Bradley T. RogersEngineerLJB, Inc.

Dayton, Ohio

Richard A. Miller, Ph. D.Associate Professor of Civil EngineeringUniversity of CincinnatiCincinnati, Ohio

As part of the FHWA (Federal Highway Admin-istration) High Performance Concrete Bridge Program, two full-scale truckload tests of Bridge GUE-22-6.57 were carried out. The main ob-jectives of these tests were to investigate the static and dynamic response of the high perfor-mance concrete (HPC) structure. A secondary objective was to investigate the load transfer between the box girders through experimental middepth shear keys. The structure was loaded using standard Ohio Department of Transporta-tion (ODOT) dump trucks. A model test of the bridge was conducted as well. It was found that the bridge behavior is well predicted using sim-ple models. The bridge behaves as a single unit and all girders share the load almost equally. The dynamic behavior of the bridge is typical for comparable structures.

Evaluation of a High Performance Concrete Box Girder Bridge

60 PCI JOURNAL

Bahram M. Shahrooz, Ph. D.Associate Professor of Civil

EngineeringUniversity of Cincinnati

Cincinnati, Ohio

T. Michael Baseheart, Ph. D.Associate Professor of Civil EngineeringUniversity of CincinnatiCincinnati, Ohio

Andreas GreuelGraduate Research AssistantUniversity of CincinnatiCincinnati, Ohio

November-December 2000 6�

The use of high performance con-crete (HPC) can lead to more economical bridge designs be-cause the designer can often eliminate girder lines, use shallower sections or extend the span of a section and elimi-nate the need for intermediate supports. While the use of HPC can lead to lower initial costs through elimination of piers or girder lines, the high durability of HPC can also lead to lower long-term costs because of reduced maintenance and a longer bridge life.

Despite the above advantages, as de-signers use longer and shallower spans with fewer girders, the structures become more flexible. This leads one to ques-tion whether the structures will deflect or vibrate excessively or if the greater flexibility will affect the load distribution between adjacent box girders.

In 1998, the Ohio Department of Transportation (ODOT) constructed a HPC, adjacent box girder bridge. In this bridge, a preliminary design for a three-span bridge was converted to a single-span design [see Figs. 1(a) and 1(b)]. Prior to construction, research-ers at the University of Cincinnati (UC) formulated the high performance concrete mixes which would be used and conducted several tests on proto-type girders.

The prototype test results showed the simple span box girders to be suf-ficiently strong and ductile. These tests confirmed that girder behavior was predictable using linear elastic theory prior to cracking and a strain compat-ibility approach for the ultimate be-havior. Satisfied with prototype girder performance, ODOT let the contract for the bridge construction, yet, there were still questions as to how the ac-tual bridge would behave.

To try to answer these questions, the UC research team tested the bridge during and after construction. This paper summarizes the result of destruc-tive and nondestructive testing of pro-totype girders and of nondestructive testing the completed bridge structure.

BRIDGE DESCRIPTIONBridge GUE-22-6.57 is a 115.5 ft

(35.2 m) long prestressed, noncompos-ite, adjacent box girder bridge [see Fig. 1(a)]. The cross section of the bridge is shown in Fig. 2. The structure is Fig. 2. Cross section of HPC bridge.

Fig. 1(a). HPC single-span design.

Fig. 1(b). Original, three-span design.

located on US Route 22, a heavily traveled two-lane highway near Cam-bridge, Ohio.

The new bridge replaced a 70 ft (21.4 m) single span, steel stringer bridge with a concrete deck. The hy-draulic requirements required the span of the new bridge to be increased to 115.5 ft (35.2 m ). Note that the bridge was originally designed as a three-span, noncomposite, adjacent box girder system using normal strength concrete [5500 psi (35 MPa)] and 0.5 in. (13 mm) diameter strands [see Fig. 1(b)].

Normally, a span of 115.5 ft (35.2 m) would not be too long for a box girder. However, the Ohio box girder (see Fig. 3) differs from the AASHTO standard

in that the Ohio box girder has only a 5 in. (127 mm ) thick bottom flange rather than the 5.5 in. (140 mm) flange used in the AASHTO box. As a result, the Ohio box can accommodate only a single, full layer of 23 strands in the bottom flange and a partial layer of four strands and several layers of two strands in the webs.

