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9h Australian Small Bridges Conference 2019
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Estimated versus measured capacity of CFA piles for Seaford Road
bridges, Melbourne
Cillian Mc Colgan, Associate Geotechnical Engineer, WSP
Suthagaran Visvalingam, Senior Geotechnical Engineer, WSP
Nicholas Withers, Geotechnical Engineer, WSP
ABSTRACT
This paper discusses the design of piles for a rail bridge in south east Melbourne which adopted a
relatively conventional approach to estimating static axial pile capacity and compares it to construction
stage validation.
On site validation included assessment of pile installation data provided by the piling rigs on board
computer. Some agreement between the interpreted ground profile and the data obtained from the
piling rig outputs was observed.
The results of fifteen dynamic pile tests and two sacrificial (non-working) pile tests are also presented.
These tests demonstrated greater capacity than may be estimated through these conventional
approaches. This presents opportunity to further refine pile design on subsequent stages of the
project leading to a more robust and cost effective design.
1 INTRODUCTION
The removal of level crossings improves safety by separating trains from traffic, with other benefits
such as reduced congestion, an upgraded road network, connected communities and support of
urban regeneration. The current Victorian State Government has initiated a level crossing removal
program to remove at least 75 of these level crossings across Melbourne by 2025. The Southern
Program Alliance (SPA) comprising WSP, Lendlease, Acciona/Coleman Rail, MTM and the Level
Crossing Removal Project (LXRP) is delivering the works associated with the Frankston Line which
includes the removal of four level crossings at Seaford Road in Seaford and at Mascot Avenue,
Bonbeach and Station Street and Eel Race Road in Carrum. The site location is shown in Figure 1.
Figure 1- Seaford Road site location (LXRP, https://levelcrossings.vic.gov.au/projects )
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At Seaford in south east Melbourne, two parallel U trough girder bridges have been constructed with
combined spans of more than 110m. These bridges convey the Frankston Line over a shared user
path and Seaford Road. These bridges are founded on groups of Continuous Flight Auger (CFA)
piles.
A photograph of the completed bridge is reproduced at Figure 2 below.
Figure 2- Seaford Road bridge (LXRP, Image Gallery of Seaford Road, Seaford)
2 GEOTECHNICAL MODEL
The ground conditions at Seaford consist of surficial fill, between 3 m to 5 m of quaternary aged
sands and swamp deposits overlying Tertiary aged Brighton Group (now officially known as the
Sandringham Sandstone) and Gellibrand Marl Units These conditions are typical of South East
Melbourne.
The salient features of these materials, as they relate to the Seaford bridges are summarized in Table
1. The geological long section and a detail of the Southern Abutment of the Seaford Road Bridge are
presented in Figure 2 and Figure 3.
Older Volcanics are also encountered at Seaford but these underlie the Gellibrand Marl and were not
considered in the bridge foundation design due to their depth.
Table 1: Geotechnical units relevant to Seaford Road Bridge
Geological Unit and ID Material Description Consistency Depth to top (m bgl)
Unit 1 Fill Shallow fill associated with
rail formation
N/A 0.01
Unit 2 2A Quaternary
Sands
Dune Sands Medium
Dense
0.0 – 1.0
2B Swamp
deposits
Sands with occasional layers
of compressible back swamp
materials
Loose 2.9 – 6.51
Unit 3 3A Brighton Group Sandy Clay Stiff to Hard 4.1 – 8.0
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3B (Sandringham
Sandstone)
Clayey Sand, occasionally
cemented
Medium
Dense to
Very Dense
10.0 – 16.5
Unit 4 Gellibrand Marl Stiff to Very Stiff Stiff to Very
Stiff
19.0 – 22.6
Notes:
1. Unit not present across all pier/abutment locations
Figure 3- Geological long section for the Seaford Bridge, Seaford
3 AXIAL CAPACITY ESTIMATION AND PILE DESIGN
Table 2 presents the adopted design parameter ranges for the different units including the ranges of
Standard Penetration Test (SPT) “N” values recorded in the geotechnical investigations.
