
system. Likely post-construction movements for these systems are typically higher than
deep foundation systems, but a GRS-IBS may be easier and more cost effective to
construct. If GRS-IBS is selected, GROUND can provide design services for the
abutment and wing walls.
To limit differential movements relative to the bridge structure, abutment and wing walls
should be founded in a similar manner to the bridge. However, we understand that
shallow foundations are being considered for the abutment and wing walls.
Geotechnical parameters for design of shallow foundations for abutment and wing walls
are provided in the Shallow Foundations section of this report. We estimate post-
construction movements will be on the order of 1 inch for abutment and wing walls
supported on properly constructed shallow foundations.
Regardless of the selected foundation type, bridge elements should be designed to
account for the design depths of scour and associated loss of support.
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DEEP FOUNDATIONS
Both driven steel H-piles and drilled piers, bearing in the bedrock underlying the
alignment appear suitable, in principle, to support the proposed bridge. We anticipate
that drilled pier may entail significantly greater difficulties to install. Geotechnical
parameters for both types of foundations are provided below.
The geotechnical parameters provided in this report are discussed in terms of the ‘North
Abutment’ and ‘South Abutment.’ If the 3-span design is selected, the parameters for
the North Abutment also are applicable to the northern bent and the parameters for the
Sorth Abutment also are applicable to the southern bent.
Note that the pile and drilled pier depths indicated herein refer to depths below existing
grades at the test hole locations at the time of drilling. The contractor should make
allowance for any grade changes between that time and the time of construction.
Lowering grades, however, may not be sufficient to result in shortening of piles,
however. We have assumed that pile caps / top-of-pile will be at depth of approximately
12 below existing grades.
Driven Pile Foundations Geotechnical criteria based on the AASHTO LRFD
methodology are provided below for design of driven, steel, H-pile foundations.
Driven piles should be installed per Section 502 of the current edition/revision of the
CDOT Standard Specifications for Road and Bridge Construction. The installation
considerations should be taken into account when preparing project documents.
Geotechnical Parameters for H-Pile Foundation Design
1) Driven steel H-piles to support the bridge and walls should be Grade 50, high
strength steel (ASTM A572) with a minimum yield strength of 50 ksi.
2) The piles should consist of heavy steel H-sections. We anticipate that HP 12x53,
HP 12x74, HP 14x73, or HP 14x89 sections will be used. Selection of pile size
should consider anticipated pile depths.
3) The piles should be driven to practical refusal, bearing in relatively un-weathered
bedrock underlying the site. The depths to which we anticipate that piles will be
driven are tabulated below:
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ANTICIPATED PILE TIP ELEVATIONS
Location Test
Holes
Approximate
Top-of-Bedrock Depth
Anticipated Pile Tip
Depth for Moderately Loaded Piles
Anticipated Pile Tip
Depth for Heavily Loaded Piles
(feet below existing grade*)
(feet below existing grade*)
(feet below existing grade*)
Southern Abutment TH 3 & 4 23 ± 28 to 34 32 to 36
Northern Abutment TH 1 & 2 26 ± 31 to 37 35 to 39
*existing grade at time of drilling operations
Because of the potential variability depth to top-of-bedrock, the contractor should
be prepared to advance the piles to elevations at least 5 feet lower than the
depths tabulated above. A test pile program with dynamic pile testing (PDA)
would be beneficial for refining anticipated driving depths/elevations. At least one
test pile should be driven at each abutment.
A minimum pile penetration may need to be specified to meet the requirements
for uplift resistance, lateral resistance, and/or the depth to resist scour.
The structural engineer and contractor also should anticipate refusal locally
during pile installation. The contractor should be prepared to install additional
piles offset from refusal locations, as necessary.
4) Drivability analysis using a wave equation analysis should be performed to
establish preliminary installation criteria for driven piles.
5) Based on the structural strength limit (AASHTO Article 6.5.4) and drivability
analysis (AASHTO Article 10.7.8) for H-piles, the maximum driving stress should
not exceed the nominal structural strength of 36 ksi.
6) In order to allow pile dynamic testing (See ‘Installation Considerations,’ below.)
welded pile splices should be specified, and compression splices not allowed.
7) Geotechnical parameters for design of piles to resist lateral loads are provided in
the Lateral Loads section of this report.
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8) Lateral resistance to horizontal forces also can be resisted by battered piles. It is
normal to assume a battered pile can resist the same axial load as a vertical pile
of the same type and size driven to the same depth. The vertical and horizontal
components of the load will depend on the batter inclinations.
Pile batters should not exceed 1:4 (horizontal : vertical).
9) For static analysis, based on the available geotechnical data and AASHTO Table
10.5.5.2.3-1, the nominal resistance values for axial compression and resistance
factors (φstat) tabulated below for each abutment may be used for pile design.
* Side resistance should be neglected for all material above the depth of design scour.
A higher soil resistance factor may be used in the pile foundation design, if the
driving criteria are established by static and/or dynamic testing. Provided that the
dynamic testing with signal matching per CDOT Section 502.05 is performed on
at least 1 pile at each abutment, then the nominal resistance values and dynamic
soil resistance factors (φdyn) tabulated below for each abutment may be used for
pile design:
Material
Depth Range
(feet)
North South
Loading
Type
Resistance
Type
Nominal
Resistance
Resistance
Factor
(φstat)
Fill 3* – 11 3* – 11 Axial Side 0.40 ksf 0.45
Uplift Side 0.40 ksf 0.35
Sands & Gravels 11* – 15 11* – 15 Axial Side 1.7 ksf 0.45
Uplift Side 1.7 ksf 0.35
Sands & Gravels 15* – 26 15* – 23 Axial Side 4.2 ksf 0.45
Uplift Side 4.2 ksf 0.35
Shale Bedrock
26+
23+ Axial
Side 10 ksf 0.35
Tip 91 ksf 0.40
Uplift Side 10 ksf 0.25
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* Side resistance should be neglected for all material above the depth of design scour.
For H-piles, the soil contact area should be applied for the skin resistance and
the box / square perimeter area should be applied for the tip resistance.
10) Groups of closely spaced piles will require an appropriate reduction of the
resistance values tabulated above. Reduction of axial capacities can be avoided
by spacing piles apart a distance of at least 3 ‘diameters’ center to center.
Reduction factors for piles spaced more closely than 3 diameters center to center
can be obtained from Figure 10.
11) Reduction of lateral capacity generally can be avoided by spacing piers at least 3
diameters apart, center to center. However, linear arrays of piles loaded in line
with the pile centers should be spaced at least 8 ‘diameters’ apart to avoid
reduction in axial capacity. Reduction factors for piles spaced more closely than
8 diameters can be obtained from Figure 11.
Installation Considerations
12) We suggest that a test pile installation program be performed to define better the
driving conditions, appropriate pile hammer energy, installation depth, and
refusal conditions. Due to the overburden soils being relatively thick at the site,
Material
Depth Range
(feet)
North South
Loading
Type
Resistance
Type
Nominal
Resistance
Resistance
Factor
(φdyn)
Fill 3* – 11 3* – 11 Axial Side 0.40 ksf 0.65
Uplift Side 0.40 ksf 0.35
Sands & Gravels 11* – 15 11* – 15 Axial Side 1.7 ksf 0.65
Uplift Side 1.7 ksf 0.35
Sands & Gravels 15* – 26 15* – 23 Axial Side 4.2 ksf 0.65
Uplift Side 4.2 ksf 0.35
Shale Bedrock 26+ 23+ Axial Side 10 ksf 0.65
Tip 91 ksf 0.65
Uplift Side 10 ksf 0.25
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the wave equation analysis list in Item 4, above, may not always be able to
model the refusal conditions properly.
13) It may be beneficial to reinforce the piles with a commercial, heavy duty, pile tip.
14) The pile-driving hammer should develop a minimum of 26,000 foot-pounds of
energy per blow for HP 12x53 piles, 42,000 foot-pounds for HP 12x74 piles,
40,000 foot-pounds for HP 14x73 piles, and 52,000 foot-pounds for HP 14x89
piles. However, the hammer generally should not develop more than 2,500 foot-
pounds per square inch of cross sectional area (e.g., 38,750 foot-pounds for HP
12x53 piles) unless it is demonstrated by wave equation analysis that the piles
can be installed safely and efficiently.
15) After the actual pile type and proposed hammer have been selected, a
geotechnical engineer should be retained to perform a Wave Equation Analysis
to determine if the driving hammer is sized adequately for the type of pile
selected and the soils and bedrock materials into which the piles will be driven, if
a test pile program will not be performed.
16) Consideration also should be given to re-striking piles to evaluate their capacity
at least 24 hours after initial driving.
17) Pile dynamic testing should be performed using a pile driving analyzer (PDA) and
signal matching at the start of pile installation to:
Assess the condition of the pile,
Evaluate the efficiency of the hammer,
Measure the driving stress in the pile,
Determine the static capacity of the pile, and
Establish the pile driving criteria for required static capacity, or sands and gravels / bedrock penetration.