If the bridge were designed using the largest Ohio box (B42-48, see Fig. 3), standard 5500 psi (39 MPa) concrete and 0.5 in. (13 mm) diam-eter strands, the maximum span of the girder would be 110 ft (33.5 m). If the steel area is increased by using 0.6 in. (15 mm) diameter strand, the girder could easily span the required length. However, the larger strand

62 PCI JOURNAL

generates a greater prestressing force and, therefore, high strength concrete is required.

Another reason for using HPC is that it has a high durability. Adja-cent box girders have shear keys cast between the girders to transfer shear between adjacent girders. Shear keys of adjacent box girder bridges tend to crack, and subsequently, allow

water from the bridge surface to leak between the sides of adjacent box girders.

Leakage can cause serious damage to the tendons and reinforcement when water and deicing chemicals penetrate the concrete. Leakage is a major prob-lem with adjacent box girder bridges, so HPC with its high durability is an ideal choice for this type of structure.

Fig. 3. Ohio B42-48 section.

Fig. 4. Installing tie rods.

This bridge utilizes an experimental shear key at middepth of the cross section. This shear key configuration was found to be less susceptible to cracking.1 After tying adjacent girders together with non-prestressed threaded rods located transversely through dia-phragms at the ends and quarter points of the bridge (see Fig. 4), the shear keys are grouted.

Note that in this configuration, only the shear key itself was to be grouted. The area above the middepth shear key was filled with sand and a sealant was applied to the top of the joint to further guard against leakage (see Fig. 5).

MATERIAL PROPERTIESThe first task for the UC research

team was developing a mix design which would produce extremely du-rable concrete that had high release and ultimate strengths. To make the girder span the required 115.5 ft (35.2 m), it was determined that the concrete would need a minimum compressive strength of 10 ksi (70 MPa ). For dura-bility, a rapid chloride permeability of less than 1000 coulombs was desired.

The mix was designed using the ma-terials which the precaster had avail-able. Normally, Type I cement is used for HPC, but the precaster usually used Type III and it was not economically desirable to change to Type I.

To improve durability and strength, a water to cementitious material ratio (w/c+p) of less than 0.3 was chosen and microsilica was added to the mix. The precaster did not have a silo avail-able to store the microsilica, so it was batched from bags. To avoid having to weigh the microsilica separately, the mix was designed using single bag increments of 25 lb (11.3 kg).

This is why there is an unusual per-centage (11.8 percent) for the micro-silica. Because of the low w/c+p ratio and presence of microsilica, a water reducer was required both to provide enough workability and to defloculate the cement particles so the microsilica would be able to fit in between them and densify the mix.

The fine aggregate was natural river sand. A No. 8, partially crushed, river gravel (3/8 in. or 10 mm max.) was used as the coarse aggregate. Be-cause the aggregate was only partially

November-December 2000 63

crushed, the aggregate/paste bond did not appear to be particularly good and this appeared to limit the concrete strength. Making the specified strength required a high cement content. The specified strength could have been obtained with lower cement contents and/or higher w/c+p ratios with the use of a better aggregate (e.g., crushed limestone).

Thirteen trial mixes were investi-gated.2 The final mix proportions are shown in Table 1. During fabrication of the prototype and final bridge gird-ers, test cylinders and modulus of rup-ture beams were cast and later tested to determine the concrete properties. Table 2 provides a list of the measured concrete properties.

GIRDER PROPERTIESThe girders are standard ODOT

B42-48 box girders (see Fig. 3) and were designed using the provisions of the AASHTO Standard Specifica-tions.3 The ODOT require bridges to be designed for an HS-20 loading, but this bridge was designed using the HS-25 loading to allow for future increases in traffic loads.

Because of the experimental aspect of the bridge, the engineer added a few extra strands for additional safety. Thirty 0.6 in. (15 mm) diameter strands were used in the bottom flange with two additional strands placed in the top flange to control tension at release.

