THE EFFECTS OF SLAKING ON THEENGINEERING BEHAVIOR

OF CLAY SHALES

by

Michael Edward Botts

B.S., Auburn University, 1976M.A., Washington University

in St. Louis, 1979

A thesis submitted to theFaculty of the Graduate School of the

University of Colorado in partial fulfillmentof the requirements for the degree of

Doctor of PhilosophyDepartment of Civil, Environmental, and

Architectural Engineering1986

Botts, Michael Edward (Ph.D., Civil, Environmental, and ArchitecturalEngineering)

The Effects of Slaking on the Engineering Behavior of Clay ShalesThesis directed by Associate Professor Stein Sture

Numerous foundation and slope stability problems worldwide have been contributed tothe presence of clay shale deposits. Present experimental and theoretical methods usedin geotechnical engineering practice are inadequate for assessing the stability of clayshales. The major difficulties with clay shales are attributed to two properties: They areintermediate in behavior between rock and soil, and they tend to transgress from rock-like to soil-like materials within relatively short time-frames.

Evidence suggests that softening along fissures is important to the rapid loss of strengthin clay shales. Yet, geotechnical literature is surprisingly devoid of studies concerningtheoretical or experimental aspects of fissure deterioration in clay shales. Additionally,clay shales invariably exhibit a strong tendency to slake, or disperse, during rewetting, aproperty which certainly contributes to the softening of clays shales. However, fewreported studies have investigated the slaking resistance of geological materials underconditions of confinement found in the field, while no reports were found whichinvestigate the effects of slaking on strength or strain behavior.

The research presented in this dissertation provides initial theoretical and experimentalassessments of the effects of fissure deterioration on the engineering behavior of clayshales. In particular, the drastic decreases in strength resulting from slaking are examinedin detail.The author introduces the possibility incorporating the effects of slaking into the criticalstate soil mechanics model. The role of slaking is seen in this context as simply anotherpath for altering the water content of the clay shale, while the strength envelop remainsconstant.

Laboratory experiments involving triaxial compression tests were performed on Pierreshale samples which had undergone various degrees of slaking under confinement of 10psi. The results from these tests show very significant reductions in strength (up to80%), resulting from a single slaking cycle. The shear strength data from both unaltered

and softened samples were successfully normalized using critical state concepts, anddisplayed a well-defined two-segment failure envelope. This suggests that drastic lateraland temporal variations in the strength of clay shale deposits, which result from slaking,might be accounted for using the critical state approach.

ACKNOWLEDGEMENTS

Seldom is anything accomplished without the assistance or encouragement of others.This dissertation is no exception. Like a rebellious teenager, I felt a need to pursue thisresearch in my own way. Yet, without the advice and encouragement of Dr. Stein Sture,there were times when this research might have gone no further. I thank him for hisseemingly infinite patience, and for always being there when I needed him the most.Likewise, I owe much to the fatherly advice of Dr. Hon-Yim Ko; his deep understandingof both research, and people, helped me through some difficult times.

The ideas for this research were inspired during brief studies of the clay shale problemsin Italy. I thank Dr. Robert Schiffman at the University of Colorado, and ProfessorsArturo Pellegrino, Arrigo Croce, and Carlo Viggiani at the Universitá di Napoli, forproviding this opportunity. Special appreciation is expressed for my dear friend andcolleague, Professor Luciano Picarelli, whose enthusiasm for this research has been asource of much of my own enthusiasm and inspiration. In addition, the encouragementand advice of Dr. Bob Fleming of the U.S.G.S., during the initial stages of this research,were invaluable, as has been the advice of my committee members, Drs. BernardAmadei, Dobroslav Znidarcic, and William Braddock, during the later stages. I wouldalso like to thank Dr. Nicolas Costes, of NASA Marshall Space Flight Center, for havingthe patience and enthusiasm required to introduce a geologist to the strange world of soilmechanics.

Of course, no one could survive the perils of graduate school without the friendship ofthe students who share in the misery and joys. Thanks to Steve Ketcham, who eased mytransition to Boulder and forced me to have fun in spite of myself, and particularly toAtes and Canan Ontuna, Roberto Azevedo, Izabel Duarte, Vincenzo Pane, SebastianoPerriello-Zampelli, Paolo Croce, and Dan Egging, all of whom gave me something I willkeep the rest of my life. No less appreciation is reserved for Bob Scavuzzo, EmirMacari, Trond Mageli, and all the others who gave me a smile from day to day.

Much appreciation goes to Mom and Dad, and the rest of my family; their support andbelief in me have always inspired me in my endeavors. Last, but certainly not least, Ithank my wife, Mary Lynn, whose love gave me the strength and encouragement to

follow the path that I felt was right, and whose patience and unselfishness allowed methe freedom to complete the task. This dissertation is lovingly dedicated to her.

CONTENTS

CHAPTER

I. INTRODUCTION 1

II ENGINEERING PROBLEMS ASSOCIATED WITH CLAY SHALEDEPOSITS 5

Characteristic Problems Related to Clay Shale DepositsAround the World 6British clay shales 6Clay shales of North America 9Clay shales along the Panama Canal 15Italian clay shales 18Residual Versus Softened Strength 24Comparative Summary and Discussion 26

III. SOFTENING MECHANISMS IN CLAY SHALE DEPOSITS 35Softening Mechanisms 36The importance of fissures 36Equilibration of negative pore water pressures 36Fissure deterioration model 38Progressive failure mechanism 41Interactions between softening mechanisms 44Slaking in Clay Shales 47The slaking process . 47Testing for susceptibility to slaking 49Summary of slaking in clay shales 50

IV. THEORETICAL ASPECTS OF FISSURE SOFTENING IN CLAYSHALES 53Model for Fissure Softening 53Implications of the Fissure Softening Model 56The shear strength of jointed masses 56The shear strength of a single, clean, filled, or softened joint 56

The strength of a clay mass consisting of a soft matrix surroundingstiff, intact cores 61The effects of progressive fissure deterioration on the engineeringbehavior of clay shales 61

V. THE ROLE OF SLAKING WITHIN THE CRITICAL STATE MODEL 68Review of the Principles of Critical State Soil Mechanics . 68The Role of "Aging" Within the Critical State Concept 77Slaking within the Critical State Concept 80Possible Complications Resulting from the Fissured Nature of ClayShales 85Testing the Critical State Model with Regard to Slaking 86Summary on the Role of Slaking within the Critical State Model 88

VI. EXPERIMENTAL METHODOLOGY 90Testing Apparatus 91Hoek cell 91Conventional triaxial cell 92Sample Preparation 97Block sample retrieval 97Sample coring and trimming 97Sample drying and the inducement of fissures 100Test Procedure 101General considerations 101Initial conventional triaxial tests 102Improved conventional triaxial tests 105Water content measurements 108

VII. TEST RESULTS AND ANALYSIS . 110Volumetric Changes During Drying and Wetting 110Drying 110Rewetting 113Summary of drying and wetting data 117Stress-Strain Response During Shear 119Axial stress-strain response 120

Volumetric strain during shear 130The Shear Strength of Progressively Softening Pierre Shale 132The Effect of Sample Orientation on the Strength of Pierre Shale 136Summary of Test Results 139

VIII. EVALUATION OF TESTS RESULTS WITHIN THE CONCEPT OF CRITICAL STATE MECHANICS 143

Assessment of the Effects of Slaking Within the Critical StateConcept 143Obtaining the normalization parameters 145Normalization of data 148Normalized strength for "high-angle" tests 153Discussion on Slaking and the Shape of Failure Surfaces forAnisotropic Material 155The effects of fissuration and anisotropy on the shape of failuresurfaces 155Slaking and the intensity of natural remolding 160Slaking and the Critical State Concept:Recommendations for Further Studies and Practical Applications 163Practical applications 163Recommendations for future studies 167

VII. CONCLUSIONS 173

VIII. BIBLIOGRAPHY 178

APPENDIXA. CLASSIFICATION SCHEMES 189B. PROCESSES ACTING DURING THE FORMATION OF CLAY

SHALES 204C. THE MECHANICS OF SINGLE FISSURES 217D. THE STRENGTH OF A CLAY MASS CONSISTING OF

INTACT CORES SURROUNDED BY A MUD MATRIX 228E. PRESENTATION OF TEST RESULTS 232

FIGURES

2.1. Tentative relationship between average shear strength along slipsurfaces and time, for cuttings & retaining walls in London clay. 07

2.2. Relationship between cohesion intercept and time between con-struction and failure. 07

2.3. Failure envelope for London clay based on back-analysis of first-time slides. 09

2.4. Location map for major dams built in clay shales in the NorthCentral United States and Canada. 12

2.5. Proposed clay shale design strength parameters for Bearpawshale at Gardner Dam. 14

2.6. Summary of the shear strength data for the Cucaracha clay shale. 172.7. Map of the landslide areas and their relationship to underlying

rock type in Italy. 192.8 Relationship between unconfined compression strength and water

content for different degrees of weathering in the Laga formation, Italy. 222.9 Variation of shear strength in an Italian clay shale as a function

of time of storage in a humidity room. 232.10. Shear strength envelopes for undisturbed core samples of firm and

weathered Lugagno clay shale. 253.1. Definition of Bishop's average pore pressure ratio. 373.2. Contours of shear strength in a slope and its relation to corresponding

stress states in an overconsolidated clay. 423.3. Schematic illustrating complex action reaction paths possible in clay

shale materials. 453.4. Modes of swelling associated with simple rebound and slaking in an

overconsolidated clay or clay shale. 514.1. Schematic of the proposed model for progressive deterioration of

clay shales. 544.2. Photo showing joint deterioration in the Pierre shale of

South Dakota. 554.3. Schematic illustrating various forms for joints. 584.4 Schematic illustrating the progression of softening in undulating and

Planar joints. 59

4.5 Schematic of a highly altered clay shale consisting of stiff lumpsof intact clay within a matrix of highly weathering clay. 62

4.6 Schematic illustrating possible changes in the failure surfaceresulting from progressive softening of a clay shale. 62

4.7 Schematic illustrating changes in the stress-strain behavior of aclay shale undergoing progressive softening as illustrated inFig. 4.6. 65

4.8 Schematic illustrating changes in the strength envelope resultingfrom progressive softening of a clay shale as illustrated in Fig. 4.6. 66

5.1. Ultimate failure points for drained and undrained tests on normallyConsolidated specimens of Weald clay. 70

5.2. Stress paths in (a) Q':P' and (b) v:P' space for undrained tests onnormally consolidated samples. 71

5.3. Stress paths in (a) Q':P' and (b) v:P' space for drained triaxial testson normally consolidated samples. 72

5.4. The critical state line in v:ln P' space. 735.5. Stress path followed in a drained triaxial compression test on an

overconsolidated clay 735.6. Normalized failure surface for drained and undrained tests on

overconsolidated samples of Weald clay. 745.7. Schematic of expected undrained test paths for samples at different

overconsolidation ratios. 755.8. The complete boundary surface in three dimensions ; Q':P':v space 765.9. The drained path in Q':P':v space. 775.10 Failure states of drained tests on samples at different overconsolidation

Ratios. 785.11 Predicted failure points for overconsolidated and normally consolidated

Clays. 795.12 Geological history and compressibility of normally consolidated

clays, showing effects of "aging". 805.13 Changes in the undrained shear strength ratio and the consolidation

pressure ratio versus the plasticity index for aged" and "young"normally consolidated clays. 81

5.14 Normalization of the undrained shear strength for "young" and "aged"Clays. 81

5.15 Possible compression and swelling paths for a clay undergoingaging and slaking. 82

5.16 Possible loading/unloading paths (a) before and (b) during drained tests. 845.17. The Hvorslev surface in normalized P':Q' space. 876.1. Schematic of the triaxial test cell and measuring apparatus. 936.2. Schematic of the test sample assembly. 956.3. Photograph of dried cores of Pierre shale showing different

orientations of fissure pattern. 996.4. Irregular volumetric data from tests on two poorly saturated samples 1047.1. Drying curve showing the rate of decrease in water content with

increased time of open-air drying of Pierre shale samples. 1117.2. Drying and wetting curves as a function of drying time for the

Pierre shale. 1167.3. Representative stress-strain plots of samples having undergone

no drying and half-an-hour prior to wetting and shear testing. 1217.4. Representative stress-strain plots of samples having undergone

similar wetting drying cycles, but tested at different confining pressures. 1227.5. Comparative stress-strain plots showing drastic softening of Pierre

shale after slaking. 1247.6. Representative stress-strain plot showing three distinct segments in

the prefailure curve. 1257.7. Two- and three-segment loading curves for other stiff, or cemented clays. 1277.8. Schematic explaining two-segment loading curves, resulting from the

combining of frictional and bonding resistance forces. 1287.9. Plot of Youngs modulus and specific volume for loading curve

segment I and II. 1297.10. Prefailure P'-v paths for all Pierre shale samples tested at 30 psi

Confining pressure. 1297.11. Peak and ultimate values of P',Q', and v for unaltered samples of

Pierre shale. 1337.12. Peak and ultimate values for P',Q', and v for Pierre shale samples

dried for 4 days and wet for 1 day. 1347.13. Final and ultimate values for P',Q', and v for all Pierre sample,

unaltered and softened. 135

7.14. Comparison of loading curves for unaltered Pierre shale samplescored perpendicular and at a highly oblique angle to the major

plane of fissuration.137

7.15. Comparison of loading curves for softened Pierre shale samplescored perpendicular and at a highly oblique angle to the majorplane of fissuration. 138

8.1. Peak and ultimate values for P',Q', and v for all samples ofPierre shale, unaltered and softened (reproduced from Fig. 7.13. 144

8.2. Relationship between the specific volume and ln P' at peakstrength for Pierre shale samples at confining pressures of 10, 30,and 50 psi. 145

8.3. Values of v and ln P' at the ultimate (final) strength of all PierreShale samples tested at 10, 30, and 50 psi. 147

8.4. Values of v and ln P' at the ultimate (final) strength ofunaltered Pierre shale samples tested at 10, 30, and 50 psi. 147

8.5. Normalized peak values of P' and Q' for all samples of Pierreshale, unaltered and softened, showing well definedfailure surface. 149

8.6. Normalized peak values of P' and Q' on log-log scale for allsamples of Pierre shale, unaltered and softened, showingextension of failure surface. 151

8.7. Normalized pre-failure loading paths for all samples of Pierreshale, unaltered and softened. 154

8.8. Schematic comparing (a) the Patton model for a sawtoothjoint, to (b) normalized failure envelopes for Pierre shale. 157

8.9. Normalized strength envelope showing low stressstress strengths, overconsolidated strengths, and criticalstate strengths for heavily-overconsolidated clays inCanada. 158

8.10. Normal consolidation and critical state lines in P':v space fornatural and remolded clays of Fig. 8.9. 159

8.11. Failure envelope for natural and remolded samples, with datanormalized according to respective normal consolidation lines. 160

8.12. Complex P':v history followed by Pierre shale sample 30/28-3

prior to testing. 162A.1. Classification scheme of Underwood (1967). 192A.2. Classification scheme of Gamble (1971), based on the relationship

between slaking durability and plastic index. 196A.3. Classification scheme of Deo (1972), based entirely on resistance

to slaking. 197A.4. Two part classification scheme of Morgen-Stern and

Eigenbrod (1974). 199A.5. Modified classification scheme of the present author showing clay

shale as a unique entity of argillaceous materials. 200B.1. Relative abundance of major groups of clay minerals in Phanerozoic

Mudrocks. 206B.2. Probable stress history of the Bearpaw sediments. 207B.3. Energies of repulsion, attraction, and net curve of interaction

forparallel flat plates. 209B.4. Schematic of the stress history of an over-consolidated clay over

geological time. 211B.5. Schematic showing void ratio response to reloading after rebound. 212B.6. Aging affects observed in the laboratory for a normally consolidated

Clay. 213B.7. Schematic illustrating aging effects resulting from sustained loading

Over geological time. 214C.1. Schematic illustrating various shapes and roughnesses that are

possible in joints. 218C.2. Schematic of joint models used by Patton. 219C.3. Comparison of normalized shear strength of a joint as predicted by

equations of Ladanyi and Archambault, and the model of Patton 221C.4. Roughness profiles and corresponding ranges of JRC values

associated with each. 222C.5. Plots showing dependency of shear strength on scale and JRC. 223C.6. Critical state model of Roberds and Einstein for the behavior of

rock joints. 224C.7. Shear strength of a rough joint as a function of the joint-fill

Thickness. 226D.1. Two-dimensional friction model for granular materials. 229

TABLES

2.1. Landslide susceptible clay shales in United States 11

2.2. Shear strength characteristics of the Crete Nere in the Sinnin Valley,Italy 25

2.3. Mineralogy and plasticity of various clay shale materials 28

2.4 Various strength parameters for unaltered and softened clay shaleunits worldwide 30

2.5. Mohr-Coulomb strength parameters for unaltered and softened clayshale units 31

2.6. Summary of clay shale data presented in Tables 2.3, 2.4, and 2.5 33

3.1. Relative susceptibility of various clay types, based on ratios defined byBjerrum (1967) 43

6.1. Calculated mineralogy mode of Pierre shale using microscopic andx-ray data 98

6.2. Procedure check-list for triaxial tests on Pierre shale samples 106

7.1. Values for void ratio, saturation, and shrinkage of Pierre shale,measured at the end of the drying period 112

7.2. Values for axial and radial strains, void ratio, water content, anddegree of saturation resulting from rewetting of Pierre shalesamples 116

7.3. Testing program on Pierre shale 119

7.3. Values of Youngs modulus and Poissons ratio for loadingsegments II and III for Pierre samples undergoing softening byslaking 126

7.4. The change of specific volume, v, with stress, P', for loadingsegments II and III in Pierre shale 131

8.1. Normalized peak strengths for all samples of Pierre shale,unaltered and softened 150

8.2. Cohesion and internal friction angle values unaltered Pierre shale,as well as for samples which have been rewetted after various periodsof drying 150

A.1. Geological classification of mudrocks by Ingram (1953) 153

A.2. Geological classification of mudrocks by Folk (1968) 191

A.3. Classification scheme of Skempton and Hutchinson (1969) 191

B.1. Estimated maximum preconsolidation loads on North Americanclay shale units 207

1

CHAPTER I

INTRODUCTION

Clay shale deposits throughout the world have become notorious as a result of thenumerous foundation and slope stability problems with which they are often associated.Past engineering failures have demonstrated that experimental and theoretical methodstypically used in geotechnical engineering practice are not adequate for determining thestability of these unique materials. Present engineering design in clay shales relies primarilyon experience obtained from past failures, and results in the liberal use of highlyconservative and costly factors of safety.

The major difficulties in assessing and predicting the engineering behavior of clay shalescan be attributed to two unique properties of these materials: (a) clay shales areintermediate in behavior between rock and soil, and (b) clay shales tend to transgress fromrock-like to soil-like materials within a relatively short time period. Changes in thestrength of clay shales can be very drastic, commonly exhibiting 40% to 80% reductions inshear strength over periods ranging from 2 to 70 years. Internal friction angles of 20o to30o in unaltered clay shales are often reduced to extremely low values of 2o to 6o aftersoftening.

The factors that control the magnitude and time frame of these changes have not been wellunderstood nor have they been seriously investigated. Evidence suggests that softeningalong fissures may play a very important role in the rapid loss of strength in clay shaledeposits. However, the geotechnical and geological literature is surprisingly devoid ofsystematic studies concerned with theoretical or experimental aspects of fissuredeterioration in clay shales.

In addition, it is widely recognized that clay shales which have been totally or partiallydried, exhibit a strong tendency to slake (i.e. disperse) during rewetting. This susceptibilityof clay shales to slaking is certainly a major factor, if not the major factor, involved in thesoftening of clay shale deposits in the field. Several techniques have been developed toassess the slake resistance of geological materials under unconfined conditions. However,slaking in the field typically occurs in the presence of confining stresses. Yet, surprisingly

2

few reported studies have investigated the slaking resistance of geological materials underconditions of confinement. Furthermore, the author has found no published systematicinvestigations into the effects of slaking on the strength and stress-strain behavior of anygeological materials.

The research presented in this dissertation was undertaken in order to provide initialtheoretical and experimental assessments of the effects of fissure deterioration on thestrength and stress-strain behavior of clay shales. In particular, the drastic changes that canoccur in response to wetting and drying cycles (i.e. slaking) are examined in detail. Aprimary contribution of this research is the incorporation of the slaking process into thecritical state soil mechanics model.

Organization of Dissertation

Primarily due to the transitional nature of clay shales, a somewhat multidisciplinaryknowledge base is necessary before one can fully understand clay shale behavior. Theauthor therefore conducted an extensive literary review, concentrating on field andlaboratory observations of clay shale behavior in various parts of the world, in addition tosuch subject matter as the mechanics of joints and jointed rock, the theory andmeasurement of slaking potential, and the critical state soil mechanics model. The authoraugmented these literary studies with personal fields studies of clay shales in southern Italyand in South Dakota, USA.

One of the intentions of the author in writing this dissertation has been to present aninsightful review of many of the problems associated with clay shale deposits. However, inorder to further preserve the coherency and continuity of this dissertation, reviews anddiscussions on tangential subject matter have been placed within separate appendices, andwill be referred to in the appropriate sections of the main body. These include reviews anddiscussions regarding the classification of clay shales (Appendix A), the formation of clayshales and the nature of clay shale bonds (Appendix B), the mechanics of clean and filledjoints (Appendix C), and the mechanics of a clay mass consisting of intact coressurrounded by a mud matrix (Appendix D). For the enthusiastic reader, these appendicesprovide concise reviews into the appropriate subject, and hopefully some insight into thecomplexities of clay shale behavior.

3

The primary background material for this dissertation is presented in Chapter II and III.Chapter II includes a review of the major difficulties encountered by engineers dealingwith clay shales, and discusses the inadequacies of present geotechnical techniques toaccurately assess the long-term stability of clay shale deposits. As discussed previously,most of these difficulties have resulted from our inability to adequately account forsoftening in clay shales. In Chapter III, the author reviews and discusses the variousmechanisms proposed to account for the dramatic softening in clay shales.

The main purpose of this dissertation is to investigate the potential changes in strength andstress- strain behavior that might occur as a result of fissure deterioration, particularly inresponse to wetting and drying cycles. The author considers the theoretical aspects of thesoftening problem from two particular viewpoints. First, in Chapter IV, a model for theprogressive deterioration of a fissured clay shale is presented. This model considers aninitially unaltered clay shale consisting of fissures along which softening agents areintroduced. Softening of the clay shale mass is initiated along the fissure walls andadvances into the mass by progressively increasing the depth of softening perpendicular tothe fissure wall. From a mechanical viewpoint, the clay shale can therefore be consideredas passing through four stages: (1) initially considered as a stiff clay with a network ofclean fissures, (2) then as a stiff clay consisting of fissures filled with soft clay, (3) next asa matrix of soft mud surrounding intact cores of stiff clay, and (4) finally as a thoroughlysoftened clay mass. The author considers, in theory, the effects that this progressivesoftening may have on the strength and stress-strain response of the clay shale mass.

Second, in Chapter V, the author introduces the possibility of incorporating the effects ofslaking into the 'critical state soil mechanics' model. The role of slaking is seen in thiscontext as simply another path for altering the water content of the clay shale, while thestrength envelope remains constant or changes predictably. The importance of thispossibility cannot be overstressed. If the process of slaking can be constrained by thecritical state model, then the engineering behavior of a clay shale which has undergone, orwill undergo, a complex history of wetting and drying, is dependent only on the finalspecific volume and stress state, and not on the wetting-drying history. The critical statemodel could therefore greatly simplify the monitoring and prediction of changes in thestrength of clay shales.

Finally, the author has performed laboratory experiments involving triaxial compression

4

tests on Pierre shale samples that have undergone various cycles of wetting and dryingwhile under confinement. The purpose of these experiments were two-fold: (1) to provideinitial assessment of the effects of slaking on the strength and stress-strain response of aclay shale, and (2) to test the feasibility of incorporating slaking into the critical state soilmechanics model. The methodology and results of these experiments are presented inChapters VI and VII, while the critical state model for slaking is evaluated and discussedin Chapter VIII. Finally, the conclusions in Chapter IX present a concise summary of themajor points of the entire dissertation.