18) It may be beneficial also to perform lateral loading testing at the start of pile
installation.
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19) Where a pile cannot be advanced to at least the approximate, anticipated tip
elevation, it should be evaluated with regard to its capacity by the geotechnical
engineer and the structural engineer.
Drilled Pier Foundations Geotechnical criteria based on the AASHTO LRFD
methodology are provided below for design of straight-shaft, drilled pier foundations.
Geotechnical Parameters for Drilled Pier Design
1) Drilled piers should bear in relatively un-weathered bedrock underlying the site.
For design purposes, ‘relatively un-weathered’ bedrock may be taken to be at
and below depths of 23 feet below existing grades at the north abutment and at
and below 26 feet at the south abutment. For bidding purposes this may vary
somewhat at each abutment.
2) Drilled piers should be at least 24 inches in diameter and generally should be
designed for a maximum length to diameter ratio of 30:1. The actual pier
diameters and length to diameter ratios should be determined by the structural
engineer, however.
3) Drilled piers should have a minimum length of 29 feet at the north abutment and
32 feet at the south abutment. The actual pier lengths should be determined by
the structural engineer based on loading, etc., with further increases in length
possibly required by the conditions encountered during installation at each pier
location.
4) Piers also should penetrate at least 6 feet into relatively un-weathered bedrock
or 3 pier diameters, whichever is greater.
Based on a the minimum length and bedrock penetration, and taking top of
competent bedrock to be 23 feet (north abutment) or 26 feet (south abutment)
below grade, pier lengths of 31 to 34 feet at the north abutment or 34 to 37 at the
south abutment are anticipated to meet the geotechnical criteria. Actual pier
lengths commonly will be greater due to structural considerations, conditions in
the pier holes, etc.
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5) Drilled piers may be designed using the nominal resistance values and
resistance factors tabulated below for each abutment.
Material
Depth Range
(feet)
North South
Loading
Type
Resistance
Type
Nominal
Resistance
Resistance
Factor
Fill 5* – 11 5* – 11 Axial Side 0.35 ksf ** 0.55
Uplift Side 0.35 ksf ** 0.45
Sands & Gravels 11* – 15 11* – 15 Axial Side 0.95 ksf ** 0.55
Uplift Side 0.95 ksf ** 0.45
Sands & Gravels 15* – 26 15* – 23 Axial Side 1.15 ksf ** 0.55
Uplift Side 1.15 ksf ** 0.45
Shale Bedrock
26+
23+ Axial
Side 11 ksf 0.45
Tip 120 ksf 0.50
Uplift Side 11 ksf 0.30
* Side resistance should be neglected for all material above the depth of design scour. In addition, AASHTO indicates that the uppermost 5 feet should be neglected for calculating side resistance.
** Assumed to be cased during installation.
6) Estimated settlement of properly constructed drilled piers will be low, on the order
of ½ inch, to mobilize side resistance. Settlement of groups of piers spaced
more closely than 3 diameters, center to center, may be larger and should be
studied under an individual basis.
7) Piers should be reinforced adequately, as determined by the structural engineer.
8) Geotechnical parameters for resisting lateral loading of piers are provided in the
Lateral Loads section of this report.
9) Rock penetration in pier holes should be roughened artificially to assist the
development of peripheral shear between the pier and bedrock. Artificially
roughening of pier holes should consist of installing shear rings 3 inches high
and 2 inches deep in the portion of each pier penetrating bedrock below depths
of 23 feet at the north abutment or 26 feet at the south abutment. The shear
rings should be installed on 18-inch centers.
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The specifications should allow a geotechnical engineer to waive the requirement
for shear rings depending on the conditions actually encountered in individual
pier holes, however.
10) Groups of closely spaced drilled piers will require an appropriate reduction of the
resistance values tabulated above. Reduction of resistance values can be
avoided by spacing piers at least 3 diameters apart, center to center. Reduction
factors for the resistance values for piers spaced more closely than 3 diameters
center to center can be obtained from Figure 10.
11) Reduction of lateral capacity generally can be avoided by spacing piers at least 3
diameters apart, center to center. However, linear arrays of drilled piers loaded
in line with the array (parallel to the line connecting the pier centers) must be
spaced at least 8 diameters apart, center to center, to avoid reductions in lateral
capacity. Reduction factors for lateral capacities for closely spaced drilled piers
can be obtained from Figure 11.
Drilled Pier Construction Considerations
12) We anticipate that due to the relatively permeable nature of the fill and native
sands overlying the bedrock at the site, and the presence of coarse gravel and
boulders in these materials associated with the Big Thompson River, advancing
the drilled pier holes to the design depths or more will be relatively difficult.
13) The top of competent bedrock should be determined in the field at each pier
location. For the purpose of assessing pier length and bedrock penetration, the
top of competent bedrock should not be counted as shallower than the design
depths discussed above, even where encountered at a shallower depth.
14) Lenses or beds of relatively soft bedrock not suitable for foundation support may
be encountered within the competent bedrock section. Construction observation
of such materials penetrated by the pier excavations will require deepening of
individual piers.
15) The bedrock beneath the bridge alignment was very hard and resistant.
Penetration-resistance values of 50 blows for 2 inches were typical. The pier-
drilling contractor should mobilize equipment of sufficient size and operating
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capability to advance the pier holes through deposits including coarse gravel and
boulders, and achieve the design lengths and bedrock penetration. The
contractor should be prepared to core highly resistant bedrock materials.
If refusal is encountered in these materials, a geotechnical engineer should be
retained to evaluate the conditions to establish whether true refusal has been
met with adequate drilling equipment.
16) Groundwater was encountered during subsurface exploration at depths between
about 12 and 20 feet. Water levels presumably will vary with stream stage.
Groundwater will be encountered during drilled pier installation. Casing likely will
be required in the pier holes to reduce water infiltration. In the event that casing
is seated into the bedrock, the minimum bedrock penetration should be taken
from the bottom of the casing.
Seating of the casing in the upper layers of the bedrock may not create positive
cutoff of water infiltration. The contractor should be prepared to address this
condition.
17) In no case should concrete be placed in more than 3 inches of water, unless
placed through an approved tremie method. The proposed tremie method be
discussed during the pre-construction meeting by the Project Team.
18) Where groundwater and unconsolidated soils are encountered, the installation
procedure of drilled piers can be a concern. Commonly in these conditions, the
drilling contractor utilizes casing and slurry during excavation of the pier holes,
which may adversely affect the axial and/or lateral capacities of the completed
piers. During casing withdrawal, the concrete have sufficient slump and must be
maintained with sufficient head above groundwater levels to displace the water or
slurry fully to prevent the creation of voids in the pier.
Because of these considerations, the drilling contractor should submit a written
procedure addressing the use of casing, slurry, and concrete placement prior to
commencement of drilled pier installation.
19) Pier holes should be properly cleaned prior to placement of concrete.
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20) Concrete utilized in the piers should be a fluid mix with sufficient slump so that it
will fill the void between reinforcing steel and the pier hole wall, and help exclude
soil, water and slurry from entering the concrete. Concrete should have a
minimum slump in the range of 5 to 7 inches.
21) Concrete should be placed by an approved tremie method or other method to
reduce mix segregation.
22) Concrete should be placed in a pier on the same day that it is drilled. Failure to
place concrete the day of drilling normally will result in a requirement for
lengthening the pier. The presence of groundwater or caving soils may require
that concrete be placed immediately after the pier hole drilling is completed.
23) The contractor should take care to prevent enlargement of the excavation at the
tops of piers, which could result in “mushrooming” of the pier top. Mushrooming
of pier tops can increase uplift pressures on the piers.
24) Sonic integrity testing (sonic echo or cross-hole sonic) should be performed for
an appropriate percentage of the drilled piers to assess the effectiveness of the
pier construction methods for installing the piers in accordance with project plans
and specifications. Testing 10 percent of drilled piers is common, with testing of
additional piers if voids or other flaws are detected.
Additional information on sonic integrity testing can be provided upon request.
SHALLOW FOUNDATIONS
The geotechnical criteria below should be observed for spread footing foundation
systems for the project retaining walls.
1) The footings should bear on dense, undisturbed, native sands and gravels, or on
‘dental’ concrete. Footing bearing elevations may need to be lowered to
account for scour or because of the conditions exposed at the actual abutment
locations.
Fine-grained, soft, loose, or otherwise deleterious materials exposed at footing
bearing elevation, e.g., a silt pocket, should be removed entirely, and replaced
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with ‘dental’ concrete. Likewise, voids created by removal of boulders should be
replaced with ‘dental’ concrete.
A geotechnical engineer should be retained to observe the surfaces on which the
footings will bear. The exposed surfaces should be approved prior to placement
of reinforcing steel or footing concrete.