DESTRUCTIVE TESTING OF PROTOTYPE GIRDERS

Prior to the fabrication of the actual bridge girders, two prototype box gird-ers were cast and tested. This was done so that the precaster could gain more experience in placing HPC and so that

Fig. 5. Middepth shear key

and tie rods.

the research team could verify the be-havior of the girders. An in-depth dis-cussion of the prototype fabrication and destructive testing can be found in a previous work.4 For completeness, a summary is presented here.

One area explored was the trans-fer length for 0.6 in. (15 mm) diam-eter strand when used with HPC. To measure transfer length, vibrating wire

strain gauges were placed between the strands at approximately 1 ft (3 m) intervals from each end of the girder prior to casting. After the girder had cured and the strands were cut, the measured strain was used to determine that somewhere between 35 in. and 48 in. (0.89 and 1.2 m) the transfer was complete.

This means that the transfer length was between 60D and 80D (where D is the strand diameter). The AASHTO Standard Specifications2 use a transfer length of 50D, the AASHTO LRFD Specifications5 use a transfer length of 60D while a transfer length of 80D has been suggested in the literature.6

The prototype girders were subjected to destructive testing. Each girder was supported on neoprene pads such that the test span was 115.5 ft (35.2 m) and loaded with two-point loads placed

Compressive strength at release* 8 ksi (56 MPa)

Compressive strength at 56 days‡ 11.8 ksi (80 MPa)

Modulus of elasticity at 56 days‡ 5800 ksi (40000 MPa)

Modulus of rupture at 56 days§ 1250 psi (8.6 MPa)

Split strength at 56 days† 620 psi (4.3 MPa)

Shrinkage after one year§ 0.1 percent

Creep coefficient after one year** 2.1

Rapid chloride permeability at 56 days 360 coulumbs

Table 2. Average concrete properties.

* 6 x 12 in. cylinders - bed cured. † 6 x 12 in. cylinders - moist cured. ‡ 6 x 6 x 14 in. beams - moist cured. § 3 x 3 x 11.25 in. prisms - 28 days moist cure** 6 x 12 in. cylinders - test started at 1 day (i.e., at release of prestressing force)

Material Lbpercuyd kg/m3 Cement 846 497 Microsilica 100 58.7 Water 262 145 Fine aggregate 927 544 Coarse aggregate 1774 1042 Air 6 percent 6 percent

Table 1. Concrete mix proportions.

Admixtures Ouncespercuyd l/m3 Air entrainer 21 0.81 Retarder 28 1.08 High range water reducer 203 7.84

64 PCI JOURNAL

47.25 ft (14.4 m) from each support (see Fig. 6). Loads were applied such that the deflection of the girder under the load points increased in 0.05 in. (1.3 mm) increments.

The first cracks in both girders ap-peared when the applied load at each point was approximately 42 kips (187 kN). Counting self weight, the total cracking moment was 3570 kip-ft (4840 kN-m). This value was larger than the 3155 kip-ft (4280 kN-m) cracking moment predicted using the provisions of the AASHTO Standard Specifications.

Note that in calculating the crack-ing moment, a loss of prestressing force of 20 percent was assumed. This loss value was calculated from the provisions of the AASHTO Standard Specifications.

One reason for the difference be-tween the calculated and measured values of the cracking moment is that the modulus of rupture was taken as 7.5 (in psi units), or 750 psi (5.3

MPa). However, HPC has a

higher modulus of rupture. If the mea-sured modulus of rupture of 1250 psi (8.6 MPa) is used, the calculated cracking moment becomes 3525 kip-ft (4780 kN-m), which is very close to the measured value.

The use of the higher modulus of rupture is important since the provi-sions of the AASHTO Standard Speci-fications require that φ Mn > 1.2 Mcr, where Mn is the nominal moment, Mcr is the cracking moment and φ is the capacity reduction factor. If Mcr is un-derestimated, the design may not be conservative.

The girders were loaded to approxi-mately 80 kips (356 kN) at each load point. Again, accounting for self weight, the applied moment was 5320 kip-ft (7220 kN-m). This exceeded the ulti-mate moment of 5130 kip-ft (6960 kN-m), calculated using the provisions of the AASHTO Standard Specifications.

Under this load, the girder had a very large deflection of 24 in. (61 m) at the midspan. This corresponds to a deflection of L/58, where L is the span.