Table 2: Geotechnical design parameters and SPT results
Geological Unit ID
Adopted SPT “N” Value
Adopted Design Parameters
Undrained
Shear
Strength, Su
(kPa)
Effective
Cohesion, c’
(kPa)
Effective
Friction
Angle, ’
(Degrees)
Ultimate
Unit Skin
Friction, fs
(kPa)
Ultimate Unit
End Bearing
Friction, fb
(kPa)
Unit 2A 16 N/A 0 29° – 32° 15 - 24 N/A
Unit 2B 0 - 23 N/A 0 28° – 33° 4 - 32 N/A
Unit 3A 12 - 33 40 - 200 - - 16 - 65 1100 - 3600
Unit 3B 12 - 33 N/A 0 30° – 36° 16 - 65 1100 - 3600
Unit 4 14 - 25 125 - 175 0 28° – 33° 39 - 43 700 – 1200
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Shaft friction and end bearing was limited to the values recommended by Decourt, L., (1995). The
average skin friction was assessed based on O’Neill & Reese (1999) for cohesive soils and Craig
(2004) for granular soils. The ultimate bearing capacity was assessed based on Fleming et. al (2009)
for cohesive soils and Berezantzev (1961) for granular soils.
Based on the above interpretation of the ground conditions and the provided loading, pile lengths
were assessed and are presented in Table 3 below. Note that the design geotechnical strength is
provided for each pile which was based on a geotechnical strength reduction factor (g) of 0.72. This
was adopted based on undertaking a dynamic load test at each pier and abutment location (i.e. 8
piles out of a total of 80 piles for the bridge foundation). The strength presented in Table 3 is also the
load that needed to be achieved in the dynamic testing.
Table 3: Pile design summary
Bridge ID Foundation Location
Pile Diameter
(mm)
Required Ultimate Capacity at Pile Head1
(kN)
Pile Length (m)
Founding Material
Pedestrian
Access Bridge
Abutment A 1050 3740 20.0 Stiff Clay
Abutment B 1050 3740 19.5 Very Stiff Clay
Seaford Road
Bridge
Abutment C 1050 3740 19.0 Very Stiff Clay
Pier 1A/1B 1050 3650 18.0 Very Stiff to Hard Clay
Pier 2A/2B 1050 3650 19.5 Stiff to Very Stiff Clay
Abutment D 1050 3740 19.0 Very Stiff Clay
Notes:
1. This does not include pile self weight and is also the pile test load
2. The capacity of the pile is measured after the pile is cast so the pile self weight during the
test, thus this can be added to the result to get the piles real capacity
4 CONSTRUCTION VALIDATION
The primary method of validation of the design was through testing piles at each bridge support
location in accordance with the requirements of VicRoads Specification 607. Periodic inspections of
the drilling operations were also carried out along with a review of the pile installation records
provided by the Continuous Flight Auger (CFA) piling rig.
4.1 Pile installation records
The piles were installed using a LB28 piling rig with a Jean Lutz on board computer. The onboard
computer measures a number of different operation parameters including drill pressure, intrusion rate
and pile radius. The latter is calculated by comparing the volume of concrete delivered with the rate of
extraction of the auger.
An example of the computer outputs compared to a borehole log is reproduced in Figure 4. Although
somewhat anecdotal, there does not appear to be a very convincing correlation between drill pressure
and material type. This was found to be consistent across all the pile installation records.
It is the opinion of the author that this data does not provide a reliable means of validating ground
conditions for floating piles. The assessment undertaken was focused on verifying that conditions
were not significantly worse than assumed in the design, i.e. no significant drop off in drill pressures
were observed and a general trend of drill pressure increasing with depth was present and the as built
pile diameter achieved was in accordance with the design.
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Figure 4- Pile installation records compared to adopted geotechnical ground model
4.2 Inspection of Pile Cuttings
Inspecting the pile cuttings was carried out at a nominal frequency of about 2 piles per group. A plan
showing the pile caps for Piers 2A and 2B and Abutment D is reproduced at Figure 5 below.
Figure 5- Plan illustrating pile layouts
Pier 2A/B
Abutment D
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There are technical difficulties associated with trying to assess the materials a CFA pile is founded in
by logging auger cuttings. These include:
• mixing of soils during drilling making it difficult to assess what level the soils on the CFA auger
come from
• disturbance of soils during drilling affecting their consistency, especially for the sandy silts which
were the soils this bridge was founded in.
There is also a logistical disadvantage to not knowing the founding material you are in until
completion of pouring the pile. Its essentially too late to do anything other than re-excavation of the
already poured pile.