5

CHAPTER II

ENGINEERING PROBLEMS ASSOCIATEDWITH CLAY SHALE DEPOSITS

Clay shales are extensively exposed throughout the world and are invariably responsiblefor numerous slope stability and foundation problems in these areas. Where clay shalesoutcrop at the surface, the terrain is generally characterized by low-lying and gently rollinghills. Where clay shales underlie more resistant materials, numerous slope failures can befound in valleys and road cuts.

Clay shales have consistencies that range from stiff clay to shale. As discussed in detail inAppendix A, the term "clay shale" is being increasingly used to define a stiff, fissured clay,or shale, which is highly susceptible to significant deterioration as a result of interactionwith water. Besides resulting in very low strength, this transitional nature of clay shalescreates some special problems with regard to the analysis of slopes and foundations.

Geotechnical engineers are generally accustom to viewing geological materials as either arock, with engineering behavior primarily controlled by fissures and joints, or as a soilswhose behavior is highly susceptible to the fabric and water content of the intact material.However, clay shales are intermediate between rock and soil, and typically exhibitproperties of both. Furthermore, the behavior of clay shales tends to transgress from rock-like to more soil-like within time frames of a few months to 70 years or more.

These peculiar properties of clay shales create difficulties in analyzing these deposits bymeans of standard laboratory and mathematical means. In this chapter, the author reviewsthe engineering problems associated with typical clay shale deposits in four areas of theworld: North America, England, Italy, and Panama.

Evidence presented in the following sections suggests the importance of fissuredeterioration as a viable model for explaining much of the long-term reduction of strengthexhibited by clay shales. As indicated below and in the following chapters, there is a great

6

need for systematic research into the effects of fissure deterioration, and particularly theeffects of softening due to slaking.

Characteristic Problems Related to Clay Shale Deposits Around the World

British clay shales. The clay shales of England include the Lias, Oxford, and Londonclays, and range in age from Lower Jurassic (Lias) to Eocene (London clay). Althoughthese clays vary significantly in mineralogy, depositional history, grain size distribution,color, and even consistency, they are all characterized by overconsolidation, the presenceof fissures, and the tendency to exhibit significant reductions in strength over time periodsfrom a few months to 70 years or more.

In his paper on the Lias clay, Cassell (1948) discussed slides that occurred 27 to 70 yearsafter the slopes were cut at angles of 20 to 26 degrees. Shear tests indicated that thestrength within the slip planes was reduced to between 1/5 and 1/26 of the compressivestrength of the nearby undisturbed material. Cassel attributed these long term failures toprogressive deterioration of the Lias clay within the zone of fluctuating ground watertables, and stated that factors of safety derived from circular arc theory and laboratorytests on undisturbed materials were not adequate for assessing the stability of these clays.

Of all the clay shales of England, the stiff fissured London clay has become the mostnotorious. Present deposits consist of a 5 to 15 meter (16 to 49 ft) mantle of brown,oxidized London clay, underlain by less altered, yet fissured, blue-grey London clay. Someof the reported failure surfaces have been forced into the blue clay by the presence ofretaining walls (Skempton, 1977). However, most of the slips occur predominantly withinthe brown London clay, while some of the slip surfaces appear even to be controlled bythe contact of the brown and blue clays (Gregory, 1844; Skempton, 1942; Henkel, 1957).

The London clay is fairly uniform with regard to its geotechnical properties. The LiquidLimit ranges from 70 to 90, and the Plastic Limit, while generally about 28, variesbetween 24 and 32 (Henkel, 1957). The natural water content is typically slightly abovethe Plastic Limit. The clay fraction of the brown London clay is about 55%, with themineralogy of the clay fraction being 47% illite, 35% montmorillonite, 15% kaolinite, and3% chlorite (Burnett and Fookes, 1974). The relative abundance of montmorillonite

7

Figure 2.1. Tentative relation between average shear strength along slip surfacesand time, for cuttings & retaining walls in London clay (Skempton, 1948).

Figure 2.2. Relationship between cohesion intercept and time betweenconstruction and failure (Henkel, 1957).

8

in the London clay is in contrast to the clay mineralogy of the other British clay shales, inwhich expandable clays are reported to be rare (Attewell and Taylor, 1973).

As in the Lias clay, slope failures have occurred in the London clay tens of years after theslopes were constructed. In slides involving cuts and retaining walls constructed in theearly 1900's, Skempton (1948) calculated a reduction of the original shear strength from2600 psf to only 700 psf after periods of 7 to 30 years. Tentative relations between theshear strength and the time after excavation are presented in Figs. 2.1 and 2.2. These plotsindicate that the London clay exhibits a 50% loss of strength after 10 to 15 years, and aloss of about 70% to 80% of its strength after some 70 years.

The loss of strength in the London clay appears to be related to a loss of cohesion. Asreviewed in Skempton, 1977), standard triaxial and shear box tests indicate strengthparameters of

c' = 14 kN/m2 (2 psi) φ' = 20o,

while triaxial tests on large diameter specimens (250mm) give values of

c' = 7 kN/m2 (1 psi) φ' = 20o.

As illustrated in Fig. 2.3 from Chandler and Skempton (1974), the results from backanalysis of first-time slides indicate that the most realistic values for the strength in thefield are given by

c' = 1 kN/m2 φ' = 20o,

with a lower limit of

c' = 0 kN/m2 φ' = 20o

(Fig. 2.3). It is important to note that the effective angle of friction calculated for first-timeslides is significantly above the residual friction angle of 13

o, and does not exhibit a

reduction with time.

9

The mechanism by which the cohesion of the London clay is reduced toward zero is notfully understood. Several authors have noted the presence of very moist, softened zonesalong fissures both within and outside of the failure zones, suggesting that the influx ofwater into the clay may result in softening with time (Gregory, 1844; Delabache, 1844;Terzaghi, 1936; Cassel, 1948; Skempton, 1942; Henkel, 1957). However, based onpiezometric measurements within "old" slopes, Skempton (1977) has proposed that thelong-term reduction of strength in the London clay may be related to the very slowequilibration of negative pore water pressures. These two models will be discussed inmore detail in the following chapter.

Clay Shales of North America. Troublesome clay shales are widespread within the U.S.and Canada, and have been responsible for costly problems in many major constructionprojects and along many highways. Many clays and shales in the U.S. have not beenrecognized in the literature as "clay shales", although they characteristically exhibit atendency to lose shear strength due to interaction with water. A list of some of the

Figure 2.3. Failure envelope for London clay based on back-analysis of first timeslides (Skempton, 1977, after Chandler and Skempton, 1974).

10

landslide-prone clay shales in the U.S. was compiled by Fleming et al (1970) and ispresented in Table 2.1. Most of these deposits are Cretaceous or younger (Pliocene,Eocene, and Miocene), although a few are much older Paleozoic deposits.

The most notorious clay shales in North America are located in the upper Missouri andSouth Saskatchewan River basins. This notoriety results partially from problemsencountered during four major construction projects at Fort Peck, Gardiner, Oahe, andGarrison Dams, and partially from the enormous areal extent of these materials (see Fig.2.4). In the U.S., the Pierre shale and its stratigraphically equivalent deposits aloneunderlie an area of approximately 600,000 square miles and outcrop in an area of 230,000square miles (Tourtelot, 1962). These deposits extent into large areas of Canada as well.

The most troublesome clay shales in this region are the Pierre, Bearpaw, and Claggettshales. Portions of the Fort Union group and Judith River formation have also causedengineering troubles to a lesser degree. These clay shales can in general be characterizedby overconsolidation, by the presence of slickensides and fissures, by high swelling andhigh plasticity resulting from the presence of montmorillonite, by extremely high slakingpotential, and by very low cohesion and friction angles as exhibited under long-termconditions in the field.

Although natural landslides occur throughout the clay shales of north-central U.S. andcentral Canada, these slides were not considered a major engineering problem until theconstruction of large dams and major highways began in 1933. In 1938, before thecompletion of the Fort Peck Dam, a major landslide in the Bearpaw shale foundationresulted in re-calculation of strength parameters, as well as a year delay in the completionof the project (Fleming et al, 1970). Subsequently, zones of weathered clay shale wereidentified at the Fort Peck Dam site, extending to depths of 30 to 50 feet. Although thewater content of the unaltered Bearpaw shale varies between 11 to 18%, the watercontent in the weathered zones were as high as 40%. The landslide at Fort Peck Damapparently resulted from failure to account for overstressing of this degraded clay shale(Middlebrook, 1942). Slopes at the dam were redesigned using shear strength parametersof c = 2.8 psi and φ = 10.5

o.

11

Table 2.1. Landslide susceptible clay shales in United States (Fleming et al,1970).

12

Similarly, the Bearpaw shale at Gardiner Dam has been arbitrarily divided into three zonesbased on consistency: soft, medium, and hard. The upper soft zone has been significantlydisturbed and softened by swelling and weathering, with fissures and slickensidesfrequently showing signs of past desiccation. The intact material between these fissurescan be readily remolded with the fingers (Peterson et al, 1960). In contrast, the lower hardzone is less disturbed, more uniform, harder, more dull in appearance, and shows fewerslickensides. The medium zone is transitional between these two zones. The water contentvaried from only 20 to 27% in the hard zone to 29 to 36% in the soft zone. Similarly, theunconfined compressive strength varied from 400 psi in the hard zone, to a remarkablelow of only 7 psi in the soft zone.

Although laboratory tests at Gardiner Dam gave very consistent strength values for thehard zone, extreme variation of water content and consistency, and the presence ofslickensides, made it virtually impossible to obtain useful strength parameters bylaboratory testing of the soft shale. Final design criteria were based primarily on analysis ofslopes that had failed in other areas or during construction at Gardiner Dam. Failure zones

Figure 2.4. Location map for major dams built in clay shales in the NorthCentral United States and Canada (Fleming et al, 1970).

13

at Gardiner Dam were invariably within the soft shale zone, and often near the bottom ofthe zone. The design strength parameters which were established based on these criteriaare presented in Fig. 2.5, and indicate values of c' = 0 and φ' = 6.5

o for the softened

material.

Similar difficulties were encountered during the construction of Oahe Dam in SouthDakota. Direct shear tests on firm samples of the underlying Pierre shale gave values of

c = 8.3 to 12.5 psi and φ′ = 20 to 22o

for residual strength. Direct shear tests on the under lying weathered Pierre shale indicatedmuch lower shear strength values of

c = 3.6 psi and φ′= 11.9o,

which were adopted as the design criteria. However, slope failures in the softened shaleduring construction required recalculation of the strength parameters to even lower valuesof

c = 2.1 psi and φ′= 8.5o.

Later slides indicated that these values also slightly overestimated the shear strength of theslope abutments. The decision to completely remove the weathered shale from one of theslide-prone abutments resulted in an extra 6.5 million cu yd of evacuated material (Fleminget al, 1970). In addition to slope failures, the Pierre shale at Oahe Dam was responsible forexcessive rebound of the outlet-works stilling basin and resulted in redesign of the basinand anchoring of the underlying floor.

The creation of these dams and their reservoirs has resulted in relocation of many roadsand bridges. In addition, modern highway design criteria for grade and alignment requirednumerous road cuts and fills. Since these changes, landsliding has become a major problemalong highways in the north central U.S. and Canada.

14

Figure 2.5. Proposed clay shale design strength parameters for Bearpaw shale atGardiner Dam (Fleming et al, 1970).

15

Bruce and Bump (1967) reported on major difficulties encountered during theconstruction of a 12-mile extension to Highway 44 in South Dakota. Within this shortdistance, major failures of the Pierre shale occurred in four natural slopes and under threefill deposits while still in the construction stage. Excavations of the failed slopes uncoveredsprings and concentrations of discontinuities. Failure zones under the 40 and 65 ft highfills extended 10 to 40 ft below the natural ground surface. At completion of the project in1966, instability of the Pierre shale was responsible for a cost increase of $409,800 overthe original contract, or an additional $25,234 per mile of finished highway. Although theslopes were flattened and all active material was removed, Scully (1973) reported thatmovement began again in 1969. In 1984, the present author similarly noted many veryrecent slope failures along this highway section. Construction and maintenance of suchhighways in areas of underlying clay shales has been very costly.

Within the clay shales of North America, failures have occurred in both natural and man-made slopes with inclinations as low as 3

o. Most major failures in these materials have

been related to the presence of very low-strength zones of softened and weathered clayshale. Laboratory testing has proved virtually useless for establishing design criteria inconstruction projects, and invariably overestimates the in-situ strength of the foundationsand slopes. Slope design in the clay shales of north central U.S. and central Canada has, bynecessity, relied primarily on the analysis of local failures in natural and man-made slopes.This has resulted in costly and dangerously unreliable engineering design.

Clay Shales Along the Panama Canal. The excavation of the Panama Canal was greatlyhindered by landsliding in the clay shale phases of the Cucaracha formation. The mosttroublesome slides were undoubtedly the East and West Culebra slides which alone addedmore than 50 million cu yd of material that needed to be excavated. These slides, whichbegan in 1907, converged in massive failures in August and September 1915, accompaniedby upheaval of the canal bottom and blockage of the canal. Since 1916, movement ofthese slides has continued in the form of slow, sporadic flows into the canal.

Binger (1948) describes the clay shale of the Cucaracha formation as

...the clay shales all appear to be disturbed, and they contain manydegrees of slickensides and fractures or joints .... Badly crushed,gougelike zones of varying thicknesses have been encountered in virtually

16

every exploratory drill hole reaching these shales. A zone of such materialhaving a thickness of more than five feet ... was found in a testpitexcavation. The material in this zone was so soft that it flowed into theexcavation even when the wall bracing’s were carried within a foot of thebottom of the shaft.

The water content of the solid Cucaracha clay shale is about 17 to 18% (Binger, 1948),and is generally 7 to 15 below the Plastic Limit (Banks, 1971). In slickensided zones at theprobable depth of sliding, within badly crushed or gouged zones, and within slide debris,the water content can range as high as 30 to 35%. Piezometric measurements indicate areduced pore water pressure (i.e. less than the canal level) in the East and West Culebraslopes, probably reflecting rebound and swelling occurring as a result of canal excavation(Banks, 1971).

Analysis of slopes at the Panama Canal indicates drastic reductions in strength haveoccurred within the Cucaracha clay shales. Within five months in 1912, the effective shearstrength dropped by 20%. From 1912 to 1915, this had been reduced 22% on the eastbank and 35% on the west bank. By March 1947, strength along the east and west bankswas about 20% of the original strength as measured in 1912 (Binger, 1948).

In recent studies, Banks (1978) observed three modes of failure in the Cucaracha clayshale slopes:

(1) first time slides under short-term conditions,(2) first time slides under long-term conditions,(3) slides along pre-existing slip surfaces.

He also performed a wide variety of shear strength tests on four groups of samples:(1) those apparently void of slickensides,(2) samples with obvious slickensides,(3) samples repeatedly sheared until residual strength was reached, and finally,(4) remolded samples which were consolidated from a slurry (‘fully softened” samples).

17

The results from these tests are plotted in Fig. 2.6, and indicate a wide range of values for"intact" specimens. In addition, slurry consolidated samples exhibit a strength that isslightly higher than residual strength.

After analyzing failed slopes in the Cucaracha clay shale, Banks concluded that:

(1) the strength mobilized in first time slides which occur immediately afterexcavation is best represented by the peak strength of unaltered, slickensidedsamples,

(2) the strength mobilized in first time slides which occur after long periods havepassed is equivalent to the "fully softened strength" obtained from slurryconsolidated samples, and

Figure 2.6. Summary of the shear strength data for the Cucaracha clay shale(Banks, 1978).

18

(3) the strength activated along pre-existing slip surfaces is of course equal to theresidual strength.

This significant observation implies that some natural "remolding" mechanism is acting onthe deposits of Cucaracha clay shale and reducing the strength toward that of a normally-consolidated clay. Other observations discussed above, further suggest that the softeningof the Cucaracha clay shale is accompanied by an increase in the water content.

Italian Clay Shales. The Italian Peninsula has had a long history of landslide problems.The vast majority of landslides in Italy occur within clay shales and rock/clay shalemelanges. As seen in Fig. 2.7, the presence of landslides in any area is strongly influencedby the presence of these deposits (Esu, 1977). In southern and central Italy, as well as inSicily, these deposits generally belong to either the Liguride Complex, an Eocene flyschconsisting of alternating layers of black clay shales with various marine clastics, or theSicilide Complex, a Cretaceous to Oligocene deposit consisting of red and green clays orclay shales, tuffite, and an arenaceous flysch unit (D'Argenio et al, 1975).

Although the clay shales of the Sicilide Complex cover only 10% of the entire Apenninearea in southern Italy, Belviso et al (1977) have estimated that this unit alone accounts forup to 90% of the landslide events in this area. The terrain underlain by thick clay shaledeposits in Italy is characterized by numerous old landslides, and gentle rolling hills thatcontrast sharply with nearby steeper peaks composed of other geological materials. Theslopes formed in the clay shales generally have inclinations no greater than 10 to 20

o

(Evangelista et al, 1977).

Villages in the countryside of southern Italy have historically been built on slopes or at thesummit of hills. Because deposits of clay shale underlie many of these villages, theirfoundations have become unstable resulting in cracking, shifting, and even sliding ofbuildings and streets. A consortium of geologist and geotechnical engineers wasestablished in the 1970's to study the foundation problems that threaten the village ofBisaccia. These problems appear to be directly related to the instability of the variegatedclay shales that underlie the foundation material of the village. Similarly, enormous,reactivated slides and flows which threaten the villages of Calitri and Senerchia are at leastpartially affected by underlying layer of variegated clay shales (Cotecchia, 1982; Maugeriet al, 1982).

19

The Italian clay shales can, in general, be described as an overconsolidated clay of mediumto firm consistency. The clay mineral fraction is predominantly kaolinite and smectite(montmorillonite) with lesser amounts of illite. These clay shales generally exhibit two sizeorders of fabric complexity, which are probably indicative of tectonic shearing. Samplescan be easily separated into relatively coherent rhombohedral lenses, on a scale of 1 to 10centimeters, bounded by curved or planar shear discontinuities that are polished and often

Figure 2.7. Map of the landslide areas and their relationship to underlying rocktype in Italy (Esu, 1977).

20

striated. With significantly more effort, these lenses can be further broken down to exposea more scaly fabric inside.

In the Sicilide Units, this complex fabric is accompanied by highly distorted reddish-brownand olive green color banding of thickness ranging from centimeters to several meters.Therefore, the clay shales are often referred to as "argille scaliose (scaly clays)", "argillevaricolori (varicolored or variegated clays), or "complesso caotico (chaotic melange)",depending on the preferences of local geologists and geotechnical engineers and on thedominant characteristics of the clay shale.

The geotechnical behavior of the Italian clay shales is characterized by high plasticity,medium to high slaking potential, high swelling potential, and a tendency to soften underenvironmental conditions found in the field. This softening, as well as the presence of thecomplex fabric of the clay shales, greatly complicates slope stability analysis in areaswhere these materials are involved. The geotechnical properties of the Italian clay shaleshave been shown to vary significantly with composition (Belviso et al, 1977), anddirection of shear (Picarelli, 1981; Fenelli et al, 1981).

However, much of the scatter of shear strength values may also result from varyingdegrees of alteration within the samples. In the field, the clay shales of Italy have beendivided into three zones based on the amount of weathering that has occurred (Fenelli etal, 1982). They are indicated as Zone I for deeper, unaltered material, Zone II for partiallyaltered clay shale, and Zone III for overlying totally altered mudstone. In Zone I, the clayshale is characterized by the complex rhombohedral structure described earlier. Incontrast, the altered material in Zone III consists of fragments smaller than one centimeter,forming a mass with higher porosity in which the original structures have been obliterated(A.G.I., 1977).

As presented in Fig. 2.8, the water content significantly increases as weatheringprogresses, while the unconfined shear strength is drastically reduced. The unaltered clayshale in zone I exhibits an unconfined compressive strength as high as 12 MN/m2

(83 kips), while the altered material and the material sampled from landslides both showvery low strengths of 0 to only .5 MN/m2 (3.4 psi). The plot also illustrates the ratherdrastic decrease in strength which can occur with very little change in water content. In

21

fact, an increase in the water content results in a 90 to 95% reduction of the unconfinedstrength.

Several researchers have alluded to the role of softening in controlling the in-situ behaviorof the Italian clay shales. D'Elia (1980) performed direct hear tests on three groups ofspecimens taken from a slide area in Italy: (1) fresh material which had not undergone anysignificant alteration, (2) "partially softened" clays which had been stored in a humidityroom for about four months and (3) "softened" material in which remained in the humidityroom for eight months. As presented in the Mohr-Coulomb plots of Fig. 2.9, the strengthparameters measured were reduced from c' = 35 KPa (5.1 psi) and φ' = 28

o for the

unsoftened material to c' = 12 KPa (1.7 psi) and φ ' = 18o for softened material. Back-

analysis of slope failures show that strength mobilized in deeper slides is similar to thestrength parameters measured for "partially softened" material and is much greater thanthe residual strength. Furthermore, some translational slides of weathered clays exhibitedstrengths as low as the strength of the "fully softened" material.

Similarly, Manfredini et al (1981) investigated the influence of softening within the blackclay shales (crete nere) of the Liguride Complex, and reported that the slaking potential ofthese montmorillonite-poor clay shales was greatly increased by the presence of a scaleyfabric. Slake durability tests indicated that for non- scaley specimens, 60 to 90% of theoriginal material remained after two cycles, whereas only 15-25% remained in specimenswhere a scaley fabric was present. In the field, weathering tends to obliterate originalstructure of the material, resulting in a homogeneous, soft, and relatively plastic claymatrix with remnants of scales and plates. The results of triaxial and direct shear tests arepresented in Table 2.2, and indicate that the strength parameters of the in-situ and earthflow materials, as measured in the laboratory, are much higher than those for completelyremolded samples. However, back- analysis of the strength mobilized along the slipsurface of the slides indicates strength parameters which are close to the peak strength ofremolded samples. Again these values are well above the residual strength of the clayshales.

Cancelli (1981) investigated the effects of softening in the Lugagnano clay shale which isresponsible for numerous slope stability problems in northern Italy. While the Lugagnanoclay shale appears to be void of the complex scaley fabric of the southern Italian clay

22

Figure 2.8. Relationship between unconfined compression strength and water contentfor the different degrees of weathering in the Laga formation, Italy (A.G.I., 1977).

23

Figure 2.9. Variation of shear strength in an Italian clay shale as a function of time of storage in ahumidity room (D'Elia, 1980).

24

and shales, it is highly fissured and undergoes significant softening in natural slopes.Results from triaxial and direct shear tests on undisturbed core samples are presented inFig. 2.10, indicate that the softening of in-situ Lugagnano clay shale results in a reductionof the effective cohesion from 55 kPa (8.0 psi) to zero, while the effective friction angleremains the same, or is only slightly reduced. The residual strength is represented by aneffective cohesion values of zero and a significantly lower effective friction angle of 11

o to

13o. Values obtained from back-analysis of several periodic slides indicate that the

mobilized strength is very close to the residual strength values. However, assuming c' = 0for back-analysis of first-time slides, indicates an effective friction angle of 24

o. This value

is slightly lower than that measured for the softened material, but is much higher than theresidual friction angle.

Residual Versus Softened Strength

In order for the strength of a clay shale to be equal to its residual value, it is generallyassumed that the material must experience a shear deformation in excess of that requiredto mobilize the peak strength. It has long been recognized that the resistance mobilized byreactivated landslides is equal to the residual strength of the material within the slip zone.