The test was stopped at this point for safety reasons.

Fig. 7 shows a load versus midspan deflection graph for one of the pro-totype girders. Also plotted is the be-havior predicted using a computer pro-gram called RESPONSE.7 For a given moment, RESPONSE generates cur-vature and strain values for the cross section by strain compatibility.

Deflections are found by first choos-ing an applied load level, plotting the moment diagram for the total moment (dead load plus applied load), and then using the result of this analysis to plot a curvature diagram. The curvature can then be double integrated to find the deflection.

Load versus strain behavior is avail-able directly from the RESPONSE results. Clearly, the girder deflection behavior is well predicted. A graph of load versus extreme fiber compressive strain (see Fig. 8) also shows good agreement between the analysis and experiments up to a strain of 0.002 (where the test was stopped).

Fig. 6. Testing the prototype girder.

Fig. 7. Load versus deflection for prototype girder. Fig. 8. Load versus top compressive strain for prototype girder.

November-December 2000 65

Measured losses of prestressing force were determined experimentally on the prototype girders. The girders were loaded to cracking and then un-loaded. When unloaded, the prestress-ing force causes the cracks to close. Clip gauges were placed across the closed cracks so that the crack open-ing could be measured when the girder was reloaded.

When the girder is reloaded, the ten-sile stress from the applied moment relieves the compressive stress caused by the prestressing force and eventu-ally overcomes it, causing the crack to open. Theoretically, the load at which the crack just begins to open (as mea-sured by the clip gauges) corresponds to a zero stress state. Knowing the applied moment which causes a state of zero stress at a known point on the girder, it is possible to back-calculate the effective prestressing force.

The measured values are shown in Table 3. Calculated values of the loss of prestressing force, from two dif-ferent methods, are also presented. One method of calculating the loss of prestress is that given in the PCI Design Handbook8 and the other is the method given in the AASHTO Stan-dard Specifications.2

The calculated values of loss of pre-stressing force are final values, which assume a large amount of time has passed and all of the creep, shrink-age and relaxation of the strands has occurred. The prototype girders were only 6 to 9 months old at the time of testing, so an exact comparison cannot be made. However, at 6 to 9 months, approximately 80 percent of the time-dependent deformations have occurred. With this in mind, the data in Table 3 suggest that the calculated values of loss of prestressing force are reasonable.

The results of the prototype testing clearly showed that the girders were sufficiently strong and predictable. With the prototype girder testing com-plete, the girders for the actual bridge were fabricated.

FABRICATION OF THE BRIDGE GIRDERS

The actual bridge girders were fab-ricated at Prestress Services, Mel-bourne, Kentucky. No major prob- Fig. 9. Girder numbering for Phase I and Phase II.

Girder1 Girder2 PCI AASHTO Crack 1 Crack 2 Crack 1 17 percent 18 percent 17.5 percent 18 percent 20 percent

Table 3. Loss of prestressing force – prototype girder.

lems occurred during the fabrication of the girders and only minor adjust-ments to the usual fabrication process were needed to accommodate the use of HPC. Most of these accommoda-tions centered around the fact that HPC tends to be sticky, does not move well under vibration and has a high rate of slump loss. The workers sim-ply needed to use greater care in the method and rate of placement to pre-vent cold joints or voids.

Girder camber was measured at re-lease and at various times while the girders were in storage. At release, the girders showed a camber of about 1/8 in. (3 mm), but growth in storage began immediately. Before the girders were shipped to the construction site, they had cambered up an average of 1.5 in. (38 mm). This behavior is in reasonable agreement with a predicted value of 1.66 in. (42 mm), which was calculated using elastic theory and ac-counting for time-dependent effects by using the multipliers given in the PCI Design Handbook.

Embedded strain gauges (see next section on Instrumentation) were read before and after detensioning and sev-eral times during girder storage. Using these readings, the strain change in concrete at strand level was calculated to be 580 microstrain at transfer. Thus, the loss at detensioning due to elastic shortening becomes 8.8 percent of the

initial prestressing force. The calcu-lated loss due to elastic shortening was approximately 6 percent.