Where floating CFA piles are proposed geotechnical investigations should be carried out at an
appropriate density. Relying on validating the materials during construction is not practical.
Complementing the investigations with pile testing is a more robust way of validating the design and
pays dividends through increased φg. and leading to a more robust and cost effective design.
4.3 Dynamic Pile Testing – Proof Tests
A purpose-built test frame equipped with a 12T drop hammer was used to test the nominated piles.
Drop heights of between 0.8 and 1.5 m were used which equates to a test energies of between 94
and 176 kJ.
Table 4 below shows the results of the pile testing versus the required test load. Capacities far in
excess of that required by the design were achieved. All piles demonstrated the required test loads on
the first attempt. Piles were left to set up for a minimum of 6 days prior to test.
The capacities demonstrated below were done so at relatively modest test energies. The testing was
not specified to demonstrate the full capacity of the piles, rather the required capacity.
Table 4: Dynamic pile load test results of Seaford Bridge foundations
Bridge Foundation Location
Estimated Pile Capacity1
(kN)
Measured Pile Capacity2 (kN)
Estimated Shaft Force (kN)
Measured
Shaft
Force
(kN)
Estimated
End
Bearing
Force
(kN)
Measured
End
Bearing
Force
(kN)
Abutment A - P09 3746 9575 3221 6112 925 3463
Abutment B - P04 3751 6487 2934 5188 1195 1299
Abutment C - P03 3741 9303 2764 6273 1360 3030
Pier 1A - P01 3651 8869 2653 7138 1359 1731
Pier 1B - P01 3651 7746 2653 6231 1359 1515
Pier 2A - P02 3641 7513 3206 5781 820 1731
Pier 2B - PP08 3641 7336 3206 5605 820 1731
Abutment D –
P04
3745 7323 1927 5592 2098 1731
Notes:
1. The proportion of the piles capacity used up by self weight has been omitted for ease of
comparison to the pile test result
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2. Measured capacity above pile cut off level has been subtracted from the overall test result
3. The capacity of the pile is measured after the pile is cast so the pile self weight is active
during the test, thus this can be added to the result to get the piles real capacity
4. Pile diameter for all the bridge foundation is 1050 mm
5. Pile length for the bridge foundation location is as per the Table 3
4.4 Dynamic Pile Testing – CFA rigid inclusions
The bridge approach embankments were founded above potentially compressible soils and as such
some ground improvement was required to manage settlement risk. This took the form of unreinforced
“rigid inclusions” constructed as unreinforced CFA piles. Like the CFA piles these demonstrated much
higher strengths than would be calculated through conventional means. These were tested with the
same apparatus but at lower energies than the CFA piles. Test results are presented in Table 5.
Table 5: Dynamic pile load test results of CFA rigid inclusion
Rigid Inclusions Location
Measured Pile Capacity1
(kN)
Diameter
(mm)
Length
(m)
Measured
Shaft Force
(kN)
Estimated End
Bearing Force
(kN)
AD87 2427 500 8.6 1691 736
AB32 1592 500 8.6 840 742
AA06 1989 500 10.6 1351 638
PN59 1200 500 8.4 660 540
PN70 4879 600 17.0 3041 1838
PN61 4520 600 17.0 3177 1343
PN55 5641 600 17.0 3945 1696
Notes:
1. Measured capacity includes working platform material
4.5 Dynamic Pile Testing – Destructive Tests
Before the completion of the pile testing it was recognized that significantly higher pile capacities than
assumed in the design were being achieved. Also, the tests on the rigid inclusions were succeeding in
demonstrating relatively high end bearing at shallower depths. Two destructive tests were undertaken
at Seaford to try to investigate if even higher capacities could be demonstrated. This investment
(approximately $50,000) was justified against the potential savings that could be realized on a bridge
to be designed in similar materials later that year.