In addition, some first-time slope failures have occurred under states of stress whichwould have mobilized only residual strengths (Krahn et al, 1979; Palladino and Peck,1972). Such slope failures can generally be contributed to slip along surfaces which havebeen pre- sheared by tectonic deformations, or by deformations resulting from unloading.These pre-sheared surfaces can result from passive failure during the erosion of overlyingsediment (Nichols, 1980) or the cutting of river valleys (Matheson and Thompson, 1973),from ice thrusting during glaciation (Krahn et al, 1979), or from differential unloadingduring progressive deglaciation of valleys (Palladino and Peck, 1972).

In his presentation at the Seminar on the Geotechnics of Clay Shales in Denver on March15, 1984, Ralph Peck presented numerous examples of slope failures in various clay shalesof the United States in which slip apparently mobilized residual strength along sub-horizontal, pre-sheared bedding planes. He further advocated that the presence of pre-sheared horizontal surfaces are so prevalent in some clay shale deposits, that their

25

existence should be assumed in projects where a potential slide would be intolerable.Unfortunately, there is a growing tendency for researchers and practicing engineers toconsider the assumption of residual strength as a "catch-all" cure for the complicationsencountered in clay shale deposits.

However, there is a vast amount of evidence indicating that most first-time slides withinclay shales occur at strengths significantly greater than residual strength. Furthermore,there is no evidence suggesting that the strength of all slopes in clay shales will be reducedto residual values within the engineering lifetime of these projects. This is certainly thecase for the London clay, in which changes in the stability of man-made slopes have beenmeasured for up to 70 years after construction.

As reported by Skempton (1977):

Figure 2.10. Shear strength envelopes for undisturbed core samples of firm (φp')and weathered (φs) Lugagnano clay shale. Residual strength is indicated by φr

(Cancelli, 1981).

26

It appears that the displacements preceding a first-time slide are sufficientto cause some progressive failure, reducing the strength toward the fullysoftened or the lower limit of fissure strength; but the displacements arenot so large as to reduce the strength to the residual value.

Evidence presented earlier in this chapter indicates that first-time slides within the clayshales of the upper Missouri and South Saskatchewan River Basins, as well as theCuacarcha clay shale in Panama and the clay shales of Italy, mobilize softened strengthswhich are significantly greater than the residual values.

In cases where old landslides are being reactivated, or where pre-sheared surfaces areknown to exist, residual strength must be assumed along these surfaces. If the location andorientation of pre-sheared surfaces cannot be accurately determined, then it is wise toassume that the residual strength will be mobilized along the entire slip surface. However,in areas where pre- sheared surfaces have not been recognized, and where a landslide,though troublesome, is tolerable, assuming residual strength will probably result inoverconservative design and can drastically increase the construction expense ofexcavation or of building slope and foundation supports.

Comparative Summary and Discussion

The above review briefly illustrates some of the extreme challenges faced by the engineerdealing with clay shales. Our present understanding of the engineering behavior of thesematerials is primitive relative to the current state-of-the-art for other geological materials.Engineering designs based on traditional methods for analyzing geological materials havebeen dangerously inadequate. In fact, most successful designs in clay shales have beenbased on past failures, rather than on adequate test results.

The author feels that most of our past inadequacies in dealing with these materials haveresulted from our lack of appreciation of their transitional nature. This has resulted indangerous and costly slope and foundation designs. The transitional nature of clay shalesis both temporal and physical. Physically, clay shales are transitional between rock andsoil, and therefore exhibit properties of both. This has been a source of problem forgeotechnical engineers, who traditionally view geological materials in terms of rockmechanics or soil mechanics, both rarely in terms of both. In addition, clay shales are

27

transitional in time, and tend to transgress from rock-like behavior to soil-like behaviorwithin a relatively short time period. Such rapid changes in material properties createchallenges in classification and engineering design, both of which are traditionally based onmaterial properties as they exist at the present and not on possible future properties.

Tables 2.3 to 2.6 list reported values for several geotechnical properties of these clayshales. The mineralogy and plasticity values presented in Table 2.3, illustrate that thecharacteristically high to very high plasticity of clay shales. Tables 2.4 and 2.5 list variousstrength parameters which have been measured for unaltered and softened clay shales, aswell as those strength values which have been calculated based on back calculation offailed slopes. Table 2.6 summarizes all of this information in a more concise format.

The strength data presented in these tables, as well as the above review, indicate that somefirst-time slides occurring immediately after excavation in clay shale have mobilized thepeak strength of jointed or slickensided samples. However, the peak strength of intactmaterial is never mobilized by clay shale deposits in the field, indicating the importance ofconsidering the jointed, and therefore rock-like, nature of clay shales. In cases whenfailure has occurred along pre- sheared surfaces, the mobilized resistance is characterizedby the residual strength. However, the growing tendency for practicing engineers toalways assume that clay shales will mobilize the residual strength within the life-time of anengineering project is not supported by the data presented, and can lead to costlyoverconservatism. Indeed, much of the evidence presented here indicates that the strengthtypically mobilized in the field during first-time, long-term slides, is characterized byvalues which are well below peak strength and well above residual strength.

The difference between the strength of unaltered clay shale and that of the fully softenedmaterial can be quite large. Thus, the amount of strength reduction that takes place overrelatively short periods of time can be quite drastic. For example, the Cucaracha shale inPanama lost 80% of its original strength within 35 years. Furthermore, this can result inextreme scatter in strength values as measured within a single clay shale deposit at any

28

Tab

le 2

.3. M

iner

alog

y an

d pl

astic

ity o

f var

ious

cla

y sh

ale

mat

eria

ls

29

Tab

le 2

.3 c

ont.

30

Tab

le 2

.4 V

ario

us s

tren

gth

para

met

ers

for

unal

tere

d an

d so

ften

ed c

lay

shal

e un

its w

orld

wid

e

31

Tab

le 2

.5.

Moh

r-C

oulo

mb

stre

ngth

par

amet

ers

for

unal

tred

and

sof

tene

d cl

ay s

hale

uni

ts w

orld

wid

e.

32

Tab

le 2

.5. C

ont.

33

Tab

le .2

.6 S

umm

ary

of c

lay

shal

e da

ta p

rese

nted

in T

able

s 2.

3, 2

.4, a

nd 2

.5.

34

time. As an example of the latter case, consider that the unconfined compressive strengthof the Pierre shale has been reported as ranging from zero to over 17,000 kPa!

The drastic, progressive softening of clay shales with time, and the extreme variability oftest results create very difficult challenges for the geotechnical engineer dealing with thesematerials. How does one account for such large changes in strength and compressibilitywith time? What material and environmental factors control the amount and rate of changethat might occur over the lifetime of an engineering project, and how does one monitorand predict these changes? Furthermore, even without considering the time factor, is itpossible for the engineer to account for the extreme variability of strength that can occurwithin a given deposit, without relying on a prohibitive amount of site investigation?

All of these questions must be answered before the softening of clay shales can beadequately accounted for. In this dissertation, the author investigates these questions boththeoretically and experimentally. The following chapter reviews the present state ofknowledge regarding the mechanisms by which clay shale softening might be occurring,and in particular, discusses the lack of systematic studies on the effects of slaking in thefield.

3535

CHAPTER III

SOFTENING MECHANISMS INCLAY SHALE DEPOSITS

The review and discussions of the previous chapter emphasized the importance ofmaterial softening in controlling the strength and compressibility of clay shale deposits.The softening of clay shale was shown to result in up to 80% loss of strength in somedeposits after 30 to 70 years, and was probably responsible for the extreme variation ofmaterial strength observed at several engineering sites. In order to be able to accountfor such changes and variations in the engineering behavior of clay shales, it isimportant to understand the mechanisms by which the material properties of clay shalesare altered. While many researchers have speculated on the causes of softening in clayshales, few have carried out extensive investigations concerning these possiblesoftening mechanisms. Therefore, there is very little understanding of the factors whichcontrol the amount and rate of softening in clay shale deposits. Certainly, the lack ofsuch studies is one of the major reasons that clay shales have remained one of the mostdifficult engineering materials.

Two primary mechanisms have been proposed as playing important roles in thesoftening of clay shales. These include the equilibration of negative pore pressures andthe deterioration of fissures by means of chemical alteration or slaking. As discussedbelow, these two mechanisms are not entirely independent of one another, and it isprobable that both act to some degree in all clay shale deposits. In addition, thesemechanisms surely act in conjunction with the progressive failure mechanism in such away that they are enhanced by, as well enhance, progressive failure. It may be difficultto uncouple the effects of one mechanism from the other.

Still, it is essential that we recognize which material and environmental factors controlthe extent and rate of softening at any given site. In addition, the engineer needs toknow the effect that a given amount of deterioration will have on the strength andstress-strain behavior of the clay shale deposit. This chapter will briefly review thepresent state of knowledge regarding the proposed softening mechanisms.

3636

Softening Mechanisms

The importance of fissures. There are three important effects of fissuration incontrolling the behavior of clay shale deposits. First, they provide inherent planes ofweakness along which shear can occur. It is for this reason that clay shales must also beconsidered from a rock mechanics point of view. Second, the fissures greatly increasethe permeability of clay shale deposits. Without the conduits provided by fissures, mostclay shales would be virtually impervious. Third, fissures significantly increase thesurface area exposed to weathering agents.

Thus, in addition to significantly weakening a clay shale mass, fissures greatly enhancethe process of deterioration in clay shales, by first allowing a greater influx of water andother weathering agents into the interior of the mass, and by then exposing moresurface area of the rock mass on which these agents can act. No matter which softeningmechanism is acting on a clay shale deposit, the rate and extent of softening is highlydependent on the presence of fissures.

In addition, whether softening occurs in response to the equilibration of negative porepressures, or by chemical alteration or slaking, softening surely weakens fissure wallmaterial first before progressing further into the clay shale mass. For this reason, theauthor proposes a simple fissure deterioration model in Chapter IV and discusses theimplications of this model with regard to the possible changes in engineering behaviorthat can occur in response to progressive softening of clay shales.

Equilibration of negative pore water pressures. As unloading occurs in response toexcavation or natural erosion, negative pressures can be created within the pore fluid ofa clayey deposit. If the permeability of the clayey material is low, these negative porepressures may persist for extended periods of time, and may act to initially strengthenthe deposit. However, as these negative pore pressures are equilibrated by the influx offluid, the apparent strength resulting from the tensile pressures is destroyed.

Base on piezometric measurements within the blue and brown zones of the Londonclay, Skempton (1977) proposed that long-term reduction in the strength of theLondon clay might be related to extremely slow dissipation of negative pore pressures

3737

which had been generated in response to the cutting of the slope. Previousmeasurements in the London clay had indicated that the value of Bishop's average porepressure ratio, ru, as defined in Fig. 3.1, was in the range between 0.25 and 0.35 for

conditions of equilibrium. This value had been reached in slope of brown London claywhich had been cut 125 years previously. However, in the facing slope which had beenrecut 19 years before the piezometric measurements were taken, the pore pressureswere only one-half of the equilibrium value. Skempton thus concluded that theequilibration of negative pore pressures in the London clay required about 40 to 50years, and could thus account for the loss of strength in slopes of London clay.

As been determined previously by several researchers, Skempton confirmed that thestrength mobilized at failure was equal to the "fully-softened strength. The very slowequilibration of negative pore pressure was therefore not proposed as a mechanism by

Figure 3.1. Definition of Bishop's average pore pressure ratio, ru (Skempton,

1977).

3838

which peak strength was exceeded or by-passed. The proposed mechanism does,however, suggest that the delays in slope failures of the London clay might becontrolled by the time required to dissipate negative pore pressures, rather than thetime necessary for deterioration to occur. It is important to note that the slidesevaluated by Skempton in his 1977 report, deliberately excluded shallow slips and slipsin zones of seasonal variation. Unlike the slides evaluated, these slips typically occurafter exceptionally heavy rainfall, especially following prolonged dry periods. For thesecases, as well as for the slides investigated by Skempton, the mechanism of softeningmust still be assessed.

That negative pore pressures still exist in the London clay 20 to 40 years afterexcavations is both surprising and significant. It had been assumed that the presence offissures would allow rapid dissipation of negative pore pressures. However, it is notnecessary to implicate extremely low permeability to account for the presence ofnegative pore pressures many years after excavation.

It is important to assess whether the negative pore pressures result entirely from theinitial slope cutting, or whether they might in addition result from soil suctionassociated with material deterioration and swelling. Furthermore, it should bedetermined whether failure of the London clay occurs primarily along fissures. If this isthe case, one must be concerned about measuring the equilibrium of pore pressurealong fissures, and not the equilibrium within the clay mass itself. Finally, equilibrationof negative pore pressure is certainly associated with the influx of more water into thematerial, and it is important to consider how the generation and dissipation of thepressures might occur in conjunction with the deterioration and swelling discussed inthe next section.

Fissure deterioration model. As early as 1844, an engineer, Gregory, and a geologist,Sir Henry Delabache, observed the softening of fissures within the London clay andsuggested that progressive softening might account for slips which occurred severalyears after construction. Soon after, Collins (1886) suggested that clay slopes alongcanals would eventually be reduced to inclinations similar to those exhibited in thenatural terrain. These observations were not expounded on further until 1936, andaccording to Skempton (1948):

3939

one of the reasons for the long delay in the development of soilmechanics can be found in the apparently 'treacherous' andunpredictable behavior of stiff-fissured clays: especially since these arewide-spread in south east England and in France.

Terzaghi (1936) reintroduced the idea that the strength of the stiff, fissured clays, suchas the London clay, depended on the spacing of the fissures and on the degree ofsoftening adjacent to the fissures. He also emphasized the behavioral differencebetween the relatively stable non-fissured clays and the troublesome, fissured clays. Theprocess, as described by Terzaghi and Peck (1948), is as follows:

Almost every stiff clay is weakened by a network of hair cracks orslickensides. If the surfaces of weakness subdivide the clay into smallfragments 1 in. or less in size, a slope may become unstable duringconstruction or shortly thereafter. ... If the spacing of the joints in theclay is greater than several inches, slopes may remain stable for manyyears or even decades after the cut is made. The lapse of time betweenthe excavation of the cut and the failure of the slope indicates agradual loss of the strength of the soil. Before excavation, the clay isvery rigid, and the fissures are completely closed. The reduction ofstress during excavation causes an expansion of the clay, and some ofthe fissures open. Water then enters and softens the clay adjoiningthese fissures. Unequal swelling produces new fissures until the largerchunks disintegrate, and the mass is transformed into a soft matrixcontaining hard cores. ... The water seems to cause only deteriorationof the clay structure; seepage pressures appear to be of noconsequence.

Since Terzaghi's statements, numerous researchers have observed softening andincreased water content of the London clay along failure zones and within unfailedfissures (Skempton, 1942; Cassel, 1948; Henkel, 1957). It is of course important infuture studies to distinguish between softening which occurs prior to failure and thatoccurring as a result of failure. Further evidence for the deterioration of fissures priorto slope failure, comes from Hutchinson (1970) who noted that mudflows in theLondon clay often consisted of a mud matrix surrounding hard clay fragments.

4040

As discussed in the previous chapter, many slides have been occurred within the upperpartially weathered layers of the clay shales of North America and Italy. In addition,numerous authors have reported localized softening occurring along failure planes andirregularly into fissures in lower zones (Widger and Frelund, 1979; Scully, 1973;Cancelli, 1981; Manfredini et al, 1981). Scully discussed in detail the very localizedexistence of "seeps" consisting of highly weathered Pierre shale with water contentsgreater than 130%. Even at depths of over 100', softened zones were found such thatthe undrained compressive strength varied from the softened strength of 3 TSF to85 TSF within a 2' distance. Certainly, the occurrence of these softened zones must bejoint controlled.

Deterioration of the strength along fissures can result from either (1) chemicalalteration, or from (2) increasing the amount of water adsorbed by the clay material.Chemical alteration of a clay shale deposit can involve either the precipitation ordeposition of new minerals, or the chemical transformation of existing minerals.Furthermore, chemical alteration can act to either decrease strength by breaking bonds,by increasing porosity, or it can actually increase strength by creating new bonds,particularly in the presence of cementing agents. The mineralogy can be altered suchthat properties such as the swelling potential, or the residual strength, are either morefavorable, or less favorable to stability of the slope or foundation.

Although the effects of chemical alteration are probably of extreme importance incontrolling the long- term strength of clay shale deposits, this dissertation will notdiscuss this process or its effects in any detail. Unfortunately, extensive investigationsregarding chemical alteration and its potentially important effects in clay shales aresurprisingly few. Our understanding of clay shale behavior would be greatly advancedby more numerous and more extensive studies on chemical alteration in clay shales,particularly if the results are incorporated into the fissure deterioration model to bediscussed in the next chapter.

However, the primary emphasis of the present investigation is on the processes ofslaking and swelling, during which bonds are destroyed and the moisture contentincreased without altering the mineralogy of the clay shale. Evidence presented abovesuggests that slaking probably play a major role in the deterioration of clay shales.

4141

Therefore, the slaking process will be reviewed and discussed in more detail toward theend of this chapter.

Progressive failure mechanism. Progressive failure is a mechanism by which thestrength of clay shale along a potential slip zone can be progressively reduced frompeak values to residual values. If the stresses at any position along this potential slipsurface exceed the peak strength of the material, and if displacement is allowed, thenlocalized failure will occur and the strength of the material at that position will decreasetoward the residual value (Skempton, 1964). In the example of a slope, illustrated inthe schematics of Fig. 3.2, the stresses at the toe of the potential slip surface weresufficient to displace the material beyond the peak strength in the zone D-F. This willcause a redistribution of stresses in the local area such that the peak strength mightlikewise be exceeded in the adjacent material, as in zone F-G. This process willcontinue until the average shear strength along the potential slip plane is no longersufficient to resist complete failure along the plane, or until the redistributed stress stateis such that the peak strength is no longer exceeded at any point within the slip plane.

Several factors determine the susceptibility of a material to progressive failure. Bjerrum(1967) discussed several of these factors as they relate to a uniform slope. In order forprogressive failure to be initiated, stresses must locally exceed the peak strength of thematerial. Thus, the danger of initiating progressive failure increases with an increase ofthe ratio, ph/sp, where ph is the horizontal effective stress and sp is the peak effective

shear strength. Furthermore, the local differential strains in the advancing failuresurface must be sufficient to strain the clay beyond peak resistance. The ratio, eh/ep, isa measure of the amount that the horizontal strain, eh, resulting from the removal oflateral load, will exceed the peak failure strain, ep, and an increase in this ratio alsoincreases the susceptibility of the slope to progressive failure. The final ratio, sp/sr, of

peak strength to residual strength, indicates that progressive failure is also favored insoils which undergo rapid and drastic reductions in strength after the peak strength ismobilized.

Table 3.1 illustrates the relative danger of progressive failure for various materials. Asdiscussed in Appendix B, clay shales can be described as overconsolidated plastic clayswith strong bonds, and thus correspond to the third and fourth columns in the

4242

Figure 3.2. Contours of shear strain in a slope, and its relation tocorresponding stress states in an overconsolidated clay (Atkinson and

Bransby, 1978).

4343

Tab

le 3

.1 R

elat

ive

susc

eptib

ility

of v

ario

us c

lay

type

s, b

ased

on

seve

ral r

atio

s de

fined

by

Bje

rrum

.

4444

table. As measured by the factors discussed above, unweathered clay shales exhibit lowsusceptibility to progressive failure. However, the weathering of clay shales breaksdown bonds and creates conditions of very high horizontal stresses relative tothe vertical stresses. Duncan and Dunlop (1968) have shown that such high horizontalstresses can cause a ten-fold increase of the maximum shear strength at the toe of aslope where progressive failure would most likely be initiated. Therefore, clay shalesthat have been weathered, or are in the process of weathering, are very susceptible toprogressive failure, and it appears that the mechanisms of progressive failure and clayshale deterioration can act in combination to greatly increase the risk of slope failure.

Interactions between softening mechanisms. In the natural environment, thesoftening mechanisms discussed above probably all act to some degree in all clay shaledeposits. In addition, the softening of clay shales may involve complex interactionsbetween these softening processes, and it may be difficult, if not impossible, touncouple the effects of one mechanism from those of another.

As an example, we will look at two fictitious cases, whereby the processes discussedabove are activated or accelerated by different "triggers". In the first case, the stressstate on a slope is altered by either excavation, or by the natural process of erosion.This action results in two responses: fissures are propagated or opened, and negativepore pressures are generated. The generation of negative pore pressures actually acts toinitially stabilize the slope, so that the eventual loss of this pore water tension can aloneresult in slope failure some time after excavation or erosion. However, other processescan in turn act to greatly complicate this simple explanation. For instance, the openingof fissures and the presence of negative pore pressures will result in increased waterinflux if a ready source is available.As illustrated in the schematic of Fig. 3.3, this influx of water can then result in one, ormore, of the following actions:(1) the negative pore pressure is reduced, thereby decreasing the effective shear

strength,(2) bonds are broken, resulting in swelling and slaking, an increase in the negative pore

pressure, an increase in water content, and a reduction in the effective shearstrength,

(3) the influx of chemical agents is increased, alteration or deposition of mineralsoccurs, and the effective shear strength is either increased or decreased.

4545

Figu

re 3

.3 S

chem

atic

com

plex

act

ion-

reac

tion

path

s po

ssib

le in

cla

y sh

ale

mat

eria

ls.

4646

The decrease in effective shear strength by any of these methods can thereby result inviscous or "immediate" local strains, which in turn cause either (1) total failure, (2) areduction of strength toward residual values, or (3) dilation. If dilation occurs, thennegative pore pressures are again increased resulting in a "positive feedback"mechanism by which the stability of the slope depends more and more on the presenceof negative pore pressures, and less on the progressively decreasing material strength.As the above cycle continues, failure can occur if the effective stresses exceed themobilized strength, regardless of whether this occurs as a result of the dissipation ofnegative pore pressure, or the reduction of material strength.

Several important observations can thus be made regarding negative pore pressures inclay shale deposits. First, negative pore pressures, as measured by piezometers in thefield, can result from material slaking and swelling, or from dilation during shear, aswell as from those stresses generated by an excavation or erosion. Second, theexcessive length of time required for "dissipation" of negative pore pressures, asreported for instance by Skempton (1977), may indicate the superposition of negativepore pressures generated by swelling and dilation onto those resulting directly from theexcavation or erosion. Third, the discussion above suggests that it is not necessary toassume very low permeability in order to account for the long times required for thedissipation of negative pore pressures.

In the second general case to be discussed, the triggering action results from wettingand drying cycles occurring in the absence of exterior stress changes. A commoncharacteristic of true clay shales is their tendency to swell and slake in response tocycles of wetting and drying. As discussed above, this results in an increase in watercontent, a decrease in material strength, an increase in negative pore pressure, andpossibly an increase in permeability as fissure walls are softened. In this case, negativepore pressures can be attributed to capillary tension in the partially saturated material,as well as the soil suction acting in the previous example. In addition to decreasing thematerial strength, swelling tends to increase horizontal stresses, creating dangerousconditions that favor progressive failure. Reduction of strength in this general case canessentially follow the same cycle illustrated in Fig. 3.3, although the cycle is initiated bya different process.

4747

The two typical cases discussed above illustrate that the progressive deterioration ofclay shales in the field involves a rather complex interplay of the effects of dissipationof negative pore pressures, chemical alteration, slaking and swelling, and the dilationand fabric reorientation associated with shear strain. Dissipation of negative porepressures involves rather well- refined principles of fluid flow through porous media. Incontrast, the changes in strength and stress-strain response associated with mineralalteration, swelling, or slaking may be more difficult to assess or predict.

In addition, the history of slaking and chemical alteration in clay shales can be verycomplex. These processes have been investigated very little, or in some cases, not atall. Any significant advances in our ability to analyze the behavior of clay shalematerials in the field, will probably result directly from investigations into the effects ofmineral alteration, swelling, and slaking.