2) Based on the conditions encountered at the test holes, footings bearing on
dense, undisturbed, native sands and gravels, or on ‘dental’ concrete, may be
designed for the nominal bearing pressure and resistance factor tabulated below.
Nominal Bearing Pressure Resistance Factor (φb)
10,500 psf 0.45
Compression of the bearing soils for the provided nominal bearing pressure is
estimated to be 1 inch, based on an assumption of drained foundation conditions.
3) In order to reduce differential settlements between footings along continuous
footings, footing loads should be as uniform as possible. Differentially loaded
footings will settle differentially. Similarly, differential fill thicknesses beneath
footings will result in increased differential settlements.
4) Spread footings should have a minimum lateral dimension of 48 inches. Actual
footing dimensions should be determined by the structural engineer.
3) Footings should bear at an elevation below the design scour elevation or
required frost protection depth. Footings should bear at least 3 feet below lowest
adjacent grade for protection against frost; bearing at a greater depth may be
necessary for scour protection.
4) Continuous foundation walls should be reinforced as determined by the structural
engineer.
5) Geotechnical parameters for lateral resistance to foundation loads are provided
in the Lateral Earth Pressures section of this report.
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Shallow Foundation Construction
6) The contractor should take adequate care when making excavations not to
compromise the bearing or lateral support for nearby improvements.
7) Footing excavation bottoms may expose fine grained soils, loose, soft, or
otherwise deleterious materials. Firm materials may become disturbed by the
excavation process. All such unsuitable materials should be excavated and
replaced with ‘dental’ concrete.
8) Fill placed against the sides of the footings should be properly compacted in
accordance with the Project Earthwork section of this report.
ABUTMENT AND WING WALLS
To achieve similar performance, any abutment or wing walls for the bridge should be
supported on drive steel H-pile foundations in the same manner as the bridge, or
supported on the bridge foundations. Geotechnical parameters for driven pile and drilled
pier foundation systems are provided in the Deep Foundations section of this report.
Shallow foundations also appear feasible to support the abutment and wing walls.
Geotechnical parameters for these are provided in the Shallow Foundations section of
this report.
Lateral load parameters for design of the abutment and wing walls for the bridge are
provided in the Lateral Loads section of this report. Wall design should incorporate any
upward sloping backfills, live loads such as traffic, construction equipment, material
stockpiles, etc., and other surcharge pressures.
Wall Drainage Abutment and/or wing walls should be provided with drains at the heels
of the walls. In addition, damp-proofing should be applied to the backs of the walls and
Tencate MiraFi® G-Series backing (or comparable wall drain provisions such as the use
of granular wall backfill) should be placed on the backs of the walls, extending to within
about 2 feet of the top-of-wall. The wall drain system should be in hydraulic continuity
with the heel drain. Each heel drain system should be tested by the contractor after
installation and after placement and compaction of the overlying backfill to verify that the
system functions properly. A typical heel drain detail is provided in Figure 12.
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Geotechnical Parameters for Wall Heel Drain Design Heel drains for the walls should be
designed in accordance with the parameters below. The actual underdrain layout,
outlets, and locations should be developed by wall designer.
1) A heel drain should consist of perforated, rigid, PVC collection pipe at least 3
inches in diameter, non-perforated, rigid, PVC discharge pipe at least 3 inches
in diameter, free-draining gravel, and filter fabric.
2) The free-draining gravel should contain 5 or less percent passing the No. 200
Sieve and 50 or more percent retained on the No. 4 Sieve, and have a
maximum particle size of 2 inches or meet the requirements of CDOT Class B
Filter Material. Each collection pipe should be surrounded on the sides and top
with 6 or more inches of free-draining gravel.
3) The gravel surrounding the collection pipe(s) should be wrapped with filter fabric
(MiraFi 140N® or the equivalent) to reduce the migration of fines into the drain
system.
4) The heel drain system should be designed to discharge at least 15 gallons per
minute of collected water.
5) The discharge point(s) for the collection pipe flow lines should be above the
design ditch stage. (Wall design should incorporate hydrostatic loads on the
portion of the wall below the ditch.)
The collection and discharge pipe for the heel drain system should be laid on a
slope sufficient for effective drainage, but a minimum of 1 percent.
6) ‘Clean-outs’ should be provided for each heel drain to facilitate maintenance of
the underdrains.
7) The discharge piping should be connected to one or more sumps from which
water can be removed by pumping, or to outlet(s) for gravity discharge.
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ABUTMENT APPROACHES
We anticipate that several feet of new fill may be placed to construct the bridge
abutment approaches. Parameters and standards for fill placement and compaction, as
well as fill settlement estimates, are provided in the Project Earthwork section of this
report. Because the abutment fills necessarily are of differential thickness, settlements
generally will be differential with distance from the abutment.
Surface settlements also will depend on the depth and compressibility of the foundation
materials. Penetration resistance values at relatively shallow depths in the test holes
indicate that the fill and native overburden materials underlying the proposed abutment
approach areas at generally are stiff to very stiff or loose to medium dense. Therefore,
we anticipate additional settlement of approximately 1 inch in the new fill soils and the
native soils underlying the abutment approach fills. Differential settlements likely will be
proportional to fill depth.
Bridge and approach design should consider including “sleeper slabs” beneath the
roadway at both abutments to accommodate the transition from bridge structure to filled
ground. “Sleeper slabs” should not be installed until settlement of the approach fill
materials is substantially complete. GROUND estimates that the majority of settlement
will occur within 2 to 3 months after the completion of approach fill placement. Should
scheduling constraints necessitate earlier construction of settlement sensitive structures,
the approach fill(s) should be monitored weekly by surveying the top of the fill near the
abutment. When three successive measurements indicate elevation changes that are
less than ¼-inch, the majority of settlement may be considered to be complete.
LATERAL LOADS
Based on the data obtained for this study and our experience with similar sites and
conditions, lateral load analysis using the Terzaghi or Brown method may take the
values tabulated below for the modulus of horizontal subgrade reaction (Kh) to be
characteristic of the soils and bedrock underlying the site, based on a simplified soil /
bedrock profile. Resistance to lateral loads by deep foundations should be neglected in
the upper 3 feet of soils, whether fill or native.
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HORIZONTAL MODULUS SUBGRADE REACTION (Kh) – TERZAGHI METHOD
Soil / Bedrock Material
Approximate Depth Range
North Abut.
Approximate Depth Range
South Abut.
Kh based on Foundation
Element Width / Diameter
12-inch 14-inch 24-inch 30-inch
Fill 3* – 11 feet 3* – 11 feet 54 tcf
(62 pci)
46 tcf
(53 pci)
27 tcf
(31 pci)
22 tcf
(25 pci)
Sands & Gravels
11* – 15 feet 11* - 15 feet 180 tcf
(208 pci)
154 tcf
(178 pci)
90 tcf
(104 pci)
72 tcf
(83 pci)
Sands & Gravels
15* – 26 feet 15* – 23 feet 300 tcf
(347 pci)
257 tcf
(297 pci)
150 tcf
(174 pci)
120 tcf
(139 pci)
Shale Bedrock
23+ feet 26+ feet 450 tcf
(520 pci)
386 tcf
(446 pci)
225 tcf
(260 pci)
180 tcf
(208 pci)
* Lateral resistance should be neglected Note that the Kh values tabulated above are dependent on deep foundation
for material above the depth of design element width or diameter. Values for other widths / diameters can be provided
scour. upon request.
If “L-Pile” or a similar computer program is used for lateral analysis of the piles, the
geotechnical parameters tabulated below may be used for input into that program, and
are for the same simplified soil / bedrock profile. These include, unit wet weights (γ'), angle of internal friction (), and cohesion (c) for the earth materials, as well as values
for strain at 50 percent of failure stress (50) and horizontal soil modulus (k). Again,
resistance to lateral loads should be neglected in the upper 3 feet of fill or native soils.
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GEOTECHNICAL PARAMETERS FOR LATERAL LOAD ANALYSIS USING L-PILE
Soil / Bedrock Material
Approximate Depth Range
North Abut.
Approximate Depth Range
South Abut.
Parameter
Value
Fill
(model as Sand without Free Water) 3* – 11 feet 3* – 11 feet
γ' 115 pcf
30 degrees
k 0.0622 x 106 pcf
(36 pci)
Sands & Gravels
above the Water Table
(model as Sand without Free Water)
11* – 15 feet
11* – 15 feet
γ' 118 pcf
33 degrees
k 0.289 x 106 pcf
(167 pci)
Sands & Gravels
below the Water Table
(model as Sand with Free Water)
15* – 26 feet
15* – 23 feet
γ' 65 pcf
36 degrees
k 0.480 x 106 pcf
(278 pci)
Claystone Bedrock
(model as Stiff Clay w/o Free Water)
26+ feet
23+ feet
γ' 127 pcf
c 8,900 psf
50 0.004
* Lateral resistance should be neglected for material above the scour depth.