Table 6: Dynamic pile load test results of destructive tests
Location Diameter
(mm)
Length
(m) Measured Capacity1
(kN)
Measured
Shaft Force
(kN)
Measured End
Bearing Force
(kN)
Test Pile
Abutment A2
1050 9.0 3472 1740 1732
Test Pile
Abutment C3
1050 14.0 9040 3412 5628
Notes:
1. Measured capacity includes working platform material
2. Pile for destructive test at Abutment A is founded in Unit 3A Brighton Group Sand
3. Pile for destructive test at Abutment C is founded in Unit 3B Gellibrand Marl
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5 ANALYSIS OF PILE TEST RESULTS
5.1 Shaft Friction
The results of the 17 dynamic tests have been compared to SPT testing results to see if the results
correlate. A summary of all of the dynamic pile load test data, for each geological units along with the
SPT test results are presented at Figure 4. The SPT tests were corrected for energy and overburden
stress (for granular soils) using the methodology presented by Skempton (1996).
Table 7 compares the average SPTN1,60 values for each of the units with the measured shaft friction.
Table 7: Summary of shaft friction
Unit N1,60 Shaft Friction (kPa) Correlation
Range Average Range Average
Unit 2 0 – 43 15 50 – 80 65 Fs = 4.3*N1,60
Unit 3A 10 – 71 23 50 - 120 80 Fs = 3.5*N1,60
Unit 3B 0 - 70 33 80 - 150 100 Fs = 3.0*N1,60
Unit 4 10 – 65 21 50 - 150 100 Fs = 4.7*N1,60
Broadly speaking there does appear to be good correlation between N1,60 and the measured shaft
friction in the pile tests for the Brighton Group Materials (Unit 3A and 3B) and the Quaternary Sands
(Unit 2). A general trend of increasing shaft friction and N1,60 with depth is clear.
The Gellibrand Marl (Unit 4) doesn’t follow this trend. It is difficult to assess exactly why the data for
the Gellibrand Marl shows such high shaft friction. The Gellibrand Marl was generally logged as
cohesive and the undrained strengths recorded were relatively low when compared to the measured
shaft friction. It could be to do with the use of the SPT as the primary form of characterizing these
soils. Silts, especially when they are sandy like we found at Seaford, tend to dilate. It is easy to
imagine this leading to the development of local excess pore pressures when they are subjected to
the first blow from an SPT. This could be responsible for the relatively low values of N1,60.
The pile test however occurs some 6 days after the pile has been set. This provides time for excess
pore pressures that may have been generated during pile installation to dissipate and for the silts to
move towards there equilibrium condition. Allowing piles to set up before testing is a common
practice. We do not do this, however, when conducting in-situ testing on soils that may be subject to
dilation.
The outcome of the above is two fold, the correlations may not be appropriate for the Gellibrand Marl
given doubts around the applicability of SPT testing. Other, slower (and more expensive) means of
testing in-situ strength may be better capable of characterizing the Gellibrand Marl but these are
expensive and come with their own challenges.
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Figure 6- Pile test results compared to SPT N1,60
0 2,000 4,000 6,000 8,000
-22
-20
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
0 40 80 120 160 200
Unit End Bearing (kPa)
RL
(m A
HD
)
Unit Skin Friction (kPa)
Unit 2 Unit 3A Unit 3B Unit 4
-22
-20
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
0 10 20 30 40 50 60
RL
(m A
HD
)
SPT N1,60
Unit 2A Unit 2C Unit 3A
Unit 3B Unit 4
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5.2 End Bearing
Table 8 compares the average SPTN1,60 values for each unit with the measured end bearing values.