INSTRUMENTATION OF THE GIRDERS

During the fabrication of the actual box girders for the bridge, a total of 69 internal sensors were embedded. A total of 57 vibrating wire strain gauges and 12 bonded foil strain gauges were embedded inside the girders. All the internal sensors were installed at mid-span.

Vibrating wire strain gauges were used to measure static truckload strains but are not capable of measuring dy-namic events. In addition, temperature changes during curing and afterwards could be measured by thermistors in the vibrating wire gauges. The foil strain gauges measure instantaneous strains under traffic loading and can capture dynamic responses from mov-ing trucks.

Depending on the girder location, the number and layout of the embedded strain gauges were varied. This scheme allowed a reasonable and adequate number of sensors. All the girders had a single foil strain gauge near the middle of the bottom flange. As a minimum, two vibrating wire strain gauges were used in each girder – one in the top flange and one in the bottom flange.

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Additional sensors were installed in some of the girders to measure the distribution of strains and temperature gradient through the depth and across the width of the girders. Considering that Girders 4 through 9 (see Fig. 9) will resist most of the traffic loads, these girders were instrumented more heavily than the other girders.

The vibrating wire strain gauges in the bottom flange were placed between the strands and gauges in the web were placed between the stirrups. To ensure adequate protection and long-term du-rability, the foil strain gauges were mounted onto auxiliary No. 4 mild steel reinforcement with adequate de-velopment length. These auxiliary bars were gauged in the laboratory. Du-rable adhesives, requiring oven curing, were used.

Following the strain gauge manufac-turer's recommended procedures, each gauge was protected from moisture. The strain-gauged bars were tied onto the bottom transverse reinforcement.

The electrical wires to the internal sensors were numbered and routed to what would be the east end of the girder when the bridge was completed. At this location, a sufficient length of the electrical wire to connect the

gauges to a data acquisition system at the bridge site was coiled and placed inside electrical boxes, which had been secured to the reinforcing cage. The end of each electrical wire was then taken out of the formwork. These elec-trical wire ends were then connected to a data acquisition system.

The instruments were checked dur-ing fabrication to ensure no damage had occurred. Later, instruments were used to measure the heat of hydra-tion during curing and monitor the response of the girders during deten-sioning. Prior to shipping the girders, the extra electrical cable was cut and the area around the electrical cable on the top flange was sealed. After the girders were placed, the instruments could be attached to the data acquisi-tion system by accessing the wires through electrical boxes embedded in the bottom of the girders.

ERECTION OF THE BRIDGE

The bridge was erected in two phases. In Phase I, the old bridge was cut in two sections and traffic was routed onto one-half of the structure. The other half of the old bridge was

Table 4. Truck dimensions.

Lengthbetweenaxles(L)Widthbetweentires(W) Trucknumber[center-to-centeraxles][center-to-centerfronttires] ft m ft m 527 14.58 4.44 6.58 2.00 585 14.58 4.44 6.58 2.00 624 14.58 4.44 6.58 2.00 707 11.50 3.51 7.33 2.24

TotalweightFrontaxleRearaxle Truck number lb kN lb kN lb kN 527 29780 132.5 8040 35.8 21740 97.7 585 27120 120.6 7120 31.7 20000 89.0 624 31680 140.9 8300 36.9 23380 104.0 707 30960 137.8 9300 41.4 21660 96.4

Table 5. Truck weights.

removed and the channel in this area was widened. One-half of each new abutment, consisting of concrete filled tube piles, a pile cap and a beam seat, was constructed at each end.

The first seven girders (see Figs. 2 and 9) were delivered and set on neoprene bearing pads. After the threaded rods were installed, the shear keys were grouted and the longitudinal joints were sealed. The top surface of the Phase I girders was covered with a waterproofing membrane and then a layer of asphalt. Traffic was then routed to the completed half of the new bridge.

In Phase II, the other half of the old bridge was removed and the channel was widened. After constructing the remaining halves of the two abut-ments, Girders 8 to 12 (see Figs. 2 and 9) were sequentially put in place, installing and tightening the threaded rods along the way. However, only the shear keys between Girders 8 to 12 were grouted. The shear key at the construction joint between Girder 7 in Phase I and Girder 8 in Phase II was not grouted because of the constant movement of Girder 7 due to the traf-fic on Phase I half of the bridge.