Shallow Foundations Resisting Lateral Loads Cast-in-place footings bearing on
dense, undisturbed, native sands and gravels, or on ‘dental’ concrete, and backfilled with
similar materials may be designed for the parameters tabulated below. The parameters
below also apply to abutment and wing walls. If select, granular imported soil is used as
wall backfill, it should meet the parameters for CDOT Class 1 Structure Backfill.
The lateral earth pressures tabulated below may be used for design of the shallow
foundations,
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The upper 1 foot of embedment should be neglected for passive resistance, however.
Where this passive soil pressure is used to resist lateral loads, it should be understood
that significant lateral strains will be required to mobilize the full value indicated above,
likely 1 inch or more. A reduced passive pressure can be used for reduced anticipated
strains, however.
The values for site soils were based on a moist unit weight (γ') of 120 pcf and an angle
of internal friction () of 32 degrees. The values for CDOT Class 1 Structure Backfill
were based on a moist unit weight (γ') of 132 pcf and an angle of internal friction () of
34 degrees.
EQUIVALENT FLUID WEIGHTS (DRAINED CONDITION)
Backfill
Material
Condition
Active
At-Rest
Passive
Site Soils 38 pcf 57 pcf 390 pcf (to a maximum of 3,900 psf)
CDOT Class 1 Structure Backfill
38 pcf 59 pcf -
EQUIVALENT FLUID WEIGHTS (SUBMERGED CONDITION)
Backfill
Material
Condition
Active
At-Rest
Passive
Site Soils 81 pcf 90 pcf 250 pcf (to a maximum of 2,500 psf)
CDOT Class 1 Structure Backfill
83 pcf 94 pcf -
RESISTANCE FACTORS
Sliding Resistance Passive Resistance
Φτ tan δ Φep
0.80 0.625 0.50
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Wall design should incorporate any upward sloping backfills, live loads such as
construction equipment, material stockpiles, etc., and other surcharge pressures. The
build-up of hydrostatic pressures behind a wall also will increase lateral earth pressures
on the walls.
If select, granular fill is placed as abutment or wing wall backfill, in order to realize the
(lower) values for that material, then the select granular fill should be placed behind the
wall to a minimum distance equal or greater than half of the wall height.
The criteria for CDOT Class 1 Structure Backfill are tabulated in the Project Earthwork
section of this report.
WATER-SOLUBLE SULFATES
The concentrations of water-soluble sulfates measured in samples of the site soils
ranged up to 0.46 percent by weight. (See Table 2.) Such concentrations of soluble
sulfates represent a severe environment for sulfate attack on concrete exposed to these
materials. Degrees of attack are based on the scale of 'negligible,' 'moderate,' 'severe'
and 'very severe' as described in the “Design and Control of Concrete Mixtures,”
published by the Portland Cement Association (PCA). The Colorado Department of
Transportation (CDOT) utilizes a corresponding scale with four classes of severity of
sulfate exposure (Class 0 to Class 3) as described in the table below.
REQUIREMENTS TO PROTECT AGAINST DAMAGE TO CONCRETE BY SULFATE ATTACK FROM EXTERNAL SOURCES OF SULFATE
Severity of Sulfate
Exposure
Water-Soluble Sulfate (SO4
=) In Dry Soil
(%)
Sulfate (SO4=)
In Water
(ppm)
Water / Cementitious Ratio
(maximum)
Cementitious Material
Requirements
Class 0 0.00 to 0.10 0 to 150 0.45 Class 0
Class 1 0.11 to 0.20 151 to 1500 0.45 Class 1
Class 2 0.21 to 2.00 1501 to 10,000 0.45 Class 2
Class 3 2.01 or greater 10,001 or greater 0.40 Class 3
Based on our test results and PCA and CDOT guidelines, sulfate-resistant cement
should be used in all concrete exposed to site soil and bedrock, conforming to one of the
following Class 2 requirements:
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(1) ASTM C 150 Type V with a minimum of a 20 percent substitution of Class F fly
ash by weight
(2) ASTM C 150 Type II or III with a minimum of a 20 percent substitution of Class F
fly ash by weight. The Type II or III cement shall have no more than 0.040
percent expansion at 14 days when tested according ASTM C 452
(3) ASTM C 1157 Type HS; Class C fly ash shall not be substituted for cement.
(4) ASTM C 1157 Type MS plus Class F fly ash where the blend has less than 0.05
percent expansion at 6 months or 0.10 percent expansion at 12 months when
tested according to ASTM C 1012.
(5) A blend of Portland cement meeting ASTM C 150 Type II or III with a minimum of
20 percent Class F fly ash by weight, where the blend has less than 0.05 percent
expansion at 6 months or 0.10 percent expansion at 12 months when tested
according to ASTM C 1012.
(6) ASTM C 595 Type IP(HS); Class C fly ash shall not be substituted for cement.
When fly ash is used to enhance sulfate resistance, it shall be used in a proportion
greater than or equal to the proportion tested in accordance to ASTM C 1012, shall be
the same source, and it shall have a calcium oxide content no more than 2.0 percent
greater than the fly ash tested according to ASTM C 1012.
All concrete exposed to site soil and bedrock should have a minimum compressive
strength of 4,500 psi.
The contractor should be aware that certain concrete mix components affecting sulfate
resistance including, but not limited to, the cement, entrained air, and fly ash, can affect
workability, set time, and other characteristics during placement, finishing and curing.
The contractor should develop mix(es) for use in project concrete which are suitable with
regard to these construction factors, as well as sulfate resistance. A reduced, but still
significant, sulfate resistance may be acceptable to the owner, in exchange for desired
construction characteristics.
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SOIL CORROSIVITY
Data were obtained to support an initial assessment of the potential for corrosion of
ferrous metals in contact with earth materials at the site, based on the conditions at the
time of GROUND’s evaluation. The test results are summarized in Table 2.
Soil Resistivity A sample of materials retrieved from the test hole was tested for
resistivity in the laboratory, at approximately in-place moisture contents. Measurement
of electrical resistivity indicated a value of approximately 953 ohm-centimeters in a
sample of site soils.
pH Where pH is less than 4.0, soil serves as an electrolyte; the pH range of about 6.5 to
7.5 indicates soil conditions that are optimum for sulfate reduction. In the pH range
above 8.5, soils are generally high in dissolved salts, yielding a low soil resistivity.2
Testing indicated a pH value of about 8.4.
Reduction-Oxidation testing indicated red-ox potentials of about -95 millivolts. Such
low potentials typically indicate a more corrosive environment.
Sulfide Reactivity testing indicated “positive” results in the site soils. The presence of
sulfides in the soils suggests a more corrosive environment.
Corrosivity Assessment The American Water Works Association (AWWA) has
developed a point system scale used to predict corrosivity. The scale is intended for
protection of ductile iron pipe but is valuable for project steel selection. When the scale
equals 10 points or higher, protective measures for ductile iron pipe are indicated. The
AWWA scale is presented below. The soil characteristics refer to the conditions at and
above pipe installation depth.
We anticipate that drainage at the site after construction will be effective. Nevertheless,
based on the values obtained for the soil parameters, the overburden soils and bedrock
appear to comprise a severely corrosive environment for ferrous metals (18½ points).
If additional information is needed regarding soil corrosivity, GROUND suggests
contacting the American Water Works Association or a Corrosion Engineer. It should be
2 American Water Works Association ANSI/AWWA C105/A21.5-05 Standard
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noted, however, that changes to the site conditions during construction, such as the
import of other soils, might alter corrosion potentials significantly.
Table A.1 Soil-test Evaluation
Soil Characteristic / Value Points
Soil Resistivity
<1,500 ohm-cm ..........................................................................................… 10 1,500 to 1,800 ohm-cm ................................................................……......…. 8 1,800 to 2,100 ohm-cm .............................................................................…. 5 2,100 to 2,500 ohm-cm ...............................................................................… 2 2,500 to 3,000 ohm-cm .................................................................................. 1 > 3,000 ohm-cm ................................................................................… 0
pH
0 to 2.0 ............................................................................................................ 5 2.0 to 4.0 ......................................................................................................... 3 4.0 to 6.5 ......................................................................................................... 0 6.5 to 7.5 ......................................................................................................... 0 * 7.5 to 8.5 ......................................................................................................... 0 > 8.5 .......................................................................................................... 3
Redox Potential
< 0 (negative values) ....................................................................................... 5 0 to +50 mV ................................................................................................…. 4 +50 to +100 mV ............................................................................................… 3½ > +100 mV ............................................................................................... 0
Sulfide Presence
Positive ........................................................................................................…. 3½ Trace .............................................................................................................… 2 Negative .......................................................................................................…. 0
Moisture
Poor drainage, continuously wet ..................................................................…. 2 Fair drainage, generally moist ....................................................................… 1 Good drainage, generally dry ........................................................................ 0 * If sulfides are present and low or negative redox-potential results (< 50 mV) are obtained, add three points for this range.