Table 8: Summary of end bearing
Pile Location ID Geological Unit
SPT N1,60 at pile toe1
Measured End Bearing, Fb (kN)
Correlation Between Fb and SPT N1,60
Test Pile at Abutment C Unit 3A 26 2000 Fb = 77 x N1,60
AD87 Unit 3A 18 3750 Fb = 208 x N1,60
AB32 Unit 3A 18 3750 Fb = 208 x N1,60
AA06 Unit 3A 68 3680 Fb = 54 x N1,60
PN59 Unit 3A 27 6000 Fb = 222 x N1,60
Averages for Unit 3A 33 4295 Fb = 131 x N1,60
Test pile Abutment A Unit 3B 29 6500 Fb = 224 x N1,60
PN70 Unit 3B 30 6500 Fb = 217 x N1,60
PN61 Unit 3B 30 4750 Fb = 158 x N1,60
PN55 Unit 3B 30 6000 Fb = 200 x N1,60
Averages for Unit 3B 30 5938 Fb = 200 x N1,60
Abutment A - P09 Unit 4 15 4000 Fb = 267 x N1,60
Abutment B - P04 Unit 4 18 1500 Fb = 83 x N1,60
Abutment C - P03 Unit 4 22 3500 Fb = 159 x N1,60
Pier 1A – P01 Unit 4 36 2000 Fb = 56 x N1,60
Pier 1B – P01 Unit 4 36 1750 Fb = 49 x N1,60
Pier 2A – P02 Unit 4 13 2000 Fb = 154 x N1,60
Pier 2B – PP08 Unit 4 13 2000 Fb = 154 x N1,60
Abutment D Unit 4 16 2000 Fb = 125 x N1,60
Averages for Unit 4 21 2344 Fb = 111 x N1,60
Notes: 1. Nominally includes SPT tests 2 diameters above and 3 diameters below toe
There is a considerable amount of scatter in the data. This is likely due to the variation in energy
during testing and the ranges of pile sizes considered. The smaller piles founded in Units 3A and 3B
from CMC’s show higher end bearing than the larger piles in Gellibrand Marl. This is more likely
associated with the difficulty of mobilizing end bearing in large and long end bearing piles than a
difference in strength. It is logical to expect that more end bearing could be available in the Gellibrand
Marl based on the relatively high amounts of shaft friction mobilized.
The destructive test at Abutment A does show that these capacities are possible for larger piles.
5.3 Test Energy
Figures 6 to 9 present the range of measured skin friction versus test energy. It appears from the data
that the test energies were generally sufficient to mobilise all the available shaft friction for the piles
tested in the upper materials. The Gellibrand Marl may be capable of demonstrating more shaft
friction at higher test energies as suggested by Figure 9.
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Figure 7- Test energy versus measured skin friction – Unit 2
Figure 8- Test energy versus measured skin friction – Unit 3A
20
30
40
50
60
70
80
90
100
110
120
130
140
150
160
0 50 100 150 200 250 300 350 400
Mea
sure
d S
kin
Fri
ctio
n (
kPa)
Applied Energy (kJ)
Unit 2 - Quaternary Sand & Swamp Deposits
AA-09 - U2 AB-04 - U2 AC-03 - U2 AD-04 - U2 CMC-AA-06 - U2 CMC-AB-32 - U2CMC-PN-59 - U2 CMC-AD-87 - U2 CMC-PN-55 - U2 CMC-PN-61 - U2 CMC-PN-70 - U2 P1-A1 - U2
20
30
40
50
60
70
80
90
100
110
120
130
140
150
160
0 50 100 150 200 250 300 350 400
Mea
sure
d S
kin
Fri
ctio
n (
kPa)
Applied Energy (kJ)
Unit 3A - Brighton Group - Sandy Clay
AA-09 - U3A AB-04 - U3A AC-03 - U3A AD-04 - U3A CMC-AA-06 - U3ACMC-AD-87 - U3A CMC-PN-55 - U3A CMC-PN-61 - U3A CMC-PN-70 - U3A P1-A1 - U3A
2 results
2 results
3 results
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Figure 9- Test energy versus measured skin friction – Unit 3B
Figure 10- Test energy versus measured skin friction – Unit 4
20
30
40
50
60
70
80
90
100
110
120
130
140
150
160
0 50 100 150 200 250 300 350 400
Mea
sure
d S
kin
Fri
ctio
n (
kPa)
Applied Energy (kJ)
Unit 3B - Brighton Group - Clayey Sand
AB-04 - U3B AA-09 - U3B AC-03 - U3B AD-04 - U3B
CMC-PN-55 - U3B CMC-PN-61 - U3B CMC-PN-70 - U3B P1-A1 - U3B
P1-B1 - U3B P2-A2 - U3B P2-B8 - U3B TP-AA - U3B
20
30
40
50
60
70
80
90
100
110
120
130
140
150
160
0 50 100 150 200 250 300 350 400
Mea
sure
d S
kin
Fri
ctio
n (
kPa)
Applied Energy (kJ)
Unit 4 - Gellibrand Marl
CMC-PN-55 - U2 CMC-PN-61 - U2 CMC-PN-70 - U2 P1-A1 - U2 P1-B1 - U2
P2-A2 - U2 P2-B8 - U2 AA-09 - U4 AC-03 - U4 AD-04 - U4
2 results
3 results
2 results
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Figures 10 to 14 show the relationship between the applied test energy and the mobilized end bearing
resistance for the different units encountered at Seaford. There is a clear trend of an increase in the
measured end bearing with an increase in the test energy. It can also be seen that the smaller
diameter piles (CFA Rigid Inclusions) that were tested we capable of demonstrating more end bearing
at lower energy, as would be expected.