Without the shear key in place, the vertical displacements of Girder 7 due to the traffic load caused spalling on the bottom edge of Girder 8 where it contacted Girder 7. This occurred because there is a small amount of tension in the transverse rods which is used to pull the girders together during erection (but is not intended as a trans-verse post-tensioning).

To arrest the spalling until the shear key could be placed, the tension in threaded rods between Girders 7 and 8 was relieved. Shortly afterwards, while the traffic was rerouted during the night, the threaded rods between Girders 7 and 8 were retightened and the shear key grouted with a fast set-ting magnesium phosphate grout.

Once the shear key between the two phases of the bridge was in place, the spalling stopped. After sealing the construction joint, the remaining half of the bridge was covered with the waterproofing and asphalt. The entire bridge was then opened to traffic.

TRUCK LOAD TESTING

November-December 2000 67

Two truck load tests were conducted on the bridge. The first occurred when Phase I was complete but before Phase I was opened to traffic. The second test was done after the entire bridge was complete and had been opened to traffic for about 2 weeks.

To measure the vertical deflection of each girder, 21 wire potentiometer displacement transducers were used. During Phase I [with Girders 1 to 7 in place (see Fig. 9)], the deflection at the quarter points and midspans were measured. For the second test with the bridge completed, only the midspan deflections for each girder were re-corded.

Each sensor was mounted onto a small wooden plate, which had been screwed to the bottom face of each girder. The wires from the wire po-tentiometers were attached to concrete blocks anchored firmly in the creek bed directly under each sensor.

The relative displacement between adjacent girders was measured using linear variable differential transformer (LVDT) displacement transducers placed at various locations. The rela-tive displacement was obtained by at-taching a displacement transducer on one girder and targeting it to the adja-cent girder (see Fig. 9).

During Phase I, the relative dis-placements between Girders 2 and 3, 4 and 5 and 6 and 7 at their midspan were measured. During Phase II, the relative displacement at the construc-tion joint between Girders 7 and 8 was measured at the quarter points and midspan.

LOADING CASESUp to four ODOT trucks filled with

gravel were used to load the bridge. The weight and dimensions (longitu-dinal as well as transverse center-to-

center distances between the wheels) were measured for each truck prior to the test (see Tables 4 and 5).

To examine various responses, dif-ferent truck positions were imple-mented. For each test, the trucks were placed as closely as possible to pre-determined locations marked on the bridge. For the Phase I test, seven different load cases with one to four trucks placed on the bridge were used. The different load cases for Phase I are shown in Fig. 10.

Load Cases 1, 2 and 6 were used to induce the maximum moments and de-flection at midspan. These cases were also used to investigate the linearity of the structure. In Load Cases 3 and 4, one truck was placed as far as pos-sible to the edge of the structure to investigate the load transfer across the width of the bridge. Load Case 5 was intended to look for the load distribu-tion between the girders. Load Case 7

Fig. 10. Truck locations for Phase I test.

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represents the maximum lane load for the structure.

The test program for the completed bridge (Phase II) consisted of six load cases (see Fig. 11). Load Cases 1 and 6 were used to induce the maximum moments and deflections at midspan. Load Cases 2 and 3 were maximum lane load in the westbound lane (Case 2) and examined the shear transfer across the shear key that joined Phase I and Phase II construction. Load Case 4 represents the maximum possible lane load for the eastbound lane. Load Case 5 was intended to look into the distribution of the reactions along the west abutment.

In addition, a moving load test was done for the completed bridge (no such test was done in Phase I as the bridge was still under construction and a mov-ing load test was deemed to be too dan-gerous). The trucks traveled over the

bridge with a speed of about 50 miles per hour (80 km/h) and the deflections of the structure were recorded at time intervals of 0.05 seconds.

TEST RESULTSDiscussed below are the major re-

sults regarding deflections, strain dis-tribution and moving load.

Deflections

The maximum deflections for both the Phase I and Phase II tests, as a fraction of span length, L, are summa-rized in Table 6. For the Phase I tests, the maximum deflection occurs for Load Case 6. Using an elastic analy-sis, it was determined that this loading would provide stresses and deflections roughly equivalent to having an HS-20 design truck in each lane with no im-pact applied.