Slaking in Clay Shales

As discussed in Appendix A, clay shales are characteristically highly susceptible toslaking. It is highly probable that the process of slaking plays an important role in thesoftening of clay shales. Numerous researchers have shown that the strength of claysand clay shales is closely related to the water content, and that fissure walls in failedclay shales often exhibit higher water contents than the adjacent clay material. In theabsence of mineralogical changes, water content can be increased by (a) dilation duringshear, (b) simple swelling related to elastic rebound following unloading, and (c)swelling and slaking related to the breaking of interparticle bonds in response towetting, or wetting and drying cycles in the absence of external load changes.

The slaking process. The term "slaking" usually implies the phenomena of materialdisruption or dispersion observed when dried or undisturbed chunks of clay or shale areimmersed in water. Moriwaki (1974) noted four modes of slaking, including

(a) swelling, described as an increase in bulk volume without visible cracking orsignificant loss of material,

(b) body slaking, which appears to originate from internal processes and which rapidlytraverses large portions of mass with no apparent deterioration between cracks,

4848

(c) surface slaking, characterized by loss of mass due to "sloughing" of tiny flakes ofgrains from the entire surface with no apparent cracks in the underlying material,and

(d) dispersion, characterized by loss of mass resulting from the separation of clay-sizedgrains which go into spontaneous suspension, rather than settling.

As discussed by Bjerrum (1967), most slaking is assumed to result from the disruptionof diagenetic bonds and the release of stored strain energy. However, it has also beenshown that slaking can result from the compression of trapped air within the clay orshale mass, particularly in soils containing highly-expansive clay minerals. As water ispulled into the clay mass by capillary forces, or suction pressures generated byexpansive clay minerals, air pockets can become trapped and compressed. Wetting ofthe mass stops when the pressure on the air pockets equals the suction or capillarypressure of the water. However, the rock or clay mass may fail in tension before thisequilibrium is reached, resulting in body slaking, or possibly surface slaking, asdescribed above. The importance of this mechanism in some clays and shales can bedemonstrated by the absence of slaking when slake testing is performed under avacuum.

Some clays and shales at natural water content slake when immersed in water. Others,when immersed, will remain stable with regard to slaking even though stresses arereleased. However, if these materials are first dried, and then rewetted, slaking mayoccur. In this case the process is referred to as "drying-induced" slaking. Nakano(1970) presented evidence which supports the hypothesis that some materials will notslake as long as the water content remains above a certain threshold, but if the watercontent is lowered below this threshold, slaking will occur during either drying orrewetting.

Moriwaki (1974) recognized three mechanisms of swelling and slaking, including (a)swelling by osmotic forces, (b) swelling by hydration of ions and clay surfaces, and (c)compressed-air slaking, and concluded from his investigations on artificial shales thatthe dominant slaking mechanism is controlled by the mineralogy:(a) sodium clays - osmotic swelling,(b) calcium montmorillonite - osmotic swelling and hydration,(c) calcium illite - hydration and compressed-air,

4949

(d) calcium kaolinite - compressed-air only.

Moriwaki further concluded that the susceptibility of any material to slaking willdepend not only on the mineralogy, but also on the "physico-chemical characteristics",such as bonding, and the chemistry of the slaking fluid.

McClure (1980) performed an extensive investigations on the slaking behavior of manynatural clays and shales, including many of the clay shales of North America. Hedetermined that, of all the natural materials he tested, (a) 40% behaved as expectedbased on physico- chemical characteristics and Moriwaki's observations on artificialshales, (b) 16% showed minor variation from expected behavior, (c) 20% deviatedsignificantly due to cementation, and (d) 20% more deviated significantly due tounknown causes. McClure therefore concluded that the forces which bind shales andclays together (i.e. diagenetic bonds) need to be further identified and understood.

Testing for susceptibility to slaking. The slake tests employed in the presentclassification schemes of argillaceous materials include (1) the modified jar slake test,(2) the slake durability test, (3) the "one- dimensional free swell test", and (4) thesulfate soundness test. In the modified jar slaking test (Moriwaki, 1974), anundisturbed or dried sample is placed on a wire mesh, which is then lowered into a jarof water. The slaking potential is indicated by the loss of weight as the materialcrumples and falls through the mesh. The slake durability test of Franklin and Chandra(1972) is performed by rotating six 40 to 60 gram samples in a wire mesh drum, whichis partly immersed in water. The weight percentage of material remaining in the drumafter one cycle (200 revolutions in ten minutes) defines the "slake durability index". Thesulfate soundness test is similar to the above tests, except that the previously driedsamples are wetted in a sodium or magnesium sulfate solution. The percentage ofmaterial retained on a 3/8" sieve indicates the sulfate soundness index.

The slaking factor of Morgenstern and Eigenbrod (1974) is unique in that it is based onthe one-dimensional free swell of a laterally-confined sample. In these tests, the changeof height, and therefore the change of water content, were measured as a function ofwetting and drying cycles. Increased swelling is assumed to indicate progressive slakingwithin the specimen.

5050

The modified jar slaking, slake durability, and sulfate soundness tests described above,all provide very useful information on the relative "weatherability" of variousargillaceous materials. However, the extent to which these tests significantly modelconditions existing in the field is limited. All of these tests measure the amount ofslaking under conditions of zero confining pressure. The previous review of slakingprocesses discusses several different mechanisms of slaking, each caused by variousmagnitudes of disruptive stresses. It is conceivable that slaking forces exhibited bysome materials may be balanced by very low confining pressures. In such materials,slaking would only affect the very top surface of deposits in the field. In addition, theslake durability test employs a tumbling factor, which generally does not exist in thefield.

More important, however, is the fact that none of these tests measure the relativereduction of shear strength that occurs as these materials undergo deterioration in thefield. Although these tests do measure the important tendency of a material to degrade,they do not necessarily measure the relative reduction of strength due to degradation.

Summary of slaking in clay shales. The present methods of measuring slakingsusceptibility have been reviewed, and their deficiencies discussed. Evidence presentedin Chapters II and III indicates that slaking and swelling may be important factors in thereduction of the strength of clay shales in the field. It is clear that the methodscommonly employed for investigating the process of slaking do not provide adequateinformation regarding the effects of slaking on the strength and stress-strain behavior ofclay shales in the field.

First, the phenomenon of slaking is generally studied only under zero confiningpressures. The potential influence of both hydrostatic and deviatoric stresses oninhibiting or enhancing the slaking process is not fully appreciated. Second, the slakingin clay shales is primarily confined to the walls of fissures and joints, and it is not wellunderstood whether the process will enhance or inhibit further slaking by increasing ordecreasing the effective permeability of the clay shale mass.

Furthermore, the relationship between the processes of slaking and swelling is not oftenfully appreciated. Swelling occurs as additional water is absorbed either within theinterlayers of the clay mineral structure or within the viscous water layer

5151

surrounding individual clay particles or aggregates. Swelling generally occurs in one oftwo modes, both involving the disruption of interlayer or interparticle bonds. The firstmode involves a decrease in external stress, which results in the breakage of bonds, andthe subsequent pulling in of water in response to repulsion between particles. This isthe mode of swelling associated with simple rebound during the formation ofoverconsolidated clay, and is represented by the path A-B in Fig. 3.4.

The second mode of swelling can occur under constant external load, and results fromthe disruption of bonds due to slaking processes. With unconfined clay chunks, like

Figure 3.4. Modes of swelling associated with simple rebound and slaking inan overconsolidated clay or clay shale.

5252

those used in slake durability tests, the disruption of bonds due to the immersion, orwetting and drying cycles, can act to break apart the specimen. This is the phenomenonthat is typically referred to as "slaking". However, in a confined clay mass, "breaking-up" of the material is inhibited and the disruption of bonds is instead accompanied byswelling and softening of the clay mass. Thus, as illustrated by the path B-C in Fig. 3.4,slaking under constant external load can result in swelling of the mass, accompanied byan increase in the overall water content.

The literature is essentially void of reported systematic investigations into the changesin strength and stress-strain that result from slaking. Since the slaking history of a clayshale deposit can be very complex, determining a relationship between the materialstrength and the degree of slaking may be very difficult. However, there are uniquerelationships between the strength of clay materials and their water contents. Inparticular, development of the critical state theory of soil mechanics has greatlyenhanced our understanding of the complex interaction between the stress state and thewater content of a saturated clay mass. In Chapter V, the author introduces the ideathat the effects of slaking might be able to be incorporated into the critical stateconcept. If successful, the critical state model would provide very important constraintson the slaking phenomena, and would greatly simplify the monitoring and predicting ofthe effects of slaking.

53

CHAPTER IV

THEORETICAL ASPECTS OF FISSURE SOFTENINGIN CLAY SHALES

In the natural environment, clay shales are typically jointed. The highest strengthmobilized by clay shales in the field is never greater than the strength of jointed orslickensided samples. In addition, much evidence has been presented which indicatesthat the softening of clay shale deposits is initiated along fissures, thereby furtherconcentrating the zones of weakness along fissure planes. Therefore, it is generallyimportant to consider the jointed nature of clay shales, and not threat these materialsstrictly as classical soils.

Since the first suggestions by Gregory (1844) and Delabache (1844) that the strengthof the London clay was being reduced by a softening of the clay material adjacent tofissures, several other researchers have alluded to this possible mechanism to accountfor the deterioration of strength in many clay shales. However, the author has found noin-depth discussions as to the mode by which deterioration progresses in clay shales,nor have detailed theoretical or experimental investigations been carried out withregard to the progressive changes in strength and stress-strain behavior that occur inresponse to fissure deterioration.

In this chapter, the author proposes a simple model for fissure deterioration, andinvestigates the implications of this model. With the aid of reviews presented inAppendices C and D, possible changes in the strength and stress-strain behavior of aprogressively softening fissured mass will be discussed. Although subsequent chapterswill concentrate primarily on the slaking mechanism in clay shales, the conceptspresented in this chapter should be relevent to any material undergoing fissuredeterioration by any softening mechanism.

Model for Fissure Softening

The model for progressive deterioration of clay shales, as envisioned by the author, isillustrated by the schematic in Fig. 4.1. In the schematic are four cross sections of atheoretical clay shale mass, each representing a different stage of deterioration. Stage 1

54

represents a fissured clay shale in which no alteration has occurred. At stage 2,alteration has been initiated along the fissures, and progresses in stages 3 and 4, untilthe mass is entirely deteriorated. This model is further illustrated by the photograph inFig. 4.2, which shows joint controlled deterioration in a portion of the Pierre shale thatis in an intermediate stage of weathering.

At stage 1, the strength and stress-strain response is probably controlled primarily bythe frequency, strength, and orientation of the fissures, and thus behaves similar to ajointed rock. At stage 4, however, the material is in theory behaving much like ahomogeneous, naturally-remolded clay. During the intermediate stage 2, the clay shalemass can be considered as a rock-like material with the strength controlled by the

Figure 4.1. Schematic of the proposed model for progressive deterioration ofclay shales.

55

orientation and shear resistance of soft, filled joints. Similarly, at stage 3, the massmight be best considered as a soft matrix surrounding large "grains" of stiff, intactmaterial, much like a "boulder clay". The major difficulty in modeling the behavior ofclay shales is that the failure of a clay shale slope or foundation can occur at any stageof the deterioration as depicted, and requires only that the shear strength is reduced toa level such that it can no longer resist the shear stresses acting on the potential slipsurface.

From this discussion of the fissure deterioration model, it is apparent that the analysisof the strength and stress-strain response of a progressively deteriorating clay shaleshould consider four aspects of soil and rock mechanics: (a) the effects of remolding onthe behavior of a homogeneous, but not necessarily isotropic, clay mass, (b) thebehavior of a material consisting of stiff cores of intact clay within a soft mud matrix,(c) the strength along joints with a filling of soft mud, and (d) the behavior of a fissuredmass.

Figure 4.2. Photo showing joint deterioration in the Pierre shale of SouthDakota (photo by author).

56

Implications of the Fissure Softening Model

The shear strength of jointed masses. The shear strength of a jointed mass isdependent not only on the strength along the individual joints, but also on such factorsas the orientation and spacing of the joint, or joint network, relative to the orientationof the stress field. Methods to analyze the engineering behavior of jointed masses arenumerous and often quite complex, and it is not within the framework of thisdissertation to discuss these models in any detail.

In their very simplest form, all of these models consider joints as planes of weaknessalong which shear failure is preferred. However, as discussed above, the ability tomobilize this low resistance depends on the spatial relationship of the joint system tothe stress field, as well as on the magnitude of stresses. Failure in a jointed mass canoccur entirely along joints, entirely through intact material, or both along joints andthrough intact material. The strength mobilized by a jointed mass generally lies betweenthat of the intact material and that along a single joint.

The shear strength of a single clean, filled, or softened joint. A detailed review onthe mechanics of single joints is presented in Appendix C. The present section willdiscuss how the knowledge presented in that review might be incorporated into thefissure softening model.

The shear behavior of a single clean joint is controlled affected by several factors: (a)the shear and tensile strengths acting between or within the joint walls, (b) theroughness of the joint surface, and (c) the orientation of the joint relative to the stressstate. A joint may refer to either a simple, well defined crack, or a complex shear zoneconsisting of interwoven cracks or an infilling of softened material or granulated"gouge".

As illustrated in Fig. 4.3, joint surfaces may be undulated or stepped, as well as planar,and on a smaller scale, these surfaces may be rough, smooth, or very polished. Theeffect of non-planar, or rough, joint surfaces is to create a certain degree ofinterlocking between the two joint surfaces, as they try to slip past each other.Therefore, in order for slip to occur, these rough peaks or "asperites" must either slideover one another, resulting in dilation normal to the joint, or the asperities must

57

themselves break by shear or tensile failure. The interlocking of asperities cancontribute to the measured cohesion observed in fissured clays.

A filling of softer material within a fissure or joint tends to, of course, separate thewalls of the fissure, and thereby prevents, or minimizes, the interlocking of smallerasperities. For most filled joints, the failure envelope is located between that for thefillings and that for a similar clean joint. As the thickness of the filling increases relativeto the amplitude of the asperites, the strength and stiffness of filled joints decrease.

The thickness of filling required to significantly reduce the strength of a joint dependson the roughness of the joint walls. For joints that are smooth and planar, the strengthof the joint can be rapidly reduced to the strength of the fill material. In addition, if thestrength of the fill material is much lower than the shear strength along the clean joint,the reduction of strength resulting from the presence of a filling can be rather drastic.Finally, a filling of swelling clay is particularly troublesome due to the loss of strengthand high pressures associated with swelling.

The softening of wall material along a fissure is essentially equivalent to the progressivethickening of a fill within the fissure. However, softening along fissures also results inthe weakening or destruction of asperities along the fissure walls. The detrimentaleffects of joint softening are therefore three-fold. First, softening creates a low-frictioncoating, or weak fill, and therefore reduces the strength and stiffness in accordancewith the above discussion. Second, the process of softening tends to greatly reduce theroughness, or JRC, of the joint walls; this reduces the amount of interlocking betweenjoint walls, and also decreases the thickness of filling required to reduce the strength ofthe joint to that of the soft fill material. Finally, softening of the joint walls, of course,reduces the shear resistance, or JCS, of the wall material, and facilitates the breaking ofany remaining asperities.

The transgression of softening in a rough, undulating joint, and a rough, planar joint,are illustrated by the schematic in Fig. 4.4. Softening acts initially to remove the higherorder, or small-scale, irregularities along the joint walls. However, if the degree ofsoftening is excessive, or the initial roughness coefficient is small, then the large-scaleirregularities may likewise be removed. In such cases, the strength would be

58

Figure 4.3. Schematic illustrating various forms for joints.

59

Figu

re 4

.4.

Sche

mat

ic il

lust

ratin

g th

e pr

ogre

ssio

n of

sof

teni

ng in

und

ulat

ing

and

plan

ar jo

ints

.

60

reduced to that of the fill material. In addition, the progressive reduction in the strengthof the joint wall material, may allow increased breakage through the irregularitiesbefore they have been fully softened, as in case 1 (t=3) of Fig. 4.4. For joints that areinitially planar and relatively smooth, as in case 2, only a very small thickness ofsoftened fill is required to totally reduce the joint strength to the fully softened value.

Much more research is essential before the effects of progressive softening of jointwalls are fully appreciated. Several factors may influence the nature by which softeningaffects the strength of a joint:

(1) Softening typically reduces the roughness of a joint wall; however, if the materialalong the joint wall is variable in its susceptibility to weathering, then softeningmay be confined to local areas, leaving other areas to remain as intactirregularities. In such cases, the joint may actually become rougher, resulting inan increase in the degree of interlocking once shear displacement has beeninitiated. This condition might be further enhanced if the rate of softening isintense.

(2) If the penetration of softening agents, such as water, or dissolved ions, into thejoint walls is slow, then the boundary between the intact wall material and thesoftened “fill” is likely to be well defined. However, if the rate of penetration israpid, or if the agents are able to act over long periods of time, then there maynot be a distinct boundary between softened and intact material.

(3) joint walls is slow, then the boundary between the intact wall material and thesoftened "fill" is likely to be well defined. However, if the rate of penetration israpid, or if the agents are able to act over long periods of time, then there maynot be a distinct boundary between softened and intact material.

(4) For the case of joint softening in response to drying-induced slaking, the natureof the boundary between intact and softened material will be determined by theextent to which drying progresses into the joint walls, and by the number ofwetting and drying cycles. If drying is limited to a thin layer of material adjacentto the joint, and particularly if numerous cycles of wetting and drying areallowed to occur, then a well defined boundary will be formed between intactmaterial and a much softer fill material. Therefore, the thickness of the softened

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zone, as well as the nature of the gradation from intact to softened material,depends on the rates of drying and wetting relative to the cyclic periods overwhich wetting and drying occur.

(5) There is considerable uncertainty as to the effect of softening and swelling onjoint permeability. Swelling in a joint may decrease the joint permeability byfilling voids between the walls. However, swelling might also increase thepermeability of the jointed mass by increasing the width of conduits throughwhich flow occurs. In the former case, swelling produces a negative feedbackmechanism by which further softening is inhibited, whereas the latter case isaccompanied by a positive feedback mechanism which enhances furthersoftening. The relative influence of these two factors may depend on whetherswelling is prohibited by external forces, as well as on the extent to which thejoint width is increased by wall deterioration.

In conclusion, for joints that are initially planar and smooth, the amount of fissuresoftening required to significantly reduce the shear resistance along the fissure may bequite small. Slaking and swelling along fissure walls reduces the degree of wallroughness, enhances shear through asperities, and creates a fill that reduces the amountof interlocking between joint walls. All of these affects act to significantly reduce thestrength along fissures.

The strength of a clay mass consisting of a soft matrix surrounding stiff, intactcores. If extensive alteration of a clay shale is allowed to continue, the material atfailure may consist of an assemblage of stiff, intact clay cores surrounded by a soft,remolded clay or mud matrix. An example of an extreme case is presented in Fig. 4.5.Such materials can typically fail at very low slope angles in the form of mudflows

Initially, the degree of softening within the clay mass may be such that there still existsome interlocking between some stiff clay cores. In this case, some significant cohesionmight exist as a result of this interlocking. With further degradation, the stiffcores are no longer close enough to interlock, but shear through the clay mass mayrequire the rotation or translation of some of these cores. The strength of such a claymass is very near, if not equal to, the strength of the softened clay. Ultimately, with

Figure 4.5. Schematic of a highly altered clay shale consistingof stiff lumps of intact clay within a matrix of highly weathered

clay.

Figure 4.6. Schematic illustrating possible changes in the failure surfaceresulting from progressive softening in a clay shale.

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continued softening, stiff lumps of clay may continue to exist within the soft claymatrix, but their presence would not significantly affect the shear strength of the mass.

Therefore it is important to note that it is not necessary to soften the entire mass of aclay shale before it will exhibit the "fully softened", or remolded, strength. It is onlynecessary that the concentration of intact clay cores is sufficiently low as to inhibitsignificant interaction between intact cores. The interlocking component of strength,which is certainly the major compenent of strength in fissured clay shales, could intheory be eliminated if only 5% of the clay shale material is softened. In reality, thepercentage of softening required to eliminate interlocking is probably somewhat higher,but may still be surprisingly low.

The Effects of Progressive Fissure Deterioration on the EngineeringBehavior of Clay Shales

The previous Fig. 4.1 schematically illustrated four stages of alteration in clay shales.This figure has been modified in Fig. 4.6, in order to illustrate how fissure deteriorationmight alter the failure plane, and thus the engineering behavior, within a clay shale.

The unaltered clay shale, should be considered as a highly jointed mass, and not as ahomogeneous soil in the classical sense. If the strength of the fissures is much less thanthat of the intact material, and if the applied stresses are low, then the failure plane maybe controlled entirely by the location of the fissures. This condition is indicated by thesolid line in stage 1 of Fig. 4.6. Like hard, jointed rocks, the strength and stress-strainresponse is therefore controlled by the shear resistance along the individual fissures,and on the geometric relationship between these fissures and the acting stress field. Incontrast, if the resistance of the intact material is not much higher than that of thefissures, and particularly if high stresses are applied, then failure may occur along theplane of maximum shear stress, indicated by the dashed line, and would therefore passentirely through intact material. Field evidence indicates that this is rarely the case inunaltered clay shales.

Shear along the failure plane of the unaltered clay shale mass in stage 1, could beconsidered analogous to shear along a single "joint" given by the dark jagged line. This"joint" is characterized as having a high first-order roughness with saw-tooth asperities,

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while the higher-order roughness is controlled by the roughness along the true fissures.The stress-strain response at low stresses would be similar to that illustrated inFig. 4.7a, which indicates low compressibility, high dilation perpendicular to the failureplane, high peak strength, and a significant post-peak drop in strength. As in Fig. 4.8a,the typical strength envelope is given by the Fairhurst equation for rock joints. Thestrength parameters are characterized at low confining stresses by a high apparentcohesion, resulting from interlocking of the large "saw-teeth" and by a high frictionangle which results from the added "i" component discussed in Appendix C. At muchhigher stresses, failure occurs through asperities, so that the strength parameters areessentially equal to the cohesion and friction angle values for intact material.

For the partially softened clay shale represented by stage 2 in Fig. 4.6, deterioration hasbeen confined to the walls adjacent to the fissures. The "analogous joint" along whichfailure is preferred is again given by the thick, dark line, and might be characterized asan undulating filled joint. As illustrated by Fig. 4.7b, the typical stress-strain behavior atlow stresses might initially exhibit low stiffness and high compressibility, as the layer ofsoft material within the joints is compressed. However, with further shear displacement,the intact walls of the fissures again make contact, resulting in increased stiffness. Thepeak resistance would be lower than that for the unaltered clay shale due to a decreasein the first-order roughness of the failure plane, as well as a significant decrease in thestrength along the true fissures. As discussed previously, the decrease in fissurestrength, in response to softening, is a result of smoothing of the joint walls, separationof the joint walls by softened fill material, and softening of the asperities along thefissure. Therefore, the strength envelope, represented in Fig. 4.8b, indicates a lowercohesion and a lower friction angle at low confining stresses, as a result of thesmoothing of both first and second-order roughness. In addition, the apparent cohesionat higher stresses may also be reduced if softening has enhanced the breaking of thelarger asperities within the analogous joint.

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Figure 4.7. Schematic illustrating changes in the stress-strain behavior of aclay shale undergoing progressive softening as illustrated in Fig. 4.6.

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Figure 4.8. Schematic illustrating changes in the strength envelope resultingfrom progressive softening of a clay shale as illustrated in Fig. 4.6.

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Once fissure softening has progressed to the third stage the plane of failure may bealmost entirely within softened material, as shown in Fig. 4.6. Although there may besome interference from some intact cores within the failure plane, shear displacementeven at high stresses, will probably occur by rotation of these cores, rather than failingthrough them. As can be deduced by comparing the curves in Figs. 4.7c and d withthose of Figs. 4.8c and d, the engineering behavior at this stage is very similar to thatfor the fully deteriorated clay shale. In such cases, the strength and stress-strainresponse is controlled entirely by the softened material.

In summary, analysis of the strength of a clay shale undergoing progressive softeningalong fissures is complex, and generally requires some understanding of principles fromboth soil and rock mechanics. However, as will be discussed in the following chapter,the concept of critical state soil mechanics may provide an important tool with whichthe analysis of softening clay shales might be greatly simplified.