PROJECT EARTHWORK
We anticipate that earthwork construction at the site will consist largely of excavation
and backfilling at the bridge abutments. The earthwork criteria and standards below are
directed toward that work and are based on our interpretation of the geotechnical
conditions encountered in the test holes. Where these standards differ from applicable
municipal or agency specifications, e.g., for compaction of utility trench backfill, the latter
should be considered to take precedence.
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General Considerations Prior to earthwork construction, existing concrete, vegetation
and other deleterious materials should be removed and disposed of off-site. Relic
underground utilities should be abandoned in accordance with applicable regulations,
removed as necessary, and properly capped.
Topsoil present on-site should not be incorporated into ordinary fills. Instead, topsoil
should be stockpiled during initial grading operations for placement in areas to be
landscaped or for other approved uses.
Existing Fill Soils Fill soils were recognized at in the test holes and likely are present
elsewhere in the vicinity, as well. All existing fill soils on which project elements bear
should be excavated entirely and replaced by properly compacted fill.
Although the majority of existing fill soils appeared suitable for re-use as compacted fill, it
is possible that some existing fill materials when excavated may not be suitable for re-
use. Debris and other deleterious materials may be encountered. Excavated fill
materials should be evaluated and tested, as appropriate, with regard to re-use.
Use of Existing Native Soils The local native soils that are free of trash, organic
material, construction debris, and other deleterious materials, are suitable, in general, for
placement as compacted fill. Significant quantities of organic materials should not be
incorporated into project fills.
Fragments of rock and boulders, (as well as inert construction debris, e.g., concrete or
asphalt) up to 3 inches in maximum dimension may be included in project fills, in
general. Such materials should be evaluated on a case-by-case basis where identified
during earthwork.
Imported Fill Materials If it is necessary to import material to the site as common fill,
the imported soils should be free of organic material, and other deleterious materials.
Imported material should consist of soils that exhibit 40 percent or less passing the No.
200 Sieve and a plasticity index of 10 or less. Representative samples of the materials
proposed for import should be tested and approved prior to transport to the alignment.
Material to be imported to the site as select, granular fill should meet the criteria for
CDOT Class 1 Structure Backfill (tabulated below).
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CDOT CLASS 1 STRUCTURE BACKFILL
Sieve Size or Parameter Acceptable Range
2-inch 100% passing
No. 4 30% to 100% passing
No. 50 10% to 60% passing
No. 200 5% to 20% passing
Liquid Limit < 35
Plasticity Index < 6
Soils proposed for import as select, granular fill should be tested and approved prior to
transport to the site.
Fill Platform Preparation Prior to filling, the top 12 inches of in-place materials on
which fill soils will be placed should be scarified, moisture conditioned, and properly
compacted in accordance with the criteria below to provide a uniform base for fill
placement.
If surfaces to receive fill expose loose, wet, soft, or otherwise deleterious material,
additional material should be excavated, or other measures taken to establish a firm
platform for filling. A surface to receive fill must be effectively stable prior to placement
of fill.
Wet, Soft, or Unstable Subgrades Where wet, soft, or unstable subgrades are
encountered, the contractor must establish a stable platform for fill placement and
achieving compaction in the overlying fill soils. Therefore, excavation of the unstable
soils and replacing them with relatively dry or granular material, possibly together with
the use of stabilization geo-textile or geo-grid, may be necessary to achieve stability.
Whereas the stabilization approach should be determined by the contractor, GROUND
offers the alternatives below for consideration. Proof-rolling can be beneficial for
identifying unstable areas.
Replacement of the existing subgrade soils with clean, coarse, aggregate (e.g.,
crushed rock or “pit run” materials) or road base. Excavation and replacement to a
depth of 1 to 2 feet commonly is sufficient, but greater depths may be necessary to
establish a stable surface.
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On very weak subgrades, an 18- to 24-inch “pioneer” lift that is not well compacted
may be beneficial to stabilize the subgrade. Where this approach is employed,
however, additional settlements of up to ½ inch may result.
Where coarse, aggregate alone does not appear sufficient to provide stable
conditions, it can be beneficial to place a layer of stabilization geo-textile or geo-
grid (e.g., Tencate Mirafi® HP370 or RS 580i, or Tensar® BX 1100) at the base of
the aggregate section.
The stabilization geo-textile / geo-grid should be selected based on the aggregate
proposed for use. It should be placed and lapped in accordance with the
manufacturer’s recommendations.
Geo-textile or geo-grid products can be disturbed by the wheels or tracks of
construction vehicles; care be taken to maintain the effectiveness of the system.
Placement of a layer of aggregate over the geo-textile / geo-grid prior to allowing
vehicle traffic over it can be beneficial in this regard.
When a given remedial approach has been selected, GROUND suggests constructing a
test section to evaluate the effectiveness of the approach prior to use over a larger area.
General Considerations for Fill Placement Fill soils should be thoroughly mixed to
achieve a uniform moisture content, placed in uniform lifts not exceeding 8 inches in
loose thickness, and properly compacted. No fill materials should be placed, worked,
rolled while they are frozen, thawing, or during poor/inclement weather conditions.
Where soils supporting foundations or on which foundation will be placed are exposed to
freezing temperatures or repeated freeze – thaw cycling during construction – commonly
due to water ponding in foundation excavations – bearing capacity typically is reduced
and/or settlements increased due to the loss of density in the supporting soils. After
periods of freezing conditions, the contractor should re-work areas affected by the
formation of ice to re-establish adequate bearing support.
Care should be taken with regard to achieving and maintaining proper moisture contents
during placement and compaction. Materials that are not properly moisture conditioned
may exhibit significant pumping, rutting, and deflection at moisture contents near
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 32
optimum and above. The contractor should be prepared to handle soils of this type,
including the use of chemical stabilization, if necessary.
Compaction areas should be kept separate, and no lift should be covered by another
until relative compaction and moisture content within the indicated ranges are obtained.
Compaction Standards Soils that classify as A-1, A-2 or A-3 soils in accordance with
the AASHTO classification system (granular materials), including select, granular fill,
should be compacted to 95 percent or more of the maximum dry density at moisture
contents at or above 2 percent below of the optimum moisture content as
determined by AASHTO T 180, the ‘modified Proctor.’
Soils that classify as A-4, A-5, A-6 or A-7 should be compacted to at least 95 percent of
the maximum dry density at moisture contents from at or above 2 percent below the
optimum moisture content as determined by AASHTO T 99, the ‘standard Proctor.’
Settlements Settlements will occur in filled ground, typically on the order of 1 to 2
percent of the fill depth. For a 12-foot fill, for example, that corresponds to a total
settlement of about 2 inches. If fill placement is performed properly, in GROUND’s
experience the majority (on the order of 60 to 80 percent) of that settlement typically will
take place during earthwork construction, provided the contractor achieves the
compaction levels indicated herein. The remaining potential settlements likely will take
several months or longer to be realized, and may be exacerbated if these fills are
subjected to changes in moisture content.
Cut and Filled Slopes Permanent, graded slopes supported by local soils up to 15 feet
in height should be constructed no steeper than 3:1 (horizontal : vertical). Minor
raveling or surficial sloughing should be anticipated on slopes cut at this angle until
vegetation is well re-established. Surface drainage should be designed to direct water
away from slope faces into designed drainage pathways or structures.
EXCAVATION CONSIDERATIONS
Excavation Difficulty Test holes for the subsurface exploration were advanced to the
depths indicated on the test hole logs by means of conventional, truck-mounted,
geotechnical, drilling equipment. We anticipate no unusual excavation difficulties in
these materials, in general, for the proposed construction with conventional, heavy duty,
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 33
excavating equipment. Excavations may encounter coarse gravel or boulders, however.
Fill soils may include coarse construction debris, as well.
Temporary Excavations and Personnel Safety Excavations in which personnel will
be working must comply with all applicable OSHA Standards and Regulations,
particularly CFR 29 Part 1926, OSHA Standards-Excavations, adopted March 5, 1990.
The contractor’s “responsible person” should evaluate the soil exposed in the
excavations as part of the contractor’s safety procedures. GROUND has provided the
information in this report solely as a service to the Interwest Consulting Group, and is not
assuming responsibility for construction site safety or the contractor’s activities.
The contractor should take care when making excavations not to compromise the
bearing or lateral support for any adjacent, existing improvements.
Unless analyzed individually, temporary, un-shored excavation slopes up to 15 feet in
height, in general, should be cut no steeper than 2 : 1 (horizontal : vertical) in the on-site
soils in the absence of seepage. Some surface sloughing may occur on the slope faces
at these angles. Should site constraints prohibit the use of the indicated slope angle,
temporary shoring should be used. GROUND is available to perform shoring design
upon request.