Although the bridge was designed for an HS-25 loading (which is 25 per-cent larger than the HS-20), the ODOT standard is HS-20 and this was used as the maximum test load. The measured deflection is about half of the L/1000 (where L is the span length) permitted by AASHTO Standard Specifications. It is still clear that the girders have adequate stiffness and, therefore, de-flection is not a concern regardless of whether the HS-20 or HS-25 design loading is used.

A similar argument can be made for the completed bridge (Phase II). In this case, the maximum deflection occurs under Load Case 1 (which is similar to Load Case 6 in the Phase I test). The maximum deflection is about 35 per-cent of the allowable, so the bridge is adequate under either loading.

The deflection profiles in the trans-verse direction show that all the gird-

Fig. 11. Truck locations for Phase II test.

November-December 2000 69

ers are working together, and the transverse deflection profiles are very similar to that which would be ex-pected in a slab bridge (see Fig. 12). More uniform load cases show almost uniform deflection while load cases where the trucks are closer to one side show larger deflections to that side.

Strain Distribution

Fig. 13 shows the strain distribution over the depth of the girders for Load Case 6, Phase I. As shown, the strain is almost identical in all the girders. The strain is also very close to the

Maximumdeflection Loadcase PhaseItest PhaseIItest 1 L/3310 L/2790 2 L/63409 L/3270 3 L/6610 L/3410 4 L/6550 L/2910 5 L/3410 L/11570 6 L/1760 L/2975 7 L/2070 –

Table 6. Maximum deflection.

Fig. 12. Transverse vertical deflection profiles – Phase II test.

Fig. 13. Strain distribution through height of girders, Load Case 6, Phase I test.

value which would be computed using a lumped girder model; i.e., treating the bridge as a single girder with the moment of inertia equal to the mo-ment of inertia of a single girder times the number of girders.

The distribution of the strains for the girders also supports the previ-ous observation that the girders are working together as a system. The strains in the top and bottom flanges are almost equal in values; hence, the

girders are essentially behaving linear elastically.

Based on the measured strains, the live load tensile strain is about 90 mi-crostrain. Using the average measured modulus of elasticity, the correspond-ing live load stress is about 520 psi (3.6 MPa). For Phase II, the maximum stress (again for Load Case 6) was about 60 microstrain and the calcu-lated live load stress would be approx-imately 350 psi (2.4 MPa).

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The measured relative displacement between adjacent girders was found to be less than the specified accuracy of the transducers. Clearly, no appre-ciable movement between the girders was measured. The shear keys and the threaded rods were adequate to ensure that the girders work as a system.

Moving Load

To measure the dynamic response, deflections at the midspan of Girder 7 were measured with ODOT trucks traveling across the bridge at about 50 miles per hour (80 km/h). The re-sponse is shown in Fig. 14. Based on the measured attenuation of the vibra-tional wave, the damping of the bridge can be calculated to be between 1.5 and 2 percent using the logarithmic decrement. This is a reasonable value for this type of structure.

The dynamic deflection under one truck is approximately equal to the value when the bridge was loaded statically with one truck. Truck 4 followed approximately 20 ft (0.61 m) directly behind Trucks 3 when they passed over the bridge. The deflection is almost doubled, as ex-pected. There is no apparent dynamic amplification.

There was one final interesting observation. When members of the

research team were walking across the bridge, they could not feel any vibration even when large trucks were crossing the bridge. To feel the bridge vibrate, the observer had to stand per-fectly still at midspan, and even then the vibration did not seem excessive. Fig. 15 shows the completed bridge.

CONCLUSIONSIn this project, the design for an ad-

jacent box girder bridge was changed from three spans to a single span. This was done by extending the maximum length of ODOT’s largest box section through the use of HPC and 0.6 in. (15 mm) diameter strands. The per-formance of the bridge was verified through testing of prototype girders and truck load testing of the actual bridge.

The results of the investigation lead to the following conclusions:

1. Testing of the prototype girders showed that the girders were suffi-ciently strong and ductile. Cracking strengths were accurately predicted only if the higher tensile strength of the HPC was taken into account. Ul-timate strength was adequately pre-dicted using the provisions of the AASHTO Standard Specifications. Common analytical methods could predict overall girder behavior.