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CHAPTER V

THE ROLE OF SLAKING WITHIN THECRITICAL STATE MODEL

Much research in the past three decades has been oriented toward defining the limitingstress states within which a soil can exist, independent of stress path history. Sufficientevidence exists to support the concept that the critical state of a homogeneous, non-fissured, saturated soil can be defined by two stress invariants and the water content (orspecific volume). This concept, generally referred to as Critical State Soil Mechanics,provides a concise framework within which the seemingly complex behavior of soil canbe understood. In this chapter, the author proposes that the effects of slaking mightsimilarly be considered within the framework of the critical state model. If so, thiswould greatly simplify the otherwise complex analysis of a clay or clay shale which isundergoing softening in response to slaking.

Review of the Principles of Critical State Soil Mechanics

In stress analysis, it is convenient to use stress parameters that are independent of thechoice of reference axes. One set of these, so-called, stress invariants are the octahedralnormal effective stress, σ'oct, and the octahedral shear effective stress, τ'oct, which can

be defined in terms of effective principal stresses as

σ'oct = (σ'1+σ'2+σ'3)/3, (5.1)

τ'oct = [(σ'1-σ'2)2+(σ'2-σ'3)

2+(σ'3-σ'1)2]1/2. (5.2)

3For the special case of standard triaxial shear states, where σ'2 = σ'3,

σ'oct = (σ'1+2σ'3)/3, (5.3)

τ'oct = √2(σ'1-σ'3)/3. (5.4)

To avoid the recurring √2/3 term, new invariants, P' and Q', are defined for the triaxialcase, as

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P' = (σ'1 + 2σ'3)/3 = σ'oct, (5.5)

Q' = (σ'1 -σ'3) = 3τ'oct/ √2 . (5.6)

The stress invariants are defined such that, for an isotropic, homogeneous material, P'defines the hydrostatic component responsible for isotropic volumetric changes, and Q'incorporates all of the shear stresses which can cause distortion.

The basic principles of critical state soil mechanics have been discussed in some detailin the text by Atkinson and Bransby (1978). Much of the review below is a summary ofthat text and will therefore not reference the numerous, individual contributions whichhave led to the development of the critical state concept.

For normally-consolidated clays which have been isotropically compressed and thenloaded to failure in drained and undrained triaxial tests, data indicate that therelationship between P', Q', and the specific volume, v, at failure can be uniquelydefined by a single line, as illustrated in Fig. 5.1. This line is referred to as the "criticalstate line". The stress paths followed during drained and undrained tests arerespectively shown in Q':P' and v:P' spaces in Figs. 5.2 and 5.3. An important propertyof the critical state line is that it uniquely defines the residual strength, regardless of thestress path followed.

The projection of the critical state line in Q':P' space is defined by a line through theorigin, and with a slope designated as M. In v:P' space the projected critical state line iscurved and similar in shape to the normal consolidation line. In fact, when plotted inv:ln P' space the critical state line can be approximated by a straight line with a slopenearly equal to the normal consolidation line, as in Fig. 5.4.

Although somewhat more complicated than normally consolidated clays, the behaviorof overconsolidated clays can still be accounted for within the critical state concept.The typical stress path for an heavily overconsolidated clay in Q':P' space during adrained triaxial test is illustrated in Fig. 5.5. Since σ3' remains constant for a typically

drained triaxial test, ∆Q' = ∆σ 1' , (5.8) ∆P' = ∆σ1'/3 , (5.9)

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Figure 5.1. Ultimate failure points for drained and undrained tests on normallyconsolidated specimens of Weald clay (Atkinson and Bransby, 1978; after Parry,

1960).

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Figure 5.2. Stress paths in (a) Q':P' and (b) v:P' space for undrained tests onnormally consolidated samples (Atkinson and Bransby, 1978).

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Figure 5.3. Stress paths in (a) Q':P' and (b) v:P' space for drained triaxialtests normally consolidated samples (Atkinson and Bransby, 1978).

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Figure 5.4. The critical state line in v:ln P' space (Atkinson and Bransby,1978; after Parry, 1960).

Figure 5.5. Stress path followed in a drained triaxial compression test on anoverconsolidated clay (Atkinson and Bransby, 1978).

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and ∆Q'/ ∆P' = 3 . (5.10)

The stress path for heavily overconsolidated clays is characterized by an increase of P'and Q' until peak strength is mobilized at stresses beyond those at the critical state line,followed by a reduction of P' and Q' toward the critical state line. Thus, foroverconsolidated clays, the critical state line represents the "ultimate" or "residual"strength.

The peak strength of overconsolidated clays is dependent on the specific volume, aswell as P' and Q'. However, it has been found that the differences in specific volume canbe accounted for if P' and Q' are normalized using the value of pe, which is the stress on

the normal consolidation line at that particular specific volume. Data shown in Fig. 5.6indicates that the peak failure surface for heavily overconsolidated clays can be definedby a straight line on a plot of Q'/Pe' versus P'/Pe'. After reaching this surface, called the

Hvorslev surface, the stresses tend toward the critical state line, as shown in Fig. 5.7for undrained triaxial tests. In contrast, lightly overconsolidated clays exhibit anincrease in the stresses until they intersect the Roscoe surface, which is defined by thestress paths of normally consolidated clays. These surfaces are bettervisualized in a three-dimensional plot of Q':P':v space, as in Fig. 5.8.

Figure 5.6. Normalized failure surface for drained and undrained tests onoverconsolidated samples of Weald clay (Atkinson and Bransby, 1978; after

Parry, 1960).

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The Roscoe, Hvorslev, and tensile failure surface define the boundaries which confinethe possible stress states for a given soil. Similarly, the critical state line defines thestate of stress at which further changes of P', Q', or v cannot occur. In addition to beinglimited by these surfaces, the possible stress path is restricted by the testing procedure.For example, in Fig. 5.9, is shown the plane along which the stresses can move during adrained test.

In summary, the behavior of normally and overconsolidated clays can be illustrated inthe Q':P' and v:P' plots of Fig. 5.10. The plot shows the stress paths and failure loci forsamples that have all been consolidated to the same preconsolidation stress, Pc, and

then allowed to swell under different confining pressures. Samples which lie on or tothe right of the critical state line in v:P' space, will exhibit a decrease in specific volumeand will fail at the critical state line. In contrast, samples which lie to the left of thecritical state line will exhibit an initial decrease in specific volume, followed by a largeincrease after failure, and will tend to fail at peak strengths which are higher than thosegiven by the critical state line. After peak failure, however, the stresses will drop untilthey are compatible with the critical state line.

Figure 5.7. Schematic of expected undrained tests paths for samples atdifferent overconsolidation ratios (Atkinson and Bransby, 1978).

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of the critical state line will exhibit an initial decrease in specific volume, followed by alarge increase after failure, and will tend to fail at peak strengths which are higher thanthose given by the critical state line. After peak failure, however, the stresses will dropuntil they are compatible with the critical state line.

Atkinson and Bransby (1978) have suggested based on theoretical arguments, that thelocus of peak failure for heavily overconsolidated clays might lie along the swelling line.However, data presented by Henkel (1959) and shown in Fig. 5.11, indicates that thepeak failure surface for artificially overconsolidated Weald clay is does not line alongthe swelling line.

Figure 5.8. The complete state boundary surface in three dimensions ; Q':P':vspace (Atkinson and Bransby, 1978).

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The Role of "Aging" Within the Critical State Concept

The process of "aging", which occurs in response to sustained consolidation loads overlong periods of time, has been discussed in Appendix B which discusses the formationof clay shales. For the sake of convenience, Fig. B.7 is repeated here as Fig. 5.12.There is some question as to effects of aging with regard to the critical state concept. Inparticular, the question arises as to whether the process of aging is accompanied by achange in the intrinsic material properties and therefore a displacement of the materialyield surfaces, or whether the effect of aging is simply to alter the "current" physicalstate without affecting the validity of the critical state concept or the position of theexisting yield surfaces.

The illustration in Fig. 5.12 shows that if a normally consolidated clay is allowed to age,and is then consolidated in the laboratory, the e-log P' curve will show an abrupt

Figure 5.9. The drained plane in Q':P':v space (Atkinson and Bransby, 1978).

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change in compressibility at a pressure, Pc', which is greater than the actual load,Po'.This gives the false impression that the clay has been subjected to a past load

greater than the load that existed. The clay is essentially behaving as if it had beenpreviously consolidated to Pc' and then allowed to swell under a reduced load,

suggesting that the normal consolidation line is still uniquely defined and not dependenton the path of loading.

Figure 5.10. Failure states of drained tests on samples at differentoverconsolidation ratios (Atkinson and Bransby, 1978).

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Bjerrum (1973) has shown, as in Fig. 5.13, that the ratio of the undrained shearstrength obtained from vane tests over the true consolidation pressure, Su/Po',varieswith the plasticity, Ip, and is greater for aged clays. He has further shown that the ratioof Pc'/Po',increases with plasticity, but is constant with depth of the deposit. Tavenasand Leroueil (1977) modified the plots of Bjerrum by combining the Su/Po' and Pc/Po'curves in order to normalize the shear strength with regard to Pc', and determined thatthe ratio of Su/Pc for the aged clay is nearly identical to Su/Po' for the young clay (see

Fig. 5.14). This behavior is identical to that for overconsolidated clays considered

Figure 5.11. Predicted failure points for overconsolidated and normallyconsolidated clays (Henkel, 1959).

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within the concept of critical state soil mechanics, and suggests that the material yieldsurfaces defined by the critical state model remain undisturbed during the agingprocess. Still, more extensive research is required to verify these findings.

Slaking within the Critical State Concept

In discussions regarding the critical state concept, it is common to consider the act ofswelling as it occurs in response to either a decrease in P' while Q' remains constant, ora decrease in both P' and Q'. However, the potentially important role of slaking inincreasing the water content of a clay, independent of stress changes, has not been fullyappreciated. The present author proposes that swelling associated with the slakingprocesses should be considered within the critical state model. The strength exhibitedby a clay shale which has undergone slaking, might be dependent only on the final watercontent and not on the slaking history.

Figure 5.12. Geological history and compressibility of normally consolidatedclays, showing effects of "aging" (Fleming et al, 1970; after Bjerrum, 1972).

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Figure 5.13. Changes in the undrained shear strength ratio and the consolidationpressure ratio versus the plasticity index for "aged" and "young" normally

consolidated clays (Bjerrum, 1973).

Figure 5.14. Normalization of the undrained shear strength for "young" and "aged"clays (after Bjerrum. 1973).

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As illustrated in Fig. 5.15, an overconsolidated clay or clay shale which has undergoneconsolidation, and possibly aging, followed by swelling due to rebound after unloading,may undergo further swelling in response to slaking near the surface. Whereas the agingprocess can cause a normally consolidated clay to behave as if it were overconsolidated,the process of slaking might, likewise, cause a clay to reverse these effects, by reducingthe degree of overconsolidation. Therefore, in much the same way that a sustained loadcan act as an "aging" mechanism, slaking serves as a mechanism by which a clay mightregain some of its youth, or be "rejuvenated".

This rejuvenation process can be demonstrated by the plots in Figs. 5.16. As discussedin the previous section on aging, a clay can become overconsolidated along at least twopaths: (a) by a reduction of pressure after consolidation, resulting in swell due torebound in accordance to the "swelling lines" in v:P' space, or (b) by the "aging"process, whereby the specific volume is reduced under sustained constant loads.Regardless of the path, we'll assume that a clay has reached the stage ofoverconsolidation represented by point 4 in the v:P' plot of Fig. 5.16a. If sheared to

Figure 5.15. Possible compression and swelling paths for a clay undergoingaging and slaking.

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failure in a drained triaxial test, this clay would exhibit a relatively high peak strengthgreater than the critical state strength, and a volume change during shear characterizedby an initial decrease followed by a large increase. This behavior is illustrated by thecurve 4 in the plot of shear stress versus specific volume in Fig. 5.16b.

Suppose however, that instead of being tested, the sample were allowed to slake andswell under a constant P', until the state represented by point 3 on the v:P' plot wereobtained. The sample would then exhibit a reduced strength which is closer to thecritical state strength, and the amount of dilation following peak strength would bemuch less. The clay is therefore behaving as though it were less overconsolidated thanit was at point 4.

The major question to pose is whether the clay which has arrived at the staterepresented by point 3 by the process of slaking, will behave exactly the same as theclay that arrived at point 3 by aging, or simple rebound following unloading. If not,then it must be assumed that the boundaries which define the critical state modelchange in response to the process of slaking. However, if the behavior of these clays isthe same, then the critical state model can truly be considered as stress-pathindependent with regard to slaking.

As illustrated in Fig. 5.16, it seems possible that an overconsolidated clay can benaturally remolded by the slaking process to such an extent that the clay is completelyrejuvenated and approaches a normally consolidated state. This may account forobservations that the strength of many clay shales tends toward the "fully softened"strength, which is essentially the strength of a remolded, normally-consolidated clay(Skempton, 1970). Two factors may prevent a clay shale from fully softening to thestrength of a normally consolidated clay in the field. Due to the increase in thehorizontal stress component that accompanies swelling in the field, the clay may fail inpassive failure before obtaining the normally consolidated state by means of slaking.

Furthermore, an important question is whether there is a limit to the extent to which aclay shale can slake and swell. It may be possible that the amount of available energyreleased during slaking is limited to only that amount of strain energy gained duringaging. If that proves to be the case, then a clay can only regain the "youth" lost duringthe "aging" process, and not lost as a result of normal consolidation.

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It has been previously stated that the critical state line represents a stress condition atwhich no further changes in the stress state can occur. It is therefore interesting toconsider the potential effects of slaking on a clay which has reached the critical state.One of three responses is possible: (a) further slaking cannot occur and the materialyield surfaces are not altered, (b) slaking does not alter the yield surfaces, but insteadalters the stress state of the soil away from the critical state line, but within the confines

Figure 5.16. Possible loading/unloading paths (a) before and (b) duringdrained tests (modified from Atkinson and Bransby, 1978).

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of the yield surfaces, resulting in a hardening of the clay, or (c) slaking alters theposition of the yield surface defined by the previous critical state model.

Possible Complications Resulting From the Fissured Nature of ClayShales

The presence of fissures in clay shales may complicate the incorporation of the slakingprocess into the critical state model. These complications result from three aspects offissuration. First, unless a clay shale deposit is intensely intersected by randomlyoriented fissures, the fissuration results in material anisotropy. The anisotropy resultingfrom fissures may be superimposed on any anisotropic components that might existwithin the intact clay shale itself. Still, most geological materials in nature areanisotropic. Therefore, except for possible differences in magnitude, the complicationsresulting from anisotropy in clay shales are no different from those resulting fromanisotropy in all natural argillaceous materials.

The other two possible complications result from the probability that the softeningwithin clay shales is not homogeneous, but is at least initially concentrated along fissurewalls. In the discussion in the previous chapter, it was suggested that the softeningalong fissures could actually alter the shape and roughness of the potential failure plane,as well as reduce the shear strength between fissure walls. Therefore, in a fissured massundergoing slaking, the peak strength of the mass might be reduced to a greater extentthan that predicted by the critical state model.

In addition, when applying the critical state model, the water content is generallyassumed to be homogeneous throughout the sample. In a progressively softening clayshale, the water content varies from its highest values near the fissures to lower valueswithin the interior cores of intact clay. Therefore, the overall water content measuredfor the clay shale sample may be different from the water content within the failurezone, particularly if failure occurs primarily along fissures. However, as will bediscussed in the next section, the change of water content is more important than theabsolute value. During the initial stages of softening of a clay shale, the change in totalwater content is probably equal to that for the failure zone. In later stages of softening,however, the failure plane may already be located entirely within the softened portion ofthe clay shale. Any additional increases in the total water content would probably occur

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within scattered intact cores and therefore might not significantly alter the strength ofthe mass.

Testing the Critical State Model With Regard to Slaking

As discussed previously, the critical state concept states that the strength of any ideal,isotropic, overconsolidated clay is dependent only on two stress invariants, P' and Q',and the specific volume, v. As has been illustrated in Fig. 5.9, the stress path of a lightlyoverconsolidated clay is constrained by the Roscoe surface, while the stress path of aheavily overconsolidated clay is similarly constrained by the Hvorslev surface. A majortest of the applicability of the critical state model to the process of slaking is whetherthe Hvorslev surface can be defined for a clay shale which is undergoing softening byslaking.

The parameter, Pe', the equivalent pressure at any specific volume, has been discussed

previously, and can be obtained from the equation for the normal consolidation line, as

Pe' = exp[(N-v)/λ], (5.11)

where v is the current specific volume, the slope of the normal consolidation line inv:ln P space, and N is typically defined as the specific volume of the soil at P' equal to1 kN/m2 on the normal consolidation line. By normalizing the stress invariants with Pe,

as in Fig. 5.17, the complete Hvorslev surface can be defined by a line whose equationis

Q'/Pe' = g + h(P'/Pe'). (5.12)

In addition, the equation for the Hvorslev surface is given by

Q' = (M-h)exp[(Γ-v)/λ] + hP', (5.13)where M is the slope of the critical state line in P':Q' space, and Γ is arbitrarily definedas the value of v corresponding to P' equal to 1 kN/m2 on the critical state line.

The value for λ is generally obtained directly from consolidation data. However, as wasillustrated in Fig. 5.4, the slope of the critical state line in v:ln P space is

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approximately equal to the slope of the normal consolidation line, λ. Therefore, thevalue for λ can be obtained from either consolidation data or data defining the criticalstate line. In the analysis of experimental data in Chapter VII, the value of λ will beobtained using the critical state line.

It should also be noted that the use of the normal consolidation line for defining theequivalent pressure, Pe' is convenient, but arbitrary. The value of Pe' could just as

successfully be defined using the critical state line, or in fact any equivalent line with aslope of λ. By rearranging equation 5.13, we obtain

Q' = (M-h)[ exp(Γ/λ) / exp(v/λ) ] + hP' (5.14)

and

Q'/exp(-v/λ) = B + hP'/exp(-v/λ) (5.15)

where B is equal to the constant

Figure 5.17. The Hvorslev surface in normalized P':Q' space (Atkinson andBransby, 1978).

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B = (M-h) exp(Γ/λ). (5.16)

Thus, to prove the existence of the Hvorslev surface, and to define the slope, h,requires only one material constant, λ. Furthermore, the value of h depends on thechanges in the specific volume, v, relative to λ rather than on its absolute values. Thus,even though the total water content (or specific volume) may be different than thatwithin the failure plane, as discussed previously, the slope of the Hvorslev surface canstill be obtained as long as the changes in total and local water contents are the same.

Summary on the Role of Slaking Within the Critical State Model

The concept of critical state soil mechanics has proved to be a very concise and usefulmethod for analyzing the strength of a homogeneous, non-fissured, saturated clay. Theauthor proposes that it may also be equally useful for assessing the changes in strengthresulting from slaking in clay shale.

The importance of considering the process of slaking within the concept of critical statesoil mechanics cannot be overemphasized. The slaking history of a clay could beextreme complex, and nearly impossible to determine or predict. Furthermore, to studythe effects of slaking under laboratory conditions, the researcher must investigateseveral variables, including, but not limited to, the number of wetting and drying cycles,the length of each cycle, and the extent of drying during each cycles. Clearly, inaccounting for the effects of slaking within the field or laboratory, is would beextremely useful, if not vital, for the results to be dependent only on the final specificvolume, or water content, and not on the path along which that specific volume wasobtained.

The major question regarding the incorporation of both aging and slaking into thecritical state concept is whether these processes simply provide alternate paths alongwhich the specific volume of a clay can be altered within the confines of the definedcritical state boundaries, or whether these boundaries are themselves altered by theprocesses of aging and slaking. The critical state model has been shown to be pathindependent with regard to P' and Q'. It is important to likewise determine the extent ofpath independence with regard to the specific volume.

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The author has carried out triaxial testing on clay shale samples which have undergonevarious degrees of softening by slaking. These will be compared to similar test onunaltered clay shale in order to assess the feasibility of incorporating the slaking processinto the critical state soil mechanics concept. The test procedures are described in thefollowing chapter, while analysis of the data is presented in Chapter VII. In addition toinvestigating other aspects of the slaking process, the analysis will in particularconcentrate on the question of whether the peak strengths of the intact and softenedclay shale define a common Hvorslev surface.

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CHAPTER VI

EXPERIMENTAL METHODOLOGY

In Chapters II and III, much evidence was presented which emphasized the importantrole of fissure softening in affecting the engineering behavior of clay shales. Thetheoretical effects of fissure softening on the strength and stress-strain response of clayshales were discussed in detail in the Chapter IV. It has been shown that a primarymechanism by which softening may occur involves the process of slaking.

However, most of the present studies have been limited to the phenomena of slaking atzero confining pressures. Very few, if any, systematic studies have measured slakedurability in the presence of confining pressures, and this author knows of nopublished data which directly relates the process of slaking to changes in theengineering behavior of clays or clay shales.

The previous chapter introduced the possibility of accounting for the effects of slakingwithin the concept of critical state mechanics, and emphasized the usefulness ofstudying slaking within this framework. If the effects of slaking can be consideredwithin the framework of the critical state concept, the analysis of progressivelysoftening clay shale would be greatly simplified. Outside of this dissertation, thepresent author has found no experimental nor theoretical investigations concerning theeffects of slaking within the critical state model.

For these reasons, an experimental program was developed by the author, with theprimary objectives of (1) providing initial investigation into the effects of drying-induced slaking on the strength and stress-strain response of a highly fissured clayshale and (2) assessing the feasibility of incorporating the slaking process into thecritical state concept. These experiments primarily involved triaxial compression testson samples of Pierre shale, which had undergone various degrees of slaking whileunder a low confining pressures. Particular care was taken to record changes in thewater content occurring within the samples before, during, and after the tests. Theapparatus, and the sampling and testing procedures will be discussed in detail below.

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Testing Apparatus

Hoek cell. The triaxial testing was initially intended to be performed within Hoek cellscommonly used for the testing of rock. This apparatus had been chosen because of therather high strength of undisturbed Pierre shale, and because the Hoek cells at theUniversity of Colorado are capable of accurately measuring the radial displacement inthree directions. The Hoek cell consists of a stiff, metal outer cylinder and a flexible,semi-permanent, cylindrical membrane which surrounds the rock specimen. Theconfining pressure, which acts in the radial direction only, is applied by pressurizing oilwithin the cavity between the membrane and the outer frame. Six cantilever straingages rest against the membrane surrounding the sample and measure the radialdisplacement of the specimen. The radial strain gages are connected in a wheatstonebridge configuration, such that the bridge resistance can be measured and analyzed bya computerized data aquisition system outside of the cell.

Unfortunately, after an extensive period of rebuilding and calibration of the Hoek cell,the apparatus was found to be inadequate for the test program. The reasons for thiswere as follows:

(1) Although the thinnest membranes possible were used, these still proved to betoo stiff for testing at the rather low confining pressures used. This at timesresulted in a significant amount of undesired membrane support on the sample,and at other times, resulted in gaps between the membrane and sample.

(2) Samples often "jammed" after fracturing or swelling, making it difficult toremove the specimen after the test. Often the membrane was distorted ordestroyed during this process, requiring messy, time-consuming repairs.

(3) Samples were always destroyed during removal, thereby prohibiting any post-test examination of the sample. In addition, the metal cylinder prohibited anyvisual examination once the sample was placed into the cell.

(4) Because of the membrane problems, and because of physical and electricalproblems with the strain gages, measurement of the radial displacements wasoften unreliable.

The Hoek cell was therefore abandoned for the conventional triaxial cell commonly inuse in soil mechanics.