Groundwater and Surface Water Groundwater was encountered in the test holes as
shallowly as about 12 feet. Due to the proximity of the stream, the contractor should
anticipate encountering groundwater in project excavations advanced to depths greater
than about 10 feet. Water levels may be still shallower, however, at some times of the
year and will likely be similar to the stage of Big Thompson River. Wet soils likely will be
encountered above the actual water table.
Should seepage or flowing groundwater be encountered in project excavations, the
slopes should be flattened as necessary to maintain stability or a geotechnical engineer
should be retained to evaluate the conditions and provide additional or alternative
measures, as appropriate. The risk of slope instability will be significantly increased in
areas of seepage along excavation slopes. The contractor should be prepared to work
in wet conditions and in the presence of seepage. De-watering may be beneficial to
excavate and backfill project excavations.
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 34
The contractor should take pro-active measures to control surface waters during
construction and maintain good surface drainage conditions to direct waters away from
excavations and into appropriate drainage structures. A properly designed drainage
swale should be provided at the tops of the excavation slopes. In no case should water
be allowed to pond near project excavations.
Temporary slopes should also be protected against erosion. Erosion along the slopes
will result in sloughing and could lead to a slope failure.
PAVEMENT SECTIONS
A pavement section is a layered system designed to distribute concentrated traffic loads
to the subgrade. Performance of the pavement structure is directly related to the
physical properties of the subgrade soils and traffic loadings. The standard care of
practice in pavement design describes the flexible pavement section as a “20-year”
design pavement: however, most flexible pavements will not remain in satisfactory
condition without routine maintenance and rehabilitation procedures performed
throughout the life of the pavement. Pavement sections were developed in general
accordance with the design guidelines and procedures of the American Association of
State Highway and Transportation Officials (AASHTO), Colorado Department of
Transportation (CDOT), and Larimer County Urban Area Street Standards.
Subgrade Materials Based on the results of our field exploration and laboratory testing,
the potential pavement subgrade materials generally classify as A-2-4 to A-4 soil in
accordance with the AASHTO classification system with Group Index values of 0 to 1.
Resilient modulus testing was performed on a composite sample of shallow site soils
and yield a result of 11,146 psi at optimum moisture content. Based on our experience
at similar sites, however, significant variability in the resilient modulus of subgrade
materials should be anticipated along the project alignment. Therefore, to account for
variability across the alignment a resilient modulus values of 9,500 psi, corresponding to
an R-value of 40, was used to develop the pavement sections. It is important to note
that significant decreases in soil support have been observed as the moisture content
increases above the optimum. Pavements that are not properly drained may experience
a loss of the soil support and subsequent reduction in pavement life.
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 35
Anticipated Traffic The traffic data provided in the Project 318 Request for Proposal
and CDOT loading parameters were used to develop the anticipated project traffic.
These data indicated an average total of 9,400 vehicles per day over the design life of
the pavement, for the subject reach of Namaqua Road. Of these vehicles, 1.5 percent
are estimated by GROUND to be combination (semi-trailer) trucks, and 3.5 percent are
estimated by GROUND to be single-unit trucks. Based on these data, a design value for
flexible pavements of 1,147,464 ESAL18's was calculated (1,572,976 ESAL18's for rigid
pavements). These traffic loadings were used in design of the pavement section for
Namaqua Road.
If design traffic loadings differ significantly from the values above, GROUND should be
notified to re-evaluate the pavement section parameters below.
Pavement Sections The soil resilient modulus and the anticipated ESAL values were
used to determine the required structural number for the project pavements. The
required structural number was then used to develop minimum, pavement sections.
Pavement sections were based on the DARWin™ computer program that solves the
1993 AASHTO pavement design equation. A Reliability Level of 90 percent was utilized
develop the pavement sections. Structural coefficients of 0.44 and 0.11 were used for
hot bituminous asphalt and high quality aggregate base course, respectively. The
pavement section calculations are presented in Appendix B and the resultant minimum
pavement sections are tabulated below.
Minimum Pavement Sections
Subgrade R-Value Composite Section Rigid Section
40 6 inches Asphalt /
8 inches Aggregate Base8½ inches Concrete /
6 inches Aggregate Base
It should be noted that the calculated minimum pavement sections do not meet the
minimum county sections. We understand, however, that those minimums are for use in
the absence of geotechnical data. Therefore, the thinner, R-Value 40 sections may be
used.
Mill and Overlay Option It may be beneficial for the project to asphalt mill sections of
the alignment and overlay them with 2 or more inches of asphalt. Where asphalt milling
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 36
is selected along the alignment, the contractor should anticipate that local areas may
require more extensive repairs (i.e., the complete removal of a damaged section). Areas
needing full reconstruction should be identified in the field once the areas for mill and
overlay have been selected. Existing joints and cracking, along with other signs of
distress, commonly extend into the asphalt overlay after placement. However, with
routine, preventative maintenance, we estimate that the design life of Namaqua Road
can be extended up to 10 years where a 2 inch mill and overlay option is selected.
Thicker overlay sections may increase design life.
Pavement Materials Asphalt pavement should consist of a bituminous plant mix
composed of a mixture of aggregate and bituminous material. Asphalt mixture(s) should
meet the requirements of a job-mix formula established by a qualified engineer as well
as applicable Larimer County and CDOT design requirements.
Aggregate base material should meet the criteria of CDOT Class 6 aggregate base
course. Base course should be placed in and compacted in accordance with the criteria
in the Project Earthwork section of this report.
Where rigid (concrete) pavements are placed, the concrete should consist of a plant mix
composed of a mixture of aggregate, portland cement and appropriate admixtures
meeting the requirements of a job-mix formula established by a qualified engineer as
well as applicable design requirements of the Larimer County and CDOT. Concrete
should have a minimum modulus of rupture of third point loading of 650 psi. Normally,
concrete with a 28-day compressive strength of 4,500 psi should develop this modulus
of rupture value. The concrete should be air-entrained with approximately 6 percent air
and should have a minimum cement content of 6 sacks per cubic yard. Maximum
allowable slump should be 4 inches.
These concrete mix design criteria should be coordinated with other project
requirements including any criteria for sulfate resistance presented in the Water-Soluble
Sulfates section of this report. To reduce surficial spalling resulting from freeze-thaw
cycling, we suggest that pavement concrete meet the requirements of CDOT Class P
concrete. In addition, the use of de-icing salts on concrete pavements during the first
winter after construction will increase the likelihood of the development of scaling.
Placement of flatwork concrete during cold weather so that it is exposed to freeze-thaw
cycling before it is fully cured also increases its vulnerability to scaling. Concrete placing
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 37
during cold weather conditions should be blanketed or tented to allow full curing.
Depending on the weather conditions, this may result in 3 to 4 weeks of curing, or more.
Concrete pavements should contain sawed or formed joints. CDOT and various industry
groups provide guidelines for proper design and concrete construction and associated
jointing. The concrete pavement joints should be fully tied and doweled. Example
layouts that may be applicable for joints, as well as ties and dowels can be found in
CDOT’s M standards, found at the CDOT website:
http://www.dot.state.co.us/DesignSupport/. PCA and ACI publications also provide
useful guidance in these regards.
Subgrade Preparation Although subgrade preparation to a depth of 12 inches is typical
in the project area, pavement performance commonly can be improved by a greater
depth of moisture-density conditioning of the soils. The contractor should be prepared to
prepare the subgrade as outlined herein even where elevated subgrade moisture
contents are encountered beneath the existing pavements.
Remedial Earthwork Shortly before paving, the pavement subgrade should be
excavated and/or scarified to a depth of 12 inches or more, moisture-conditioned and
properly re-compacted. However, as noted in the Geotechnical Considerations for
Design section of this report, all existing, undocumented fill exposed soils in the
subgrade should be excavated and replaced with properly compacted fill. If the owner
opts to leave some of the existing fill soils in-place beneath paved areas, additional
settlements, accelerated pavement distress, and additional maintenance should be
anticipated. Similarly, where existing utility lines or other site constraints limit the depth
to which remedial earthwork can be accomplished, additional maintenance should be
anticipated.
Subgrade preparation should extend the full width of the pavement from back-of-curb to
back-of-curb. The subgrade for sidewalks and other project hardscaping also should be
prepared in the same manner.
Criteria and standards for fill placement and compaction are provided in the Project
Earthwork section of this report. The contractor should be prepared to either dry the
subgrade materials or to moisten them, as needed, prior to compaction.
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 38
Where adequate drainage cannot be achieved or maintained, excavation and
replacement should be undertaken to a greater depth, in addition to the edge drains
discussed below.
Proof Rolling Immediately prior to paving, the subgrade should be proof rolled with a
heavily loaded, pneumatic tired vehicle. Areas that show excessive deflection during
proof rolling should be excavated and replaced and/or stabilized. Areas allowed to pond
prior to paving will require significant re-working prior to proof-rolling. Establishment of a
firm paving platform (as indicated by proof rolling) is an additional requirement beyond
proper fill placement and compaction. It is possible for soils to be compacted within the
limits indicated in the Project Earthwork section of this report and fail proof rolling,
particularly in the upper range of specified moisture contents.