Figure 11- Test energy versus measured end bearing for all units
Figure 12- Test energy versus measured end bearing – Unit 4
0
500
1000
1500
2000
2500
3000
3500
4000
4500
5000
5500
6000
6500
7000
0 50 100 150 200 250 300 350 400Mea
sure
d E
nd
Bea
rin
g(kP
a)
Applied Energy (kJ)
All Units
AA-09 - U4 AB-04 - U4 AC-03 - U4 AD-04 - U4 P1-A1- U4
P1-B1 - U4 P2-A2 - U4 P2-B8 - U4 TP-AA - U3B TP-AC - U3A
CMC-AA-06 - U3A CMC-AD-87 - U3A CMC-PN-55 - U3B CMC-PN-61 - U3B CMC-PN-70 - U3B
0
500
1000
1500
2000
2500
3000
3500
4000
4500
5000
5500
6000
6500
7000
0 50 100 150 200 250 300 350 400
Mer
asu
red
En
d B
eari
ng(
kPa)
Applied Energy (kJ)
Unit 4 - Gellibrand Marl
AA-09 - U4 AB-04 - U4 AC-03 - U4 AD-04 - U4 P1-A1- U4 P1-B1 - U4 P2-A2 - U4 P2-B8 - U4
Rigid Inclusions
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6 CONCLUSIONS
In summary the pile test data has shown:
- Conventional design methods can be overly conservative for pile design in the Brighton
Group and Gellibrand Marl Units
- The Brighton Group and Gellibrand Marl Units both displayed skin friction far greater than
conventional design would suggest.
- The Brighton Group units showed a similar trend for end bearing though the data to support
this is not as comprehensive as that for shaft friction and it is limited to smaller diameter piles
- There is likely more end bearing capacity available in the Gellibrand Marl than was
encountered at Seaford bridge
- Potential to prove greater capacities with destructive testing, which may not be limited to
particular geological units.
The results clearly demonstrate that conventional pile design methods are conservative for the
Brighton Group and Gellibrand Marl units. The difference between the test results and the
conventional design approach is striking and should encourage geotechnical engineers to pursue pile
testing more aggressively. It is the authors opinion that in the long term this will result in a net saving
for industry, especially for major infrastructure projects.
The Southern Program Alliance is currently pursuing the aforementioned opportunity with the design
of a larger bridge founded in similar materials and has been able to use the data obtained at Seaford
to achieve a more economical foundation design.
7 ACKNOWLEDGEMENTS
The author would like to thank the Level Crossing Removal Project and the Southern Program
Alliance for permission to present this work.
8 REFERENCES
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Proceedings 5th International Conference SMFE, Paris, Vol. 2, pp. 11-15.
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York, USA, pp. 313-315.
Clarke and Leonard (2004), Regional variations in neo-tectonic fault behaviour in Australia, as they
pertain to the seismic hazard in capital cities, Australian Earthquake Engineering Society 2014
Conference, Nov 21-23, Lorne, Vic.
Fleming W.G.K. (1992), A New Method for Single Pile Settlement Prediction and Analysis,
Geotechnique 42, No. 3, pp. 411-425.
Fleming, K. et. al. (2009), Piling Engineering, 3rd edition, Taylor & Francis Group, New York, USA, pp.
108.
O’Neill, M.W. and Reese, L.C. (1999), Drilled Shafts: Construction Procedures and Design Methods,
Publication No. FHWA-IF-99-025, Federal Highway Administration, Washington, D.C., pp 758.
Poulos, H.G and Davis, E.H. (1980) Pile Foundation Analysis and Design, Rainbow-Bridge Book Co,
Canada, pp. 31.
Decourt L. (1995), Prediction of load - settlement relationship for foundations on the basis of the SPT,
Ciclo de Conferencias International, Leonardo Zeevaert, UNAM, Mexico, pp.85-104.
Skemption, A. W., Standard penetration test procedures and the effects in sands of overburden
pressure, relative density, particle size, ageing and overconsolidation. Geotechnique 36, No 3, 425 -
447