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Conventional triaxial cell. The conventional triaxial test cell was chosen instead ofother test cells for several reasons:a) the triaxial tests provides complete information regarding the stress state acting on

the sample and is flexible with regard to stress paths;b) unlike the direct shear test, the triaxial test accounts for the interaction between

fissure systems rather than strictly measuring the strength along a single joint;direct shear tests would therefore not be able to account for the significantreduction in interlocking which results from the rounding of the sharp corners atthe intersection of two fissures;

c) similarly, in the triaxial test the plane of failure is not predetermined, as is donewith the direct shear test; as has been shown, alteration of the failure plane duringfissure softening can act to greatly reduce the strength of the jointed mass;

d) the triaxial test results are not significantly altered by boundary effects, as are thoseof the direct shear test;

e) the triaxial test allows complete control over changes in the water content of thesample;

f) visual contact can be maintained with the triaxial cell, allowing assessment of thevarious modes of failure;

g) the classical research and discussions regarding critical state soil mechanics areclosely associated with the conventional triaxial test.

A schematic of the triaxial cell is shown in Fig. 6.1. The triaxial cell consists of aplexiglass outer cylinder and close-fitting, metal top and bottom. These are heldtightly in place by vertical bolts which resist the hydrostatic pressures created bypressurized water inside the cell. Within the triaxial cell, the sample is supported byupper and lower stainless steel pistons, which are also used to apply axial loads on thesample.

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Figure 6.1. Schematic of the triaxial test cell and measuring apparatus.

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The "blow-up" of the sample in Fig. 6.2, illustrates that the clay shale sample isenclosed in two very thin flexible membranes. Between the two membranes is a layerof silicone grease to prevent permeation of water through the membranes during longtests. Sample drainage and wetting takes place through the porous, ceramic stones atthe top and bottom of the sample. In later tests, water flow within the sample wasenhanced by four 3/4" strips of filter paper (VWR grade 615) running verticallybetween the sample and membrane. The plexiglass loading caps at the top and bottomof the sample, hold the porous stones and allow drainage to tubes at the side of theheads. The greased membranes are held tightly to the sides of the plexiglass loadingheads by rubber sealing rings, preventing flow of water out of the membranes. Thesteel loading heads, which fit into the plexiglass heads, are designed to evenly spreadthe vertical load applied by the upper and lower stainless steel pistons. A steel ballbetween the upper piston and metal head minimizes eccentric loads on the sample.

As shown in Fig. 6.1, the vertical load is measured by a calibrated loading ring on theupper piston, while the vertical displacement is recorded from a dial gauge betweenthe upper piston and cell body. A constant all-around pressure is applied to the sampleby means of pressurized water within the cell. This pressure is supplied by an air pumpand is held constant by a pressure regulator.

Any changes in the volume of water in the cell can be measured from the pipette,which also acts as an air-water interface. At constant cell pressure, this change in cellpressure results primarily from changes in sample volume, and from changes in thedisplacement of water by the piston as it moves in or out of the cell. A constant "backpressure" is similarly applied to the water within the sample, such that the volume ofwater flowing in or out of the sample can be recorded from a second pipette. This dualpipette system provided a means of "cross checking" the changes in sample volume,and as will be discussed later, proved very useful for recognizing problems in thetesting procedure. The glass, graduated pipettes were enclosed on both ends usingcoupled brass pipe joints, with a rubber "O" ring between. Tightening of the brassjoints onto the "O"

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Figure 6.2. Schematic of the test sample assembly.

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The process of slaking was carried out under conditions of constant hydrostatic stress.During these periods, the upper piston was loaded with weights in order tocompensate the upward force generated by the hydrostatic pressure of the cell water.Therefore, the sample was able to deform both radially and vertically in response towetting, while the stresses remained constant. However, after slaking, the clay shalewas then loaded vertically to failure using a displacement- controlled Wykeham-Farrance loading frame. Using displacement-controlled rather than stress-controlledloading allowed the assessment of post peak stress- strain behavior.

With this load frame, various sized gears can be set to provide for different rates ofloading. For the first series of tests in the triaxial cell, the loading gears were set toprovide a loading rate of 0.0018"/min. For later tests, the gears were adjusted so thatthe loading rate would initially be 0.0018"/min and then adjusted to 0.0008"/min priorto sample failure. Primarily because of the flexibility of the load ring, the loadingsystem was not very stiff, resulting in some strain energy being stored by the loadingsystem during loading.

One effect of this lack of stiffness was that the actual initial rate of loading experiencedby the sample was greatly reduced to approximately half of that predicted by the gearratios. The actual rate of loading gradually increased at higher stresses, but never quitereached the rates given above. Although this slower rate of loading was notparticularly undesirable, the presence of strain energy stored by the machine did createthe undesired effect of rapidly loading the sample after peak strength was reached.This primarily occurred with unaltered samples which were able to resist high stressesbefore failing, and was somewhat minimized by turning the loading machine off untilequilibrium was again maintained.

For the present tests, the post-peak strength was more important than the stress-straincurve after failure. Therefore, this lack of machine stiffness did not significantly affectthe desired results. However, for those test programs where the post-peak stress-strain behavior is important, it is strongly beneficial to employ either anelectronic pressure transducer in place of load rings, or a variable stiffness loadingsystem, such as that presented by Sture (1976) and Sture and Ko (1978).

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Sample Preparation

The Pierre shale is well-suited for the desired test program for several reasons: (a) inits undisturbed state, the material has a high peak strength, and exhibits a high ratio ofpeak to residual strengths, (b) the difference in strength between the unaltered andsoftened states is very drastic, (c) the material is highly susceptible to slaking, rapidlyslaking from a rock consistency to mud after only one cycle of drying and wetting, (d)the shale exhibits high swelling pressures in response to rewetting after drying, and (e)the shale readily forms horizontal and vertical fissures during drying.

Block sample retrieval. Several blocks of Pierre shale were dislodged by backhoefrom an open pit quarry of the Ideal Cement Company located about 15 miles south ofBoulder, Colorado along state highway 93, in sec 5, T 1S, R 70W. The composition ofthe Pierre shale was determined by Braddock and Machette (1976) and is given inTable 6.1. The excavated blocks ranged in size from one to a few feet per side. Inorder to prevent deterioration due to drying, these blocks were immediately placed inseveral ten-gallon drums filled with water. The drums were then placed in a wet roomat the University of Colorado where they remained for some ten years before beingcored for the present test program. The blocks of Pierre shale do not appear to havesuffered any significant alteration as a result of stress unloading or soaking in water.However, if any of the blocks or fragments of the Pierre shale is allowed to evenpartially dry, and is then rewetted, the shale will rapidly break down into small chunksand then to soft mud in a matter of 15 to 20 minutes.

Sample coring and trimming. In order to perform the desired triaxial tests, it wasnecessary to obtain NX sized cores (2-1/8in or 56mm) from the large blocks. Becausethe Pierre shale is both hard and highly susceptible to softening, this proved to be verydifficult. The hardness of the Pierre shale required the use of a diamond-studded, rotary coring drill. The cores used in this study were obtained using a coringdrill mounted in the laboratory.

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During the initial attempts at coring, the downward driving force was manually appliedby the machine operator. However, it was difficult to manually maintain a constantmoderate force, and this method proved fruitless for the Pierre shale. During even briefperiods when excessive force was unavoidably applied, two undesired effectsoccurred: (a) the core experienced high torque stresses and twisted apart, and (b) largeamounts of heat was generated by friction, resulting in drying of the clay shale, andsubsequent softening to mud with rewetting. Some success of obtaining cores of Pierreshale in the field has been achieved using pressurized air as the drilling "fluid".However, a crude attempt at air drilling in the laboratory resulted in drying of thematerial and "locking up" of the sample inside the core tube.

Table 6.1. Calculated mineralogy mode of Pierre shale using microscopic andx-ray data (Braddock and Machette, 1976).

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Finally, successful core samples were obtained in the laboratory, using moderateamounts of water, and a constant, moderate downward force supplied by weightsattached to the coring drill. This configuration resulted in a coring rate of about oneinch of core length for every three minutes of drilling time. After washing off the mud

Figure 6.3. Photograph and sketches of dried cores of Pierre shale showingdifferent orientations of fissure pattern (photo by author).

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along the sides, the core samples exhibited very smooth and straight sides, with nofurther sign of alteration.

In order to eliminate any anistropic effects during the primary test program, all ofthese samples were cored perpendicular to the dominant plane of fissuration. Thisplane corresponds to either the original bedding plane, the plane perpendicular tovertical unloading, or probably both. The dominant plane of fissuration is not obviousin wet, undisturbed samples, but can be readily distinguished by the pattern offissuration resulting from drying. As illustrated by the photos in Fig. 6.3, dried Pierreshale samples exhibit one well- defined, primary direction of uninterrupted fissuresaccompanied by one or more sets of secondary fissures. Although the orientation ofthe secondary set is varied within the same sample, it is always roughly perpendicularto the primary set. The samples used for the test program were all cored perpendicularto the primary plane of fissuration, with the exception of two samples which werecored about 70o from horizontal in order to assess the effects of sample orientation.

After coring, the samples were trimmed on the top and bottom in two stages: the firstusing a rotary diamond rock saw to make initial cuts, and the second involving levelingand smoothing of the top and bottom, using a lapidary polishing table. In order tomaintain a length to diameter ratio of about two, the samples were trimmed to lengthsof about 4.2 inches. Due to overpolishing to remove chipped corners and to level thetops and bottoms, most samples were slightly shorter than 4.2 inches, although nonewere less than 4.02 inches.

Great care was taken to prevent any drying of the clay shale at all times during andafter the sample preparation procedures. The material was continually wetted duringall operations and samples were immediately resubmerged in tap water afterward.

Sample drying and the inducement of fissures. The proposed test program required a statistically significant number of fissureswithin the samples in order to enhance wetting and to, of course, examine the effectsof fissure deterioration within the clay shale. Since the Pierre shale does not naturallyexhibit an abundance of open fissures at the scale of the samples, it was necessary toinduce these fissures by some reproducible method.

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Preliminary studies determined that an extensive network of fissures could best becreated and reproduced by the drying process itself. It was desirable to dry the samplesunder a confining pressure while in the cell. However, because of the extremely lowpermeability of unfissured Pierre shale, it proved impossible to initially dry the samplesin the cell within a reasonable period of time. In addition, the pattern and surfaces ofthe fissures produced by oven- or air-drying the material were found to be very similarto those observed for the Pierre shale in the field. It was therefore decided that thesamples would first be dried outside of the cell, and then wetted under confinementwithin the cell.

It was further determined that the spacing of the fissures varied with the intensity ofthe drying process. Oven-dried samples created fissures with spacings of 1-2 mm(0.04-0.08 inches), while air-drying produced fissures every 6-15 mm, or every 0.2-0.6inches. The oven-dried samples appeared too intensely fissured; the samples used inthis test program were therefore air- dried for at least 30 minutes to induce fissuration.This was carried out inside of a laboratory cabinet in order to minimize the influence ofdrafts and sudden temperature changes.

It is important to re-emphasize that the purpose of these tests was to assess thepotential effects of slaking and swelling with fissures of clay shale, and not necessarilyto provide design parameters for the Pierre shale. Although it is desirable to modelconditions in the field as closely as possible, this was not a requirement of the testprogram. Still, the author feels that the conditions created in these tests are verysimilar to those occurring in the field at a larger scale and over a longer time. Thesamples in this test program might therefore be considered as miniature models of ajointed clay shale mass, on which is imposed an accelerated version of the slakingprocess.

Test Procedure

General Consideration. The effect of slaking on a given material is dependent tosome extent on several external variables, as well as those inherent material factorssuch as clay mineralogy or amount of stored strain energy. These external factorsinclude: (a) the hydrostatic and deviatoric states of stress acting on the material, (b)the length of drying periods, (c) the length of wetting periods, (d) the number of

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wetting and drying cycles, and (e) the chemistry and temperature of the slaking fluid.The two variables investigated in this test program were the time periods of drying andwetting. All slaking was carried out under isotropic confining pressures of 10 psi (69kPa) for one cycle of drying and wetting, using distilled, de-aired water at roomtemperature.

The testing procedure involved four stages consisting of: (a) drying of the sample to a predetermined water content, under zero confinement,in order to induce fissuration and to initiate the drying-wetting cycle, (b) wetting of the sample under constant confining pressure of 10 psi, to induceslaking and softening, (c) strain-controlled deviatoric loading of the sample under confining pressures of10, 30, and 50 psi, in order to determine the strength and stress-strain response, and (d) post-testing treatment, consisting of either the determination of the variation ofwater content within the specimen, or the impregnation of the specimen withCarbowax, in order to preserve it for fabric studies.

Throughout all stages, changes in the sample volume, sample weight, or both, weremeasured so that the water content of the specimen could be determined at any time.Two sets of tests were performed in the conventional triaxial cell. The data from thefirst series of tests were affected by some undesirable consequences of the testingprocedure. The test procedure was then improved, resulting in some very informativedata from the second series of testing.

Initial conventional triaxial tests. During the initial testing series in the conventionaltriaxial cell, eight tests were performed in order to assess the feasibility of the testprogram, and to determine which variables should be investigated during the primarytest program. The three variables which were studied during these initial tests included(a) the confining pressure acting during the slaking stage, (b) the length of the dryingand wetting period, and (c) the number of wetting and drying cycles. After the slakingstage, all of these samples were loaded to failure under a confining pressure of 30 psi,with a back pressure of 15 psi in the sample. The maximum rate of loading was0.0018"/min.

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As will be discussed in the following chapter, these initial tests provided usefulinformation regarding the effects of slaking on the strain response of Pierre shale.However, the utility of these initial results is somewhat limited by two factors. First,portions of some of the samples were not thoroughly saturated before testing, as waslater evident by the presence of well defined saturation lines near the top of a few ofthe specimens. Failure in these samples was typically restricted to the saturated zones.

Furthermore, evidence indicates that negative pore pressures, generated duringdilation, were not sufficiently dissipated by the samples. As discussed previously,volume changes were measured for both the cell and sample water. If the volumechanges in the cell water are corrected for the volume of the piston, then these changesresult solely from compression or dilation of the sample. Changes in the samplevolume similarly result from the flow of water out of and into the sample, in responseto compression and dilation, respectively. Under ideal test conditions, in which thedegree of saturation of the sample does not change, the changes in these two volumesshould be of equal magnitude but opposite in signs.

Typical changes in the volume of sample water are compared to the corrected volumechanges of cell water for several of the initial tests are plotted in Fig. 6.4. As shown inthe figure, these volume changes are rarely equal and indicate that the volume of waterin the sample changes at a slower rate than the corrected volume of cell water.Therefore, the sample is decreasing in volume without the removal of an equal amountof pore water. Since water is essentially incompressible, these results probably indicatethe presence of gases within the pores, and therefore the incomplete saturation of thesample. In addition, after the peak strength has been reached, dilation of the sample isevident from the sharp increase in volume as exhibited by the cell water. However, inmost of the initial test, the volume of water in the sample changes very little afterdilation begins, possibly indicating that the flow of water back into the sample is notrapid enough to balance the negative pore pressures created during dilation.

In summary, the results from these initial tests indicated that the wetting was in generalnot adequate to sufficiently saturate the samples. Furthermore, the negative porepressures generated during dilation at failure, were not dissipated rapidly enough

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by the sample. These two undesired effects result from two major causes: (a) the backpressure acting on the pore water was too low, and (b) the rate of loading at failure istoo high relative to the rate of pore pressure dissipation.

Figure 6.4. Irregular volumetric data from tests on two poorly saturatedsamples.

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The benefits of increasing the back pressure of the pore water are three-fold. First, athigher pore pressures, it has been shown experimentally, as well as theoretically, thatless gas will come out of solution given an equal drop in fluid pressure (Bishop andHenkel, 1962; Lowe and Johnson, 1969). Second, cavitation is essentially prevented athigher back pressures since these pressures are unlikely to become tensile withdilation. Finally, higher back pressures allow for greater pressure gradients during thepercolation process, and therefore more effectively counteract any gravitational effectswhich might prevent complete saturation.

Pore pressures can be eliminated or minimized during testing either by reducing therate of pore pressure generation or by increasing the rate of pressuredissipation. In practical terms, these can be accomplished by decreasing the rate ofloading or by increasing the flow of water into and out of the sample. The problemsencountered during these initial triaxial tests were avoided in the final test series byimprovements in the testing procedure.

Improved conventional triaxial tests. The final series of tests similarly employed theconventional triaxial cell, but included several improvements in the testing procedure:(a) the back pressure was increased from 15 psi to 60 psi (or 414 kPa),(b) the initial rate of loading remained at 0.0018" per minute, but was significantly

reduced to 0.0008"/min well before failure,(c) filter paper strips were placed along the length of the sample in order to greatly

facilitate flow to and from the sample,(d) the time periods allowed for wetting and consolidation were increased, and(e) the procedure for measuring the final water content of the sample after failure was

refined.

The changes in back pressure and the addition of filter paper strips resulted in rapidand more complete saturation of the initially dried samples, as well as more rapiddissipation of excess pore pressures imposed by hydrostatic and axial loading.According to Bishop and Henkel (1962), the radial drainage generated by the presenceof the vertical filter strips can alone increase the rate of flow into and out of

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the sample by a factor of 10. In addition, the rate of loading near failure, and thereforethe rate of excess pore pressure generation, was reduced to less than half of the initialvalue.

In contrast to previous tests, the data from this third series of tests showed nosignificant deviations between the amounts of water flowing into and out of thesample, and the amounts flowing to and from the cell. It can therefore be assumed

Table 6.2. Procedure checklist for triaxial tests on intact and softened Pierreshales samples.

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that the samples were sufficiently saturated and that any dissolved air in the pore fluidremained in solution during the dilation of the sample. This at the least indicates thatany excess pore water pressures (negative or positive) that might have been generatedwere not high enough to alter the solubility of the pore water. For reference, the

Table 6.2 (cont)

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procedure check-list for the final series of tests is presented in Table 6.2. All stepsinvolving critical measurements are boldfaced. Great care was given to themeasurement of sample volume changes before, during, and after deviatoric loading ofthe sample. The procedure for determining the water contents will be discussed inmore detail in the next section.

Water content measurements. In order to determine the initial volumes of water andsolids in the samples, water content measurements were performed on six randomlysampled disks of saturated Pierre shale. The samples were weighed before and afteroven drying. Initial sample volumes were determined both by direct measurement withcalipers and by recording the volume displaced as the sample is immersed in calibratedcylinder. The results of these measurements gave the following average values:

unit density = 2.23 - 2.26 gm/cm3

porosity (n) = 26%

void ratio (e) = 0.35 - 0.36

water content = 13.2%

specific gravity of solids 2.65 .

As indicated by the procedures check-list in Table 6.2, four critical values weredetermined at various stages of testing: (1) the initial sample volume, (2) the initialsample weight, (3) the final sample weight after removal of the sample from the cell,and (4) the volumetric changes recorded by the pipettes during deviatoric loading ofthe sample. By assuming the initial porosity (n) of 0.26, as determined above, thesample's specific volume can be calculated at any point during the consolidation orshear testing procedures, as follows:

(VW)i = (VT)i * n (7.1)

(VS)i = (VT)i - (VW)I (7.2)

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(VW)f = (VW)i + [(WT)f - (WT)i] /γ w (7.3)

VW(t) = (VW)f - Vp(t), (7.4)

where V and W are volume and weight, respectively; the subscripts i and f indicateinitial and final values, respectively, while the inner subscripts T, W, and S indicatetotal, water, or solids parameters. The values n and γw are the initial porosity (=0.26)and the unit density of water; Vp(t) is the difference between the final volume readingof the pipette and the pipette reading at any time, t; and VW(t) is similarly the water

volume of the sample at any time. In addition, the amount of water lost during thedrying stage can be determined from the volume and weight measurements recordedimmediately before the sample is placed in the cell.

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CHAPTER VII

TEST RESULTS AND ANALYSIS

The test program described in the previous chapter provided several types of data,including volumetric changes during drying and rewetting, volumetric and stress-strainresponse during shear, and both peak and ultimate strength values. Analyses usingthese data examined: (a) anisotropic shrinkage and swelling in response to drying andwetting, respectively, (b) rate of drying, (c) changes in Youngs Modulus and PoissonsRatio in response to slaking and swelling, (d) reduction of strength in response toslaking and swelling, (e) normalization of stress-invariants, P and Q, to establish theexistence of a Hvorslev surface and thus assess the role of slaking in the critical statemodel, and (f) assessment of strength anisotropy resulting from fissure orientation.

In this dissertation, the individual tests will be coded according to the followingscheme:

CP/TD-TW (for example 30/4-23),

where CP indicates the confining pressure during shear loading in psi units, TD theapproximate time of drying in days, and TW the approximate time of wetting in days.Therefore, the example above indicates that the sample was dried for 4 days, wetted for1 day, and then tested at a confining pressure of 30 psi. In addition, a value of TD ="H" indicates that the time of drying was one half hour, while the letter "A" or "B" atthe end of the test code designates the first or second run of duplicate tests. Finally,"HA" at the beginning of the code indicates one of two special "high angle" tests, inwhich the samples were cored at a highly oblique angle.

All data processing and analyses were performed on a Corona PC computer (IBM PCcompatible). Most plots were produced on a Hewlett-Packard HP7470A plotter.

Volumetric Changes During Drying and Wetting

Drying. As previously discussed samples of Pierre shale were air dried underconditions of zero confinement, for periods of up to 28 days. These samples were

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weighed at various times during the drying process, and their dimensions measuredwith caliber’s at the beginning and end of the drying period.

Since the drying procedure was independent of the test apparatus, the shrinkage dataincludes results from all three-test series. In addition, continuous weighing of thesamples during drying allowed fairly extensive records of the changes in water contentwith time. These results, plotted in Fig. 7.1, exhibit a relatively rapid exponentialdecrease in water content from the initial value of 13.2% to a value of 4% to 6% after100 hours, or about 4 days. After this, the rate of drying is greatly reduced, such thatthe water content is only 3% after 677 hours, or about 28 days.

Values for void ratio, saturation, and shrinkage strains, measured at the end of thedrying periods, are presented in Table 7.1. These results indicate that the dryingprocess is accompanied by a decrease in the void ratio from an initial value of .35 toabout .24 after 4 days, and .22 after 28 days. Again, the major shrinkage is shown tooccur within a 4 day period. Similarly, the saturation decreases rapidly in 4 days to 50or 60%, but continues to show a significant decrease in 28 days to a value of 34%.

Figure 7.1. Drying curve showing the rate of decrease in water content withincreased time of open-air drying of Pierre shale samples; based on data from

several test

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The values of saturation and void ratio measured during the drying process aredetermined from measurements of the sample volume, as well as its weight, and aresomewhat more scattered than the values for water content. This scatter resultsprimarily from the opening and closing of fissures which occur during the drying

Table 7.1 Water content (w%), specific volume (v), saturation (S%), andshrinkage strain data for Pierre shale samples after open-air drying.

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process. Although fissures are generally not evident in intact cores of Pierre shale,microfissures may exist at a scale too small to see with the unaided eye. Visualinspection during drying indicates that these fissure tend to open, or possibly form,within 10 to 30 minutes after the initiation of drying. Many of the fissures open about 1mm during the drying period between 1 hour to 40 hours, presumably as a result ofdifferential shrinkage between the exterior and interior of the sample. After 45 hours,the fissures begin to close, such that by 70 hours, or less than 4 days, the fissures areonce again tightly closed.

Thus, three observations suggest that although drying may be initially concentratedalong fissures, drying within the sample is for the most part uniform after four days.These include the rapid drying to low water contentswithin four days, the absence ofsignificant changes in sample volume after four days, and the uniform shrinkageindicated by the reclosing of the fissures. In addition, the lack of "barreling" in the driedsamples indicates that the drying along the vertical axis was uniform, even though thetop and bottom of the samples acted as free surfaces.

As was demonstrated in Fig. 5.3, the Pierre shale exhibits a preferred direction forfissuration upon drying. All samples except two were cored perpendicular to thisdirection, which probably represents the horizontal plane of deposition. The datapresented in Table 7.1, indicate that the shrinkage strain is anisotropic, with axialshrinkage being about two times greater than in the radial direction. The magnitudes ofshrinkage are rather high with about 2% vertical strain in one day, about 4.5% in 4days, and 5% in 28 days. Some negative strains, indicating expansion, were measured in samples which had dried for less than a day, and probably represent the influence offissure opening during initial drying. The shrinkage ratios reported above are inagreement with those reported by Fleming et al (1970) for the Pierre shale in SouthDakota. The implications of these shrinkage strains will be discussed later in thissection.