Additional Observations The collection and diversion of surface drainage away from
paved areas is extremely important to the satisfactory performance of the pavements.
The subsurface and surface drainage systems should be carefully designed to ensure
removal of the water from paved areas and subgrade soils. Allowing surface waters to
pond on pavements will cause premature pavement deterioration. Where topography,
site constraints, or other factors limit or preclude adequate surface drainage, pavements
should be provided with edge drains to reduce loss of subgrade support. The long-term
performance of the pavement also can be improved greatly by proper backfilling and
compaction behind curbs, gutters, and sidewalks so that ponding is not permitted and
water infiltration is reduced.
Landscape irrigation in planters adjacent to pavements and in “island” planters within
paved areas should be carefully controlled or differential heave and/or rutting of the
nearby pavements will result. Drip irrigation systems are suggested for such planters to
reduce over-spray and water infiltration beyond the planters. Enclosing the soil in the
planters with plastic liners and providing them with positive drainage also will reduce
differential moisture increases in the surrounding subgrade soils.
In our experience, infiltration from planters adjacent to pavements is a principal source of
moisture increase beneath those pavements. This wetting of the subgrade soils from
infiltrating irrigation commonly leads to loss of subgrade support for the pavement with
resultant accelerating distress, loss of pavement life and increased maintenance costs.
This is particularly the case in the later stages of project construction after landscaping
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 39
has been emplaced but heavy construction traffic has not ended. Heavy vehicle traffic
over wetted subgrade commonly results in rutting and pushing of flexible pavements,
and cracking of rigid pavements. In relatively flat areas where design drainage gradients
necessarily are small, subgrade settlement can obstruct proper drainage and yield
increased infiltration, exaggerated distress, etc. (These considerations apply to project
flatwork, as well.)
Also, GROUND’s experience indicates that longitudinal cracking is common in asphalt-
pavements generally parallel to the interface between the asphalt and concrete
structures such as curbs, gutters, or drain pans. Distress of this type is likely to occur
even where the subgrade has been prepared properly and the asphalt has been
compacted properly.
The anticipated traffic loading does not include excess loading conditions imposed by
heavy construction vehicles. Consequently, heavily loaded concrete, lumber, and
building material trucks can have a detrimental effect on the pavement.
Most pavements will not remain in satisfactory condition and achieve their “design lives”
without regular maintenance and rehabilitation procedures performed throughout the life
of the pavement. Maintenance and rehabilitation measures preserve, rather than
improve, the structural capacity of the pavement structure. Therefore, an effective
program of regular maintenance should be developed and implemented to seal cracks,
repair distressed areas, and perform thin overlays throughout the lives of the pavements.
The greatest benefit of pavement overlaying will be achieved by overlaying sound
pavements that exhibit little or no distress.
Crack sealing should be performed at least annually and a fog seal/chip seal program
should be performed on the pavements every 3 to 4 years. After approximately 8 to 10
years after construction, patching, additional crack sealing, and asphalt overlay may be
required. Prior to overlays, it is important that all cracks be sealed with a flexible,
rubberized crack sealant in order to reduce the potential for propagation of the crack
through the overlay. If actual traffic loadings exceed the values used for development of
the pavement sections, however, pavement maintenance measures will be needed on
an accelerated schedule.
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 40
CLOSURE
Geotechnical Review GROUND should review project plans and specifications to
evaluate whether they comply with the intent of the measures discussed in this report.
The author of this report or a GROUND principal should be contacted to provide that
review. The review should be requested in writing.
The geotechnical conclusions and parameters presented in this report are contingent
upon observation and testing of project earthworks by representatives of GROUND. If
another geotechnical consultant is selected to provide materials testing, then that
consultant must assume all responsibility for the geotechnical aspects of the project by
concurring in writing with the parameters in this report, or by providing alternative
parameters.
Materials Testing Interwest Consulting Group should consider retaining a geotechnical
engineer to perform materials testing during construction. The performance of such
testing or lack thereof, however, in no way alleviates the burden of the contractor or
subcontractor from constructing in a manner that conforms to applicable project
documents and industry standards. The contractor or pertinent subcontractor is
ultimately responsible for managing the quality of his work; furthermore, testing by the
geotechnical engineer does not preclude the contractor from obtaining or providing
whatever services that he deems necessary to complete the project in accordance with
applicable documents.
Limitations This report has been prepared for Interwest Consulting Group as it pertains
to design and construction of the proposed bridge and related improvements as
described herein. It may not contain sufficient information for other parties or other
purposes.
In addition, GROUND has assumed that project construction will commence by Spring,
2017. Any changes in project plans or schedule should be brought to the attention of a
geotechnical engineer, in order that the geotechnical conclusions in this report may be
re-evaluated and, as necessary, modified.
The geotechnical conclusions and criteria in this report relied upon subsurface
exploration at a limited number of exploration points, as shown in Figure 1, as well as
the means and methods described herein. Subsurface conditions were interpolated
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 41
between and extrapolated beyond these locations. It is not possible to guarantee the
subsurface conditions are as indicated in this report. Actual conditions encountered
during construction may differ from those encountered during site exploration.
If during construction, surface, soil, bedrock, or groundwater conditions appear to be at
variance with those described herein, a geotechnical engineer should be retained at
once, so that re-evaluation of the conclusions for this site may be made in a timely
manner. In addition, a contractor who relies upon this report for development of his
scope of work or cost estimates may find the geotechnical information in this report to be
inadequate for his purposes or find the geotechnical conditions described herein to be at
variance with his experience in the greater project area. The contractor is responsible
for obtaining the additional geotechnical information that is necessary to develop his
workscope and cost estimates with sufficient precision. This includes current depths to
groundwater, etc.
If any information referred to herein is not well understood, then Interwest Consulting
Group, the owner, or anyone using this report, should contact the author or a GROUND
principal immediately. We will be available to meet to discuss the risks and remedial
approaches presented in this report, as well as other potential approaches, upon
request.
This report was prepared in accordance with generally accepted soil and foundation
engineering practice in the Larimer County area at the date of preparation. Current
applicable codes may contain criteria regarding performance of structures and/or site
improvements which may differ from those provided herein. Our office should be
contacted regarding any apparent disparity. GROUND makes no warranties, either
expressed or implied, as to the professional data, opinions or conclusions contained
herein.
This document, together with the concepts and conclusions presented herein, as an
instrument of service, is intended only for the specific purpose and client for which it was
prepared. Reuse of or improper reliance on this document without written authorization
and adaption by GROUND Engineering Consultants, Inc., shall be without liability to
GROUND Engineering Consultants, Inc.
Namaqua Avenue Bridge Over the Big Thompson River
Loveland, Colorado Revised 4
Job No. 16-3048 GROUND Engineering Consultants, Inc. Page 42
GROUND appreciates the opportunity to complete this portion of the project and
welcomes the opportunity to provide Interwest Consulting Group or the owner with a
proposal for construction observation and materials testing.
Sincerely,
GROUND Engineering Consultants, Inc.
Ben Fellbaum Reviewed by Brian H. Reck, P.G., C.E.G., P.E.
DIRECT SHEAR TEST REPORT
GROUND ENGINEERING CONSULTANTS, INC.ENGLEWOOD, CO.
Client: Interwest Consulting Group
Project: Replacement Namaqua Avenue Bridge Replacement over the Big
Thompson River
Location: TH-2
Depth: 17 feet
Proj. No.: 16-3048 Date Sampled: 8/15/16
Sample Type: Liner
Description: Silty SAND
LL= NV PI= NP
Specific Gravity= 2.69
Remarks:
Figure 8
Sample No.
Water Content, %
Dry Density, pcf
Saturation, %
Void Ratio
Diameter, in.
Height, in.
Water Content, %
Dry Density, pcf
Saturation, %
Void Ratio
Diameter, in.
Height, in.
Normal Stress, psi
Fail. Stress, psi
Strain, %
Ult. Stress, psi
Strain, %
Strain rate, in./min.
Initia
lA
t T
est
Sh
ea
r S
tre
ss,
psi
0
10
20
30
40
50
60
Strain, %
0 5 10 15 20
1
2
3
Fail.