Fig. 14. Dynamic response to moving trucks.

2. Loss of prestressing force was measured in the prototype girders and found to be about 18 percent at 6 to 9 months. The AASHTO Standard Spec-ifications predict a loss of 20 percent (calculated at time equals infinity).

3. The bridge was constructed in two phases – Phase I where Girders 1 to 7 were erected and Phase II where the remaining five girders were added. A truck load test was conducted after the completion of each phase. These tests consisted of using up to four dump trucks, each weighing approximately 30 kips (133 kN), placed on the bridge in different configurations. The test results showed:

(a) The bridge did not deflect exces-sively under load. Maximum static deflections were 35 to 50 percent of that allowed by the AASHTO Standard Specifications.

(b) The maximum live load stress (computed from measured strain) under static loading was about 500 psi (3.5 MPa). From elastic analysis, the stress caused by an HS-20 truck (without impact) would be 550 psi (3.8 MPa), so the measured value is in reason-able agreement with design cal-culations.

(c) Under dynamic load, the bridge did not vibrate excessively. The

November-December 2000 7�

damping was calculated to be about 1.5 to 2 percent of critical damping, reasonable for this type of structure. Dynamic deflections were about equal to the static de-flections and there was no appar-ent dynamic magnification.

ACKNOWLEDGMENTSThis project was funded by the Fed-

eral Highway Administration through the Ohio Department of Transporta-tion as part of the FHWA HPC Bridge Showcase Program. The University

of Cincinnati Research Team grate-fully acknowledges the following people: Terry Halkyard, James Hoblit-zell, Suneel Vanikar and Susan Lane of FHWA; Brad Fagrell, Vik Dalal, Roger Green and Lloyd Welker of ODOT; and Charles Goodspeed of the University of New Hampshire.

The authors are grateful for the excellent assistance of the fabrica-tors, Don Bosse, Joe Roche and Jack McDonald of Prestress Services of Melbourne, Kentucky. The authors also want to express their gratitude to the PCI JOURNAL reviewers for

their thoughtful and constructive comments.

The contents of this paper reflect the views of the authors, who are solely responsible for the accuracy of the facts, data and interpretations presented herein. This paper does not necessarily reflect the policies or offi-cial views of the Federal Highway Ad-ministration or the Ohio Department of Transportation and this paper does not constitute a standard, regulation or specification.

Fig. 15. Completed bridge.

REFERENCES

1. Long, Ellen, "High Performance Concrete Mix Design for an Adjacent Box Girder Bridge," A Thesis Submitted in Partial Fulfillment of the Requirements for the Degree of Master of Science, Department of Civil and Environmental Engineering, University of Cincinnati, Cincinnati, OH, March 1998.

2. AASHTO, Standard Specification for Highway Bridges, 16th Edition, American Association of State Highway and Trans-portation Officials, Washington, DC, 1998.

3. Miller, R. A., Hlavacs, G., Long, T., and Greuel, A., “Full Scale Testing of Shear Keys for Adjacent Box Girder Bridges,” PCI JOURNAL, V. 44, No. 6, November-December 1999, pp. 80-90.

4. Miller, R., Shahrooz, B., Baseheart, T. M., Eberenz, E., Jones, J., Knarr, R., Sprague, R., "Testing of a High-Performance Concrete, Single-Span Box Girder,” Transportation Research Record, No. 1624, 1998.

5. AASHTO, LRFD Bridge Design Specifications, Second Edition, American Association of State Highway and Transportation Officials, Washington, DC, 1998.

6. Russell, B., and Burns, N., “Measured Transfer Lengths of 0.5 and 0.6 in. Strands in Pretensioned Concrete,” PCI JOUR-NAL, V. 41, No. 5, September-October 1996, pp. 44-63.

7. Collins, M., Mitchell, D., Felber, A., and Kuchma, D., RE-SPONSE, V. 1, in Prestressed Concrete Structures, Prentice Hall, Englewood Cliffs, NJ, 1991.

8. PCI Design Handbook, Fifth Edition, Precast/Prestressed Con-crete Institute, Chicago, IL, 1999.