Rewetting. As discussed previously, the samples which had undergone drying, werethen placed in the triaxial cell and rewetted for periods ranging from one to 23 days.Because of testing difficulties, only the data from the final series of tests are consideredreliable, and thus, all rewetting of samples occurred under 10 psi confining pressures.

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The samples were free to swell in all directions in response to rewetting in the triaxialcell.

The vertical displacement of the sample undergoing wetting was measured directlyusing strain gages. In addition, the volumes of the samples at the beginning of shear,were calculated using the method described in Chapter VI. Thus, the axial and radialstrains resulting from rewetting were determined from these measurements, as were thevalues for void ratio, water content, and degree of saturation. The results of theseanalyses are presented in Table 7.2, along with the results reported previously for thedrying phase. By necessity, the values were calculated for test conditions immediatelybefore the application of shear stresses, and therefore include the effects ofconsolidating the sample from 10 psi to the confining pressure applied during sheartesting. Therefore, the values for samples tested at 30 and 50 psi, probably indicatesomewhat less swell than what had actually occurred.

The data in Table 7.2 indicate that the amount of swelling during rewetting is moreaffected by the length of the previous drying period, rather than the length of thewetting period. Samples which had dried for periods of only half an hour, merelyregained the volume lost during the drying period. In fact, the final water content andvoid ratio values for these samples are essentially equivalent to those for samples whichhad not experienced any drying period.

However, samples which had been previously dried for 4 days or more, swelledsignificantly upon rewetting, and in fact swelled to water content values which weremuch higher than the initial value of 0.35. This increased water content is undoubtedlythe result of the breaking of bonds associated with drying-induced slaking. The amountof swell resulting from slaking was high, indicated by water content values of about .40for samples which had dried for 4 days, and a value of .51 for the sample dried for 28days.

In Fig. 7.2, the rate of drying is indicated by plots of void ratio and degree of saturationover time. As indicated previously, the data show a rapid rate of drying prior to 4-5days, followed by a much slower rate. As the degree of saturation returns to 100%during rewetting, the sample swells beyond the initial void ratio of 0.35. It is

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Table 7.2 Water content (w%),specific volume (v), saturation (S%), andshrinkage strain data for Pierre shale samples after open-air drying and

rewetting.

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Also apparent that the minor amounts of water lost after four days of drying can have avery significant effect of the amount of swelling during rewetting.

The axial and radial strains presented in Table 7.2 are scattered and somewhatunreliable. However, it appears that the axial strain is again greater than the radialstrain, although the swell ratio is generally less than the value of 2 to 3 reported for the

Figure 7.2. Drying and wetting curves as a function of drying time for thePierre shale.

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shrinkage ratio. Finally, the rather large value of almost 9% for the axial swell ofsample 30/28-3 should be noted.

Summary of the drying and wetting data. Core samples of Pierre shale exhibitedsignificant shrinkage and swelling in response to drying and rewetting, respectively.Axial strains of over 4% were measured after 4 days of drying, while one sampleexhibited axial strain of over 5% during 28 days of drying. Rewetting of these samplesresulted in excessive swell beyond that required to return the sample to its originalstate. Axial swell of almost 9% was measured for the sample which had undergone 28days of drying, resulting in an increase in the void ratio from an initial value of 35% tothe final value of 51%. This significant increase in water content after drying andrewetting is certainly the result of bond breakage during drying-induced slaking.

The length of the period of drying experienced prior to rewetting was far moresignificant than the subsequent time of wetting. Samples which had dried for periodsless than one hour, behaved essentially the same as those samples which had not beenpreviously dried. Still, the most significant changes in void ratio, water content, andstrain occurred within 4 days of drying. On the other hand, the degree of saturationcontinued to decrease significantly even after 28 days of drying.

The Pierre shale exhibits a strong preferred orientation of fissuration during the dryingprocess. In addition, anisotropic behavior was observed during drying and rewetting.Shrinkage in the axial direction was generally 2 to 3 times that in the radial direction,while the axial to radial swell ratio during rewetting was typically slightly less than 2.Visual inspection of the samples during drying did not indicate any distortion of thecore samples, implying that the Pierre shale may be transversely isotropic with respectto the coring axis. This axis is assumed to correspond to the vertical direction in thefield.

The strains observed during drying are probably elastic, and can thus be represented bythe following equations:

εr = [σr(1-ν ) - νσ a ]/E (7.1)

εa = [ σa - 2 νσr ]/E (7.2)

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where σr and σa are the radial and axial stresses, respectively, εr and εa are the radial

and axial strains,and E and ν are Youngs Modulus and Poissons Ratio, respectively. Thus, assumingthat ν and E are constant for both directions

εa/εr = [ σa - 2 νσr ]/[ σr(1-ν ) - νσa ]. (7.3)

Using values of εa/εr = 2, and ν = .25, which

are consistent with data presented in this dissertation, rearranging and solving eqn. 7.3for σa/ σr indicates that the axial stress acting on the sample would need to be 1/3

greater than the radial stress in order to obtain the difference in strain observed duringdrying.

However, the axial and radial stresses are presumably equal during the drying process,implying that the differences in axial and radial strain must result from directionalvariations in ν and E. This anisotropic nature could indicate (1) the transversealignment of clay particles during deposition and compaction, (2) previous in-situswelling in the vertical direction during unloading, or both.

Stress-Strain Response During Shear

After drying and rewetting, the samples of Pierre shale were tested in triaxialcompression under conditions of vertical strain-controlled loading. In Table 7.3 is anoutline of the series III tests, from which the remaining information has been derived.The primary set of tests (Set A) involved systematic variations of two parameters: thetime of drying and the confining pressure during shear loading. For Set A tests, theconfining pressure varied between 10, 30, and 50 psi, while the length of drying variedbetween zero, one half hour, and 4 days. The period of rewetting during these tests was1 day. In addition, a second set of tests (Set B), were all performed at 30 psi confiningpressure on samples which had undergone other variations in the drying and wettingtimes. These allowed evaluation of the relative significance of the length of

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drying and wetting periods on the engineering behavior of the Pierre shale. Finally, twosamples (Set C) were cored at angles which were highly oblique to the other samples,thus allowing assessment of the effects of fissure orientation.

Various results for all series III tests are plotted in Appendix E, including plots relatingaxial, shear, and volumetric stresses to relevant strains. In addition, other test resultswill be presented and discussed in the following sections.

Table 7.3. Table of testing program on Pierre shale.

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Axial stress-strain response. In Fig. 7.3 are presented representative plots of axialstress, axial strain, and volumetric strain, which indicate that there is essentially nodifference in response between samples which have experienced no previous drying andthose which have undergone only half an hour of drying. This implies that thefissuration which becomes visible after about 15 minutes of drying, may actuallyalready be present, but tightly closed, in unaltered samples.

The stress-strain response for unaltered Pierre shale is typical for heavilyoverconsolidated clays and shales. It is characterized by a steep increase of axial stressup to a load sufficient for abrupt failure. With continued shear, the strength rapidlydecreases from this peak value to a lower value, sometimes referred to as the "ultimate"strength (note: the term "residual" strength is not used here, because it is questionablewhether the true residual strength is obtained in any such triaxial compression tests). Inthe unaltered Pierre shale, this decrease in strength after failure is quite drastic, with theultimate strength ranging from 28% to as low as 20% of the peak strength.

The volumetric strain during shear is characterized by an initial compression of up to1.5 to 2.5% at failure, followed by sample dilation immediately before and followingfailure. In most cases, the sample dilates to volumes well above the initial samplevolume. In heavily overconsolidated or bonded materials, the dilation following failureis generally concentrated along the plane or zone of failure, and is therefore indicativeof the opening of a crack rather than a general expansion of the entire sample. Forthese reasons, volumetric measurements after peak failure are somewhat misleading,and should be examined with caution.

In Fig. 7.4, overlapping stress-strain plots of "well-behaved" samples illustrate that, forsamples which have undergone similar drying and rewetting schedules, the curves arevery similar even at different confining pressures. However, with higher confiningpressures, the samples tend to fail at higher axial strains, and thus sustain greater axialstress and volumetric strain. The relationship between the axial stress and axial strain isbest described by the Youngs modulus, E, defined here as the ratio of the change ofaxial stress to the change of axial strain, while ratio of the change in radial to axialstrain is given by Poissons ratio, ν . Thus, for samples with the same drying and

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Figure 7.3. Representative stress-strain plots of samples having undergone nodrying (50/0-1) and half-an-hour of drying (50/H-1) prior to wetting and shear

testing.

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Figure 7.4. Representative stress-strain plots of samples having undergonesimilar wetting- drying cycles but tested at different confining pressures.

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rewetting history, the values for Youngs modulus and Poissons ratio essentially remainconstant.

However, samples which have undergone longer periods of drying and rewettingexhibit significantly lower values of E, while the values of ν do not change asdrastically. This is further illustrated by the plots presented in Fig. 7.5, which show verysignificant softening of the samples as the time of drying is increased from half an hourto 28 days.

Upon closer inspection, the prefailure portions of these curves can be divided into threeparts, as in Fig. 7.6. The first stage of the stress-strain curve is generally short and veryshallow, and probably represents initial fissure closing during loading. The remainder ofthe loading curve exhibits two well-defined linear segments. These segments appear inthe plots of axial strain versus volumetric strain, as well, although the location of thebreak between these curves may be slightly shifted.

The values of E and ν have been calulated for the second and third segments of thesecurves, and are presented in Table 7.4. For segment II, the value of Youngs modulus,E, decreases with softening from an average of about 28,000 psi for unaltered Pierreshale, to about 9,000 psi for samples which had undergone 4 days drying, and finally toa low of only 4,500 psi with 28 days of drying. This represents about an 84% increasein the sample compressibility. Youngs modulus for segment III exhibits a similardecrease from 22,000 psi to about 5,000 psi with softening.

As illustrated in Fig. 7.7, similar two- or three-segment loading curves have beenobserved for other stiff, or cemented, clays (Uriel and Sarrano, 1973; Conlon, 1966;Davidson, 1977). It has been suggested that the shape of such curves results from thesuperposition of frictional and bonding components of material strength. This effect isillustrated schematically in Fig. 7.8. From both Table 7.4 and the plots discussed above,the loading curves for unaltered samples of Pierre shale are similar to curves for otherbonded clays, in that the initial loading segment is steeper than the subsequent segment.Thus, in view of the interpretation discussed above, shearing of the unaltered Pierreshale is initially resisted by bonding between the clay particles. After this bonding isdisrupted, probably as a result of shearing through asperities, further resistance isprovided by the frictional component of strength.

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Figure 7.5. Comparative stress-strain plots showing drastic softening of Pierreshale after slaking.

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Figure 7.6. Representative stress-strain plot showing three distinct segmentsin the prefailure curve.

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However, for samples of Pierre shale which have previously undergone 4 days ofdrying, the slopes of these two segments are essentially the same. With furthersoftening, the loading curve actually becomes steeper in segment III than it was insegment II. In Chapter IV, it was proposed that the softening of the fissure walls mightprogress at a faster rate than that of the interior cores of the clay shale. If one considers

Table 7.4. Values for Youngs modulus, E and Poissons ratios, ν, for loadingsegments II and III for unaltered and softened Pierre shale samples.

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Figure 7.7. Two- and three-segment loading curves for other stiff, orcemented, clays (Uriel and Sarrano, 1973).

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the theoretical implications discussed in Chapter IV, the strain hardening observed insoftened Pierre shale may still be in agreement with the interpretation presented above.However, in such samples, the bonded material remaining between fissures is separatedby the softened fill along the fissures. Thus, the shear load is initially resisted only bythe weaker frictional component of the softened fissure material, as represented bysegment II. Then with further shear displacement, the less softened clay clumpsbetween fissures begin to come in contact, thereby adding a slightly stronger bondingcomponent to the overall strength. This would result in a slight steepening of theloading curve, as observed in the samples of softened Pierre shale.

The values of Young’s modulus discussed above were calculated using the least-squares method for each loading segment. However, Young’s modulus was alsocalculated incrementally for each step of axial loading. In Fig. 7.9, these values for all

Figure 7.8. Schematic explaining two-segment loading curves, resulting fromthe combining of frictional and bonding resistance forces.

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Figure 7.9 Plot showing the relationship of Young’s modulus to specificvolume for loading curve segment III in Pierre shale.

Figure 7.10. Prefailure P'-v paths for all Pierre samples tested at 30 psiconfining pressure

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tests are plotted against the specific volume, v, indicating a well defined relationshipbetween Young’s modulus and the specific volume. The modulus for segment IIIremains greater than that for segment II, for specific volumes less than about 1.4. Atvalues greater than 1.4, the Young’s moduli are essentially equal for both segments.

Volumetric strain during shear. The changes in sample volume during shear arereflected in the changes of specific volume and volumetric strain. In Appendix E, thespecific volume measured during testing is plotted against the natural log of the stressinvariant, P, for each test. In addition, the prefailure P-v paths are plotted collectively inFig. 7.10 for samples tested at 30 psi confining pressure.

These curves exhibit a relatively stiff behavior during initial loading, followed by anabrupt decrease in specific volume with continued loading. This abrupt change involumetric compression probably indicates a change from relatively stiff elasticbehavior to more plastic deformation. With confining pressures of 30 psi, the changefrom elastic behavior occurs at P'= 90 psi (Q'= 180 psi) for unaltered Pierre shale, andat very low values of P'= 35 psi (Q'= 15 psi) for softened samples. In addition, thesecurves indicate that the volumetric strains associated with plastic behavior, increase forsamples which have undergone softening in response to slaking. This is evident fromthe increased slopes for samples with higher specific volumes.

The values for ∆v/∆P and ∆v/∆ln(P) have been tabulated in Table 7.5, for the loadingcurve segments II and III. These data give some indication of the relative volumechanges that occur during the shear tests. However, the volumetric changes measuredby the specific volume, v, reflect changes resulting from both shear and isotropic ("all-around") stresses. In contrast, the values for the volumetric strain reflect the changes involume resulting only from changes in the all-around stress and are given by

εv = - ∆V/Vo = εa + 2 εr , (7.4)

where V and Vo are the present and initial sample volumes, respectively. Values for

∆εv/∆ln(P) which are also tabulated in Table 7.5, range from a low of .0033 psi-1 for

unaltered samples to a high of .0296 psi-1 for the highly softened sample 30/28-3.

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Table 7.5. Volumetric changes related to P′ for loading segments II and III inPierre shale.

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The Shear Strength of Progressively Softening Pierre Shale

The stress-strain plots presented in Appendix E and in the previous Fig. 7.5, exhibit avery significant decrease in shear strength associated with progressive slaking in thePierre shale. Results from tests at 30 psi confining pressure, indicate a 45% reduction inthe peak strength upon rewetting of a samples which have been dried only 4 days, and adrastic 75% reduction in strength for a sample having undergone only one cycle ofdrying for 28 days and rewetting for 3 days. As observed for sample 30/28-3 inFig. 7.5, the peak strength has effectively been reduced to a value very near the valueof the ultimate strength of unaltered samples. It is quite remarkable and significant thatsuch drastic changes can be simply reproduced by a single cycle of drying andrewetting!

In Fig. 7.11, the peak and ultimate values of P', Q', and specific volume, v, are plottedfor unaltered samples. The peak values of stresses and specific volume exhibitsignificant scatter for unaltered samples. In contrast, the same values for samples whichhad previously undergone 4 days of drying are well behaved, as seen in Fig. 7.12, andexhibit a well defined slope of 1.8 and an intercept of 90 psi. These values can berelated to the more familiar Mohr-Coulomb parameters, c' and φ' by the equations

sin φ' = 3 M / ( 6 + M ) , (7.5)

and

c' = [(Q'/M)-P'] tan φ' (7.6)

where M is the slope, ∆Q'/∆P'. These equations give the values of φ' = 43.8o and c' =54 psi. These values will be discussed in more detail in a later section.

The values for all of the tests are plotted in Fig. 7.13. Of particular interest is the factthat the peak value for test 30/28-3 is near the value of the ultimate strength asdiscussed above. The scatter of data in this plot is indicative of the scatter of shearstrength values typically measured in the field for the Pierre shale. With such a scatterof data, the difficulties of establishing strength parameters for design become quite

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Figure 7.11. Peak and ultimate values of P', Q', and v for unaltered samples ofPierre shale.

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Figure 7.12. Peak and ultimate values for P', Q', and v for Pierre shale samplesdried for 4 days and wet for 1 day.

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Figure 7.13. Final and ultimate values for P', Q', and v for all Pierre samples,unaltered and softened.

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apparent. It is important to be able to relate the variation of strength values observed inFig. 7.13, to some parameter which might be more easily measured and predicted in thefield.

The Effect of Sample Orientation On the Strength of Pierre Shale

All of the results discussed thus far have concerned tests on samples which were allcored such that the plane of primary fissuration was oriented perpendicular to the coreaxis. However, as discussed previously, two samples, designated as "high angle", or"HA", were cored such that the main fissure direction was oriented at an angle of 57o

to 61o from the axis. In Figs. 7.14 and 7.15, the stress-strain curves for these tests arecompared with the curves from samples with a similar slaking history, but differentorientation. The curves for the unaltered Pierre shale in Fig. 7.14, indicate that the"high angle" sample exhibits a higher stiffness, but fails at a 40% lower stress. Thislower strength results from the orientation of the main fissures, being more coincidentto the direction of maximum shear. Thus, failure probably occurs entirely along this lowstrength fissure.

The curve of the volumetric strain for the "high angle" sample is identical to that for theother sample, except that failure is initiatiated at an earlier stage. The fact that failure ofthe "high angle" sample occurs at the exact location of the break of the εv: εa curve for

the other sample, may indicate that the component of strength responsible for segmentIII curves in other samples, is lacking or very weak for samples fissured along themaximum shear direction.

Thus, for the "high angle" sample, the value of Young’s modulus is significantly higherthan samples cut perpendicular to the main fissure direction. However, the values forthe Poissons ratio are identical. This data supports the implication of the drying data,that Young’s modulus for Pierre shale varies with the orientation of the sample.

The loading curves in Fig. 7.15 for softened Pierre shale exhibit similar results. The"high angle" sample, HA30/4-1, is again stiffer than the sample, 30/4-1A, coredperpendicular to the major plane of fissuration, while the initial Poissons ratio is againequal for both samples. Similarly, the strength of sample, HA30/4-1, is 50% lower

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Figure 7.14. Comparison of loading curves for unaltered Pierre shale samplescored perpendicular (30/H-1) and at a highly oblique angle (HA30/H-1) to

major plane of fissuration

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Figure 7.15. Comparison of loading curves for softened Pierre shale samplescored perpendicular (30/4-1) and at a highly oblique angle (HA30/4-1) to

major plane of fissuration

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than that of sample, 30/4-1A, which has undergone similar slaking history, but isoriented differently. The data indicate, however, that the reduction in shear strength,after a four day drying and wetting cycle, is higher in "high-angle" samples. In otherwords, the strength of 30/4-1A is 60% of the unaltered sample, 30/H-1, while thestrength of sample, HA30/4-1, is only 43% of the unaltered, "high-angle" sample,HA30/H-1.

Summary of the Test Results

The drying of Pierre shale results in significant shrinkage. Initially during the dryingprocess, fissures are either produced or already existing microfissures are opened. Withfurther drying, these fissures begin to close tightly. Upon drying, the Pierre shaleexhibits a strong preferred orientation for fissuration. The fissure pattern is transverselyisotropic, with one well-defined plane of continuous fissures, crossed at right angles byone or more sets of rougher, more irregular, and less continuous fissures.

The anisotropic behavior of Pierre shale is further exemplified by the directionalvariation of shrinkage and swelling strains. For samples cored perpendicular to themajor plane of fissuration, the shrinkage strain in the vertical direction is 2 to 3 timesgreater than the radial strain. Upon rewetting, the vertical swell is slightly less thantwice the radial swell. These results indicate that the values for Young’s modulus, andpossibly the Poissons ratio, vary with orientation for the Pierre shale.

The drying process is accompanied by a significant decrease in the void ratio from aninitial value of .35 to about .24 after 4 days, and .22 after 28 days. Most of the changein void ratio and water content occur within 4 days drying time. However, the degreeof saturation continued to show a substantial decrease even up to 28 days drying. Thedegree of saturation dropped from about 100% to 50% in 4 days, and to 34% after 28days.

Samples of Pierre shale which have dried longer than about 24 hours, exhibit a strongtendency to swell beyond that necessary to return the material to its original state, evenunder confining pressures of 10 psi. This excessive swell results from the breaking ofbonds during drying-induced slaking, and is probably a major factor in the reduction ofstrength observed in the field. The amount of swell during rewetting is much more

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dependent on the length of time over which drying occurred, than on the time periodfor wetting. The amount of swell resulting from slaking is remarkably high. Swellingstrains of almost 9% were observed for one sample which had been previously dried for28 days. The final void ratio after drying and rewetting changed from the initial value of.35 to .40 for samples which had dried for 4 days, to .51 for the sample which had driedfor 28 days.

Triaxial compression tests were performed on unaltered samples of Pierre shale, as wellas samples which had undergone various histories of drying and rewetting. The stress-strain curves for these tests exhibited a two- to three-segment linear curve up to anabrupt failure. The first segment exhibited low stiffness, and probably reflected theinitial closing of fissures. With further shear along segment II, the loading curveexhibited an abrupt change in the value of Young’s modulus, which was similarlyreflected in the value of Poissons ratio. The loading curves further exhibited a drasticdecrease in Young’s modulus in response to slaking, while Poissons ratio was notsignificantly affected.

After peak failure, unaltered samples of Pierre shale experienced a rapid decrease instrength to an ultimate value as low as 20% of the peak strength. The peak strengths ofunaltered Pierre shale are rather high, with a value of about 750 psi for 30 psi confiningpressures. However, this strength was shown to be drastically reduced by a single cycleof partial drying, followed by rewetting at 10 psi. For the sample which had dried for28 days, the peak strength exhibited an 80% reduction to values very near the ultimatestrength.

The results of all tests showed a significant amount of scatter, even for samples ofunaltered Pierre shale. The scatter was similar to that observed even at single sites inthe field. The results from these tests re-emphasized the need for a model which relatesthese data to some parameter which is easily and inexpensively measured in the field.Without such a model, the assessment of design parameters from such data proves tobe very unreliable, and in many cases impossible.

The critical state model has proved quite successful in relating the strength of normaland overconsolidated clays to the specific volume of the soil. The following chapter

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will evaluate the test results within the concept of critical state soil mechanics, in orderto assess whether such a model can account for the effects of slaking in clay shales.

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CHAPTER VIII

EVALUATION OF TESTS RESULTS WITHINTHE CONCEPT OF CRITICAL STATE MECHANICS

As introduced in the discussions of Chapter V, the effects of slaking on the engineeringbehavior of clay shales might be closely related to the water content, or the specificvolume, of the softened sample. Indeed, closer examination of Fig. 7.13, reproduced asFig. 8.1, suggest that the scatter in the values for strength might be correlated with thesimilar scatter observed for specific volume. In Fig. 8.2, the peak values for specificvolume are plotted against P'. The data, although somewhat sparse, seem to furthersuggest the existence of a relationship between peak strength and the specific volume atfailure. In this chapter, the triaxial compression data is analyzed within the context ofthe critical state model. Recommendations for further studies and for practicalapplication of these results is also included at the end of the chapter.

Assessment of the Effects of Slaking Within the Critical State Concept

As discussed previously, the critical state model is based on well-defined relationshipsbetween the peak and ultimate strengths of clays and their specific volume at failure.This concept has proved extremely useful in evaluating the engineering behavior ofclays which have undergone consolidation and swelling in response to changes in thestress state. It would be equally useful if the effects of the swelling which results fromslaking, could be similarly incorporated into the critical state concept.