Str
ess,
psi
0
20
40
60
Normal Stress, psi
0 20 40 60 80 100 120
C, psi
, deg
Tan()
Results
4.31
33.2
0.65
1
11.6
105.1
52.0
0.5973
1.94
1.00
7.7
106.2
35.6
0.5819
1.94
0.9910.0010.72
3.6
0.01
2
11.6
105.7
52.8
0.5885
1.94
1.00
7.7
108.1
37.2
0.5541
1.94
0.9820.0017.59
6.2
0.01
3
11.6
26.4
5.8
5.3542
1.94
4.00
7.3
26.6
3.7
5.3086
1.94
3.9740.0030.41
8.8
0.01
0.5
0.6
0.7
0.8
0.9
1
1.1
1 1.5 2 2.5 3 3.5
Center to Center Spacing (Pier / Pile Diameter)
Ax
ial
Ca
pa
cit
yR
ed
uc
tio
n F
ac
tor
-0.100
0.000
0.100
0.200
0.300
0.400
0.500
0.600
0.700
0.800
0.900
1.000
1.100
0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0 4.5 5.0 5.5 6.0 6.5 7.0 7.5 8.0 8.5 9.0
1st Pier / Pile
2nd Pier / Pile
3rd & Subsequent Piers / Piles
Center to Center Spacing (Pier / Pile Diameter)
La
tera
l C
ap
ac
ity
Re
du
cti
on
Fa
cto
r (p
mu
ltip
lier)
TABLE 1
SUMMARY OF LABORATORY TEST RESULTS
Sample Location Natural Natural Percent Atterberg Limits Percent Unconfined USCS AASHTO
Test Moisture Dry Passing Liquid Plasticity Swell Compressive Classifi- Classifi- Soil or
Hole Depth Content Density Gravel Sand No. 200 Limit Index (Surcharge Strength cation cation Bedrock Type
No. (feet) (%) (pcf) (%) (%) Sieve Pressure) (psf) (GI)
1 24 14.9 SD - - 10 21 5 - - SP-SC A-2-4 (0) SAND with Silty, Clay
2 7 4.7 SD - - 14 NV NP - - SM A-2-4 (0) FILL: Silty Sand
2 17 11.6 SD - - 13 NV NP - - SM A-2-4 (0) Silty SAND
2 30 6.0 120.3 - - 91 29 8 - 22,360 CL A-2-4 (0) Clay SHALE Bedrock
2 38 5.5 120.5 - - 88 28 9 - 42,160 CL A-2-4 (0) Clay SHALE Bedrock
2 - 6.1 SD - - 87 27 7 - - CL A-2-4 (0) Clay SHALE Bedrock
3 5 19.3 96.5 - - 75 25 3 - - ML A-4 (1) FILL: Silt with Sand
3 15 14.8 - 1 79 20 - - - - SM - Silty SAND
3 30 9.7 124.2 - - 90 31 11 - 11,620 CL A-6 (1) Clay SHALE Bedrock
4 3 8.4 101.2 - - 19 NV NP - - SM A-2-4 (0) FILL: Silty Sand
4 13 11.0 106.8 4 88 8 NV NP - - SP-SM A-2-4 (0) Silty SAND
4 29 7.3 113.3 - - 92 30 8 - - CL A-2-4 (0) Clay SHALE Bedrock
4 45 6.7 113.1 - - 90 30 10 - 8,760 CL A-2-4 (0) Clay SHALE Bedrock
P-1 1 6.0 122.9 - - 18 NV NP - - SM A-2-4 (0) FILL: Silty Sand
P-2 3 16.3 102.9 - - 39 23 4 -0.1 (150 psf) - SC-SM A-4 (0) FILL: Silty, Clayey Sand
P-3 2 11.5 105.9 - - 40 21 3 - - SM A-4 (0) FILL: Silty Sand
P-4 2 14.9 111.8 - - 47 25 5 - - SC-SM A-4 (0) FILL: Silty, Clayey Sand
P-5 3 18.9 98.0 - - 47 24 3 -0.2 (400 psf) - SM A-4 (0) FILL: Silty Sand
P-6 1 5.7 99.2 - - 15 20 2 - - SM A-2-4 (0) FILL: Silty Sand
Job No. 16-3048
Gradation
TABLE 2
SUMMARY OF SOIL CORROSION TEST RESULTS
Sample Location Water Redox Sulfides USCS
Test Soluble pH Potential Content Resistivity Classifi- Soil or
Hole Depth Sulfates cation Bedrock Type
No. (feet) (%) (mV) (ohm-cm)
3 5 0.46 8.6 -95 Positive 953 ML SILT with Sand
P-2 3 0.02 - - - - SM FILL: Silty Sand
P-6 1 0.06 - - - - SM FILL: Silty Sand
Job No. 16-3048
APPENDIX A
Core Hole Logs
APPENDIX B
Pavement Section Calculations
Page 1
1993 AASHTO Pavement Design
DARWin Pavement Design and Analysis System
A Proprietary AASHTOWareComputer Software Product
Network Administrator
Flexible Structural Design Module
16-3048Namaqua Bridge ReplacementAsphalt - Composite Section
R-40
Flexible Structural Design
18-kip ESALs Over Initial Performance Period 1,147,464 Initial Serviceability 4.5 Terminal Serviceability 2.5 Reliability Level 90 %Overall Standard Deviation 0.44 Roadbed Soil Resilient Modulus 9,500 psiStage Construction 1
Calculated Design Structural Number 3.15 in
Specified Layer Design
Layer Material Description
StructCoef.(Ai)
DrainCoef.(Mi)
Thickness(Di)(in)
Width(ft)
CalculatedSN (in)
1 Asphalt 0.44 1 6 - 2.642 Road Base (CDOT Class 6) 0.11 1 8 - 0.88
Total - - - 14.00 - 3.52
Page 1
1993 AASHTO Pavement Design
DARWin Pavement Design and Analysis System
A Proprietary AASHTOWareComputer Software Product
Network Administrator
Flexible Structural Design Module
16-3048Namaqua Bridge ReplacementAsphalt - Full Depth Section
R-40
Flexible Structural Design
18-kip ESALs Over Initial Performance Period 1,147,464 Initial Serviceability 4.5 Terminal Serviceability 2.5 Reliability Level 90 %Overall Standard Deviation 0.44 Roadbed Soil Resilient Modulus 9,500 psiStage Construction 1
Calculated Design Structural Number 3.15 in
Specified Layer Design
Layer Material Description
StructCoef.(Ai)
DrainCoef.(Mi)
Thickness(Di)(in)
Width(ft)
CalculatedSN (in)
1 Asphalt 0.44 1 7.5 - 3.30Total - - - 7.50 - 3.30
Rigid Pavement Design - Based on AASHTO Supplemental Guide
Reference: LTPP DATA ANALYSIS - Phase I: Validation of Guidelines for k-Value Selection and Concrete Pavement Performance Prediction
Results
Project # 16-3048Description: Namaqua Bridge Replacement
Location: Loveland, CO
Slab Thickness Design
Pavement Type JPCP18-kip ESALs Over Initial Performance Period (million) 1.60 millionInitial Serviceability 4.5Terminal Serviceability 2.528-day Mean PCC Modulus of Rupture 650 psiElastic Modulus of Slab 3,400,000 psiElastic Modulus of Base 30,000 psiBase Thickness 6.0 in.Mean Effective k-Value 25 psi/inReliability Level 90 %Overall Standard Deviation 0.34
Calculated Design Thickness 7.85 in
Temperature Differential
Mean Annual Wind Speed 8.8 mph
Mean Annual Air Temperature 48 oFMean Annual Precipitation 15.7 in
Maximum Positive Temperature Differential 6.02 oF
Modulus of Subgrade Reaction
Period Description Subgrade k-Value, psi
3 Winter 253 Spring 253 Summer 253 Fall 25
Seasonally Adjusted Modulus of Subgrade Reaction 25 psi/in
Modulus of Subgrade Reaction Adjusted for Rigid Layer
and Fill Section 25 psi/in
Traffic
Performance Period yearsTwo-Way ADTNumber of Lanes in Design DirectionPercent of All Trucks in Design LanePercent Trucks in Design Direction
Vehicle Class Percent of Annual Initial Annual AccumulatedADT Growth Truck Factor Growth in 18-kip ESALs
Truck Factor (millions)
Total Calculated Cumulative ESALs million
Faulting
Doweled
Dowel Diameter 1 inDrainage Coefficient 1.00
Average Fault for Design Years with Design Inputs 0.05 inCriteria Check PASS
Nondoweled
Drainage Coefficient 1
Average Fault for Design Years with Design Inputs inCriteria Check
Traffic Loading Calculations GROUND Job No. 16-3048
18 kip Loadings, S.I. = 2.5 Equivalent 18 kip Axle Loadings
High Average Method
Alignment: Namaqua Bridge
ESAL18's
Long Term Traffic Flexible Rigid
Pavement Pavement
Total Vehicles per day: 9,400
% Combination Trucks: 1.5 % 141 vehicles 671,309 1,044,945
% Single Unit Trucks: 3.5 % 329 vehicles 358,814 410,691
% Cars & Pickups: 95 % 8930 vehicles 117,340 117,340
No. Of Traffic Lanes = 2 Total: 1,147,464 Total: 1,572,976
(total, both ways) Lane Factor: 0.60 EDLA: 157 EDLA: 215
Design Life (Years): 20 Source of Traffic Data: Larimer County RFP for Project 318