For heavily overconsolidated clays and clay shales, the peak strength is constrained bythe Hvorslev surface, which has been defined in previous chapters. It is thereforeassumed that proving the existence of a common Hvorslev surface for unaltered andsoftened clay shale, is a major step toward assessing the ability of the critical stateconcept to account for the effects of slaking in these materials. As discussed inChapter V, the existence of the Hvorslev surface can be proved if the values of P and

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Figure 8.1. Peak and ultimate values for P', Q', and v for all samples of Pierreshale, unaltered and softened (reproduced from Fig. 7.13).

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Q at failure can be successfully normalized by the function, exp(Γ-v)/λ. Thus, if thepeak values for P' and Q' in Fig. 8.1 can be normalized such that they lie along acommon failure surface, this would suggest that the peak strength of the clay shale isdependent only on the final water content, and therefore independent on whether thiswater content resulted from slaking, simple swell, compression, or a combination of allof these processes.

Obtaining the normalization parameters. The normalization of P' and Q' commonlyemploys the parameters Γ and λ . The values of these parameters define the criticalstate line in v:ln P' space, and can best be obtained from triaxial compression tests onnormally consolidated, or lightly overconsolidated samples. Since such test results werenot presently available to the author, it was necessary to obtain these parameters in aslightly modified manner, using triaxial compression data from the heavily-overconsolidated samples tested here.

After peak failure has occurred during shear of an overconsolidated clay, the values forP' and Q' tend to move toward the critical state line in both P':Q' and P':v space.

Figure 8.2. Relationship between the specific volume and ln P’ at peakstrength for Pierre shale samples at confining pressures of 10, 30, 50 psi.

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However, the amount of shear during triaxial compression tests on highlyoverconsolidated clays is generally not sufficient for the material to reach the residualstrength defined by the critical state. In addition, because failure in these materials isoften concentrated along a narrow shear zone, the ultimate values obtained for P', Q',and especially v, may not necessarily define the true critical state line, even afterextensive shear displacement. Still, the values obtained at the end of these tests, mayallow the calculation of equivalent parameters necessary to normalize the stresses, P'and Q'.

As shown in Fig. 8.1, the P' and Q' values measured at the end of the tests appear to beprogressing toward a common straight line in P':Q' space. The approximate slope of theline along the lower bound equals 1.4. Using Eqn. 7.5, these results indicate an ultimatefriction angle of about 34o.

However, the final values for v and ln(P') shown in Fig. 8.3, do not all appear to beprogressing toward a common line. As discussed previously, this is primarily areflection of the uneven distribution of water content within the sample, and the factthat the value of specific volume measured indicates an average value, rather than thevalue along the failure plane. If the actual specific volume along the failure planes couldbe measured instead of the average values, then the data would probably indicate acommon critical state line in v:ln P' space. Indeed, the scattering of final specificvolumes is minimized for samples whose values of specific volume at the beginning ofthe tests were nearly equal.

For example, the data for unaltered samples in Fig. 8.4, define a critical state line withslope of -0.020, and an intercept value, vo, value of 1.455. Thus, for the purpose of

normalization, these values will taken as equivalent values for Γ and λ, respectively.However, because of the difficulties in measuring the actual specific volume along thefailure zone, and to avoid possible confusion, the designation, vo, is used here to

replace the more tightly defined parameter, Γ, but is similarly defined as the value ofthe final specific volume at ln(P) = 0. Similarly, while the slope of the line in Fig. 8.4may be functionally equivalent to λ, it may or may not be the slope of the "true"

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Figure 8.4. Values of v and ln P’ at the ultimate (final) strength of unalteredPierre shale samples tested at 10, 30, and 50 psi.

Figure 8.3 Values of v and ln P’ at the ultimate (final) strength of all Pierreshale samples tested at 10, 30, and 50 psi.

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critical state line. Therefore, the slope of this line will be designated as capital lambda, Λ, instead of the more conventional, λ.

The equation for the Hvorslev equation has been previously derived as

Q = (M-h) exp[(Γ-v)/λ] + hP , (8.1)

or by normalizing P and Q, as

Q/exp[(Γ-v)/λ] = (M-h) + hP/exp[( Γv)/λ] . (8.2)

Thus, if the peak values of P and Q are normalized using the exp[(Γ-v)/λ], the resultingvalues should define a straight line along which all peak strength of a heavilyoverconsolidated should lie. By replacing Γ and λ with vo and Λ, Eqn. 8.2 becomes

Q/exp[(vo-v)/Λ] = b + m P/exp[(vo-v)/Λ] . (8.3)

Thus, the peak values of P and Q, which were measured for the unaltered and softenedsamples of Pierre shale, can be normalized by the exp[(vo-v)/Λ]. If the resulting values

define a common failure surface for both unaltered and softened Pierre shale, then thecritical state concept appears to account for the effects of drying-induced slaking in thisclay shale.

Normalization of data. The results of the normalization, using the values of vo =

1.455 and Λ = 0.020, are listed in Table 8.1 and plotted in Fig. 8.5. As demonstrated bythis plot, the normalized data strongly define a single surface for the peak strengths ofunaltered samples of Pierre shale, as well as for samples which have undergone highlyvarying degrees of slaking. Considering the extreme scatter and drastic differences instrength shown in the previous Fig. 8.1, the degree of fit of the normalized data is quiteremarkable.

The plot in Fig. 8.5 does not include results from the extremely softened sample 30/28-3, since its normalized strength is more than 50 times greater than the highest value inthe plot. In order to observe the results from this test, the normalized data have beenreplotted on a log-log scale in Fig. 8.6. Even with such extreme softening,

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the normalized strengths for sample 30/28-3 lie along the previously defined line. Theseresults strongly support the hypothesis that the effects of slaking can be consideredwithin the established critical state concept.

A least-squares fit of the normalized data for unaltered samples indicates a correlationcoefficient, r2, of 0.995 and the parameters

m1 = 2.3

Figure 8.5. Normalized peak values of P' and Q' for all samples of Pierre shale,unaltered and softened (excluding 30/28-3), showing well defined failure surface.

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Table 8.1. P’ and Q’ values for Pierre shale samples, along with normalizedvalues for P’ and Q’ using vο = 1.455 and Λ=0.020.

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B1 = 0.2 psi,

where b and m are defined in Eqn. 8.3. As demonstrated in Fig. 8.5, the normalizeddata for the moderately softened samples (i.e. xx/4-1, xx/4-3) also lie along this line.The three more intensely softened samples, however, lie along a slightly shallower line,having a correlation coefficient of 1.000, and defined by the parameters

m2 = 2.0

B2 = 0.9 psi.

Figure 8.6. Normalized peak values of P' and Q' on log-log scale for allsamples of Pierre shale, unaltered and softened (including 30/28-3), showing

extension of failure surface.

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It is significant that the segment of the failure surface for unaltered samples alsocontains data points for samples which have undergone slaking. This suggest that theprocess of slaking merely acts to altered the water content of the sample within theconfines of the critical state model, and does not alter the model itself. If the break inthe failure surface occurred between the unaltered and softened samples, such aconclusion might have been suspect.

The parameter m can be related to the Hvorslev parameters, φe and ce, which are

defined as the effective cohesion and internal friction angle for samples at the samespecific volume. As with Eqns. 7.5 and 7.6,

sin φe = 3m / (6+m) , (8.4)

and

ce = [(Q/m) - P'] tan φe . (8.5)

From Eqn. 8.4, the value for (φe)1 is calculated as 55o and (φe)2 as 48o. The value ofthe effective cohesion, ce, varies with the value of Qo and thus depends on the specificvolume of the sample. Using the test data for σ3 = 30 psi, values of ce for the Pierre

shale have been calculated. Table 8.2 tabulates these values for the Hvorslevparameters, as well as for the more conventional Mohr-Coulomb parameters.

These results again illustrate the drastic reduction in the cohesion that can occur withthe softening of the Pierre shale. The major reduction of strength during these tests,results from a drastic decrease in the cohesion from 123 psi to 0 psi. However, insamples which have dried for longer than 4 days, the strength is further reduced by adecrease in the friction angle. These reductions in shear strength will be discussed inmore detail in the following chapter.

In addition to the normalization of P' and Q' at peak failure, the values for P' and Q' canbe normalized for the entire loading path during testing. Fig. 8.7 shows the normalizedstress paths from the beginning of shear until peak failure. These paths are similar inshape for all samples, and are characterized by an initial steep rise in

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Q'/exp[(vo-v)/Λ], followed by a gradual decrease in P'/exp[(vo-v)/Λ] up to a distinct

peak. These peaks probably define a prefailure yield surface for the Pierre shale. Afterthis peak, both of the normalized stresses decrease rapidly until they intercept theHvorslev surface where peak failure occurs. Interestingly, many of these paths appearto briefly follow the Hvorslev surface before experiencing total loss of shear strengthBecause of the localization of dilation within the failure zone, the post-peak paths arehighly unreliable and inconsistent, and have not been plotted.

Normalized strengths for "high-angle" tests. Normalization of the peak stresses forthe "high angle" samples gave the values

P/exp[(ν0 -v)/ ] Q/exp[(ν0 -v)/Λ]

HA30/H-1 0.5 1.3 HA30/4-1 4.6 9.2 .

Table 8.2. Cohesion and internal friction angle values for unaltered Pierreshale, as well as for samples which have been rewetted after various periods of

drying.

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Figure 8.7. Normalized pre-failure loading paths for all samples of Pierreshale, unaltered and softened (excluding 30/28-3).

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Comparison of these results with the failure surface for the samples in Fig. 8.5,indicates that the strength of the unaltered "high angle" sample lies along the same lineas the previous data, but the softened "high angle" sample lies somewhat below.Assuming that these samples are representative, and that their positions define a failuresurface for "high angle" samples, then this failure surface has an intercept, B = 0.3 psiand a slope, m = 2.0, which is equal to the slope, m2, for obtained for highly softened

samples in the previous tests. These limited results imply an effective Hvorslev frictionangle of 48o and an effective cohesion of 3.1 psi. The significance of these "high angle"test will be discussed in more detail in the following section.

Discussion on Slaking and the Shape of Failure Surfaces for AnisotropicMaterials

The previous section demonstrated that the critical state concept can successfullyaccount for the often drastic effects of slaking in the Pierre shale. It is important tostress, however, that the present data is insufficient to conclusively relate these resultsto those that might be obtained on remolded and artificially consolidated samples ofPierre shale material.

The effects of fissuration and anisotropy on the shape of failure surfaces. Thesamples tested in this program were initially highly overconsolidated, anisotropic, andfissured clay shales. As noted before, the rather high values obtained for the effectivefriction angle are more indicative of rock than soil. The materials typically consideredwithin critical state models are homogeneous and isotropic. In such materials, thefailure occurs along a simple plane of maximum shear stress, whose orientation isdetermined only by the surrounding stress state. The strength of the material istherefore independent of sample orientation.

However, in fissured materials, or in any anisotropic materials containing directionalelements of weakness, two additional factors may influence the measured shearstrength. First, the plane of failure may be controlled by the position and orientation offissures or other weaknesses, and might therefore not coincide with the plane ofmaximum shear strength. Secondly, for the same reason, the failure plane may not besimple and smooth, but may be rough or "sawtooth", similar to the joint model ofPatton discussed in Appendix C. Both of these factors tend to add a extra component

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of strength, depending on the orientation of the sample, the magnitude of the stressstate, and the relative strength of the "intact" material.

It is quite informative to compare the simple model of Patton with the results that thepresent author has obtained for the Pierre shale. The model of Patton is graphicallyillustrated in the strength envelopes of Fig. 8.8a, while the normalized failure surfacefor the Pierre shale is presented for comparison in Fig. 8.8b. In the Patton model, thestrength of a simple joint (Curve B) is dependent only on the frictional resistance alongthe walls, and is given by φ. However, for a sawtooth joint, the resistance to shear(Curve A) has an additional component, i, equal to the angle of the tooth, and resultingfrom the sliding of one tooth over the other (i.e. dilation). At low stresses, this addedcomponent results in a steeper failure envelope, with an total friction angle of (φ+ i). Athigher stresses, however, the shear stress exceeds the strength of the teeth (i.e.asperities), and shear displacement is accomplished by breaking through the teeth ratherthan sliding over them. At this point, the friction angle decreases to φ, the value of thefriction angle for the smooth joint.

The normalized shear strength data for the Pierre shale (Curve C) show a similar two-segment failure envelope, suggesting the influence of fissuration. For the samples coredperpendicular to the major plane of fissuration, the failure plane is probably step-like, asillustrated in Fig. 8.8b. As in the Patton model, these irregularities in the failure planeadd an extra component of shear resistance at lower stress levels, resulting in a steepfailure envelope, and therefore a higher friction angle. At higher stresses, this frictionangle is reduced as shearing occurs through the irregularities, rather than over them.

This model is further supported by the data for the "high angle" tests illustrated byCurve D. When the orientation of the major fissures are coincident with the plane ofmaximum shear stress, as is almost the case for the "high-angle" tests, then the failureplane is essentially equivalent to the planar joint in the Patton model. As would beexpected if such were the case, the limited data from the "high angle" tests suggest afailure envelope which is void of a steeper portion and is parallel to the upper portionof Curve C.

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Of greater interest to the present study, however, is the fact that the failure envelopefor the Pierre shale in Fig. 8.8b, is normalized using a function of the specific volume.Therefore, an increase in the normalized stresses can also result from an increase in the

Figure 8.8. Schematic comparing (a) the Patton model for a sawtooth joint, to(b) normalized failure envelopes for Pierre shale.

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specific volume (i.e. water content), even if the actual effective stresses are constant ordecreased. In effect, an increase in the specific volume indicates an increase in thedegree of softening within the material. Therefore, with regard to the Patton model, anincrease in the stress state or an increase in the water content, favors the tendency forshear through irregularities, rather than over them.

The results from the triaxial tests on Pierre shale, therefore support the fissuredeterioration model that the author proposed in Chapter IV. They further indicate thatthe influence from fissures is minimized by the either increasing the stress state orincreasing the water content of the clay shale.

Similar two-segment boundary surfaces have been reported in two recent papers fornatural heavily overconsolidated clays in Canada (Graham and Li, 1985; Graham andAu, 1985). In Fig. 8.9, the strength values were normalized using the preconsolidation

Figure 8.9. Normalized strength envelope showing low stress strengths,overconsolidated strengths, and critical state strengths for heavily-

overconsolidated clays in Canada (Graham and Li, 1985).

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pressure, σvc, defined as the maximum pressure to which the soil has been subjected to

in the past or present. Although the actual values of the normalized strengths aredifferent, this is equivalent to the normalization procedure used in the present paper.The normalized data for natural samples defines a two-segment boundary surface, A-B-C, for overconsolidated samples. As predicted above, the initial steep failure surfacechanges abruptly to a more shallow surface, and at its extreme, curves over to intersectthe critical state line. At lower stresses, or lower specific volume, the material strengthis greatly influenced by the presence of fissuration, and is defined by the steepersurface. With higher stresses, or at higher specific volumes, the influence of thesefissures is greatly reduced, and the strengths can be defined by the flatter slope.

In these two papers, the authors have also investigated the relationship between thestrengths of natural clays, and that of clays which have been artificially remolded andconsolidated in the laboratory. As is evident in Fig. 8.9, if the strengths of theartificially remolded samples are normalized by their preconsolidation pressures, thenthe resulting values lie well above the failure envelope for the natural materials.

Figure 8.10. Normal consolidation and critical state lines in P':v space fornatural and remolded clays of Fig. 8.9 (Graham and Li, 1985).

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However, as shown in Fig. 8.10, the normal consolidation lines for the natural andartificially remolded samples are essentially parallel, but are shifted in position. This hasbeen attributed to aging or cementation. Interestingly, if the artificially remoldedsamples are normalized in reference to their own consolidation lines, then thenormalized strength lie on the same failure surface as the natural samples (Fig. 8.11).This should be considered when attempting to compare results from natural andartificially-remolded materials.

Slaking and the intensity of natural remolding. The test results presented in ChapterVII, indicate that a 75% decrease in strength can occur in Pierre shale after only asingle cycle of drying and rewetting. In the present chapter, this has been interpreted asa reduction of 6o in the internal angle of friction, and a very significant loss of cohesion

Figure 8.11. Failure envelope for natural and remolded samples, with datanormalized according to respective normal consolidation lines (Graham and Li,

1985).

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from 123 psi to zero. Although the reductions in shear strength experienced in thesesamples are rather drastic, the calculated friction angles for the softened clay shale arestill more indicative of rock-like materials than of soft soil. Even the ultimate strengthof these samples is defined by a rather high friction angle of about 34o.

Although these angles are similar to limited strength data for firm Pierre shale at OaheDam (Fleming et al, 1970), they are still much higher than those obtained for highlyweathered Pierre shale in the field. This indicates that the degree of natural remoldingexperienced in the field must be much more intense than that which has beenaccomplished with the samples in this test program. Indeed, the softening, which wasproduced in the laboratory by a single cycle of drying and wetting, was probablyprimarily confined to zones along the vertical and horizontal fissures. Therefore, thehigher values for friction angles observed in these tests probably reflect the influence ofremaining zones of stiff, intact material, and therefore the continued influence of fissureorientation and roughness.

In summary, more intense remolding is required before the influence of fissures istotally destroyed in the Pierre shale. Furthermore, the intensity of remolding in the fieldis apparently greater than has been observed in these test. This does not suggest thatthe processes other than slaking are required to achieve this intensely remolded state.Indeed, more intense natural remolding might be achieved in the Pierre shale, and otherclay shales as well, by (a) increasing the number of cycles of drying and wetting, (b)increasing the amount of drying for each cycle, (c) allowing more time for the claystructure to adjust itself to the new state of increased water content, or (d) slaking thematerial in the presence of shear stresses.

Since the sample 30/28-3 has already reached a state of zero cohesion, with moreintense slaking and softening of the clay shale, we should expect to begin to see afurther reduction in the internal friction angle. This would be reflected by either agradual bending over of the failure envelope, or an abrupt break in its slope to values,more indicative of soft clay soil.

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Figure 8.12. Complex P':v history followed by Pierre shale sample 30/28-3,prior to and during triaxial testing.

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Slaking and the Critical State Concept: Recommendations for FurtherStudies and Practical Application

The test results presented in this dissertation are very encouraging. Considering thecomplex history of these clay shale samples illustrated in Fig. 8.12, it is significant anduseful that the resulting peak strengths are dependent only on the final states of stressand specific volume, and not on the path itself.

Still, these findings are preliminary, and general conclusions based on these results arelimited. More extensive investigations are required before the critical state conceptproves applicable to all clay shales undergoing slaking under various conditions. In thefollowing sections, the author presents suggestions regarding the potential applicationof these results to practical problems in the field, and discusses recommendations forfuture studies, both in the laboratory and in the field.

Practical application of the Critical State Concept in Clay Shales. The literaturereviews and discussions in Chapters II and III, attest to the great difficultiesencountered by engineers who attempt to accurately and reliably assess the engineeringbehavior of clay shales in the field. Two major complications are recognized: (a) theextreme variability in the strength and stress-strain parameters of a clay shale, evenwithin very localized areas, and (b) the drastic changes in the engineering behaviorwhich clay shales can undergo within geologically-short time periods. The authorbelieves, and has supported the belief in this dissertation, that in many clay shales, suchcomplications result primarily from the swelling associated with slaking.

In order to account for the softening effects of slaking, the engineer must either be ableto associate the slaking history to the final engineering parameters of the clay shale, orhe must be able to measure these final parameters directly. Unfortunately, the processof slaking in a clay shale deposit can be extremely variable and complex, depending onsuch factors as the local topography and stratigraphic layering, the frequency andorientation of jointing, and the local climatic and hydrologic environment, including theamount of rainfall, the length and frequency of wetting and drying cycles, theinfiltration and evaporation characteristics of the overlying soil, and the flowcharacteristics and chemical composition of the ground water, for example. Clearly,

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reliable assessment of the slaking history, or slaking future, of any soil is presentlyimpractical.

Therefore, in order to assess the present engineering behavior of a clay shale, or topredict future changes in this behavior, we must obtain useful strength and stress-strainparameters by direct measurement, or indirectly using some other model. For practicalfield applications, direct measurement of strength parameters through laboratory testingof clay shale samples has proved to be costly, inadequate, and unreliable. The extremespatial and temporal variability of clay shale strength requires a very extensive testingprogram if the strength and stress-strain behavior of the clay shale deposit is to beadequately determined. Many of the past engineering difficulties with clay shaledeposits can probably be attributed to inadequate sampling programs.

Because of high expense and limited range, in-situ testing is generally impractical formost engineering projects in clay shales. Even with laboratory testing, the cost ofcoring, retrieving, and testing of core samples of clay shale generally prohibits asampling program which is extensive enough to adequately assess the strength of theclay shale deposit. This becomes particularly evident when one considers the need forcontinuous monitoring of the material strength with time.

However, test results presented in this dissertation suggest that the effects of slakingwithin a clay shale are restrained by the critical state model. Therefore, once the criticalstate model has been defined, the strength of a clay shale which has undergone slakingcan be determined if one knows the water content and the surrounding stress state. If,as suggested, the strength of a clay shale can be indirectly determined from its watercontent, then mapping and monitoring of the strength within a clay shale deposit mightbe accomplished, adequately and inexpensively.

In contrast to the testing of strength, the sample retrieval and laboratory proceduresassociated with the determination of water content are uncomplicated and inexpensive.Because of such low expense, literally thousands of samples could feasibly be retrievedand their water contents determined. This could allow very extensive mapping of clayshale strength, both laterally and with depth, as well as provide an inexpensive means ofmonitoring strength changes with time. In addition, the mapping and monitoring ofstrength variations might be further enhanced by the use of electrical resistivity, self

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potential, or seismic measurements, which can provide indirect means of assessingvariations in water content. These methods are discussed in detail by Scully (1973), aswell as in several geology texts.

Thus, the major advantage of using the critical state model to constrain the effects ofslaking, is that the model provides us with a parameter (i.e. water content) which caneasily and cheaply be measured at a large number of locations within a clay shaledeposit. In addition, changes in strength with time can more cheaply be monitoredusing water content measurements, rather than direct measurements of materialstrength.

This relationship between water content and strength in a slaking clay shale does not atthe present allow for the prediction of strength changes which might occur in the clayshale deposit with time. However, such this relationship to water content may facilitatefuture studies attempting to relate environmental conditions to the rate of slaking in aclay shale deposit. This is a result of two factors: (1) the ease with which watercontents can be monitored with time could result in more field data relating watercontent changes to environmental parameters, and (2) laboratory studies which relatewater content to various slaking processes should be much easier to perform than thosewhich require shear strength measurements after slaking.

As envisioned by the author, the analysis of a project site within a clay shale depositmight proceed as follows:

(a) preliminary electrical resistivity, self-potential, and seismic refraction studiescarried out at the surface, in order to detect possible stratigraphic and hydrologicboundaries, as well as potential zones of abnormally high water contents;

(b) an extensive program of coring to obtain samples to be used for water contentmeasurements; in addition, these samples should be used to recognize stratigraphicboundaries by determining mineralogy, grain size, etc.; retrieved samples large enoughto perform shear strength tests should be set aside; however, an infinite number ofdeterminations of water content should be possible along each core since the integrityof the sample is not as critical for these measurements as it is for shear strength testing;

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furthermore, the shafts produced by this coring should be employed for furtherelectrical resistivity logging; such studies would help recognize potential failure planes;(c) if possible, measurement of the natural stress state within the deposit, usingrelaxation determinations on core samples or other more sophisticated techniques;though such measurements may not be necessary, it should be kept in mind thatexcessive horizontal stresses are common in slaking and swelling materials, such as clayshales; these stress state determinations would be important for input of initial stressconditions in any later finite element analysis;

(d) determination of the normalized failure envelope, using intact samples retrievedfrom the coring program, and possibly remolded samples; if the material behavior canbe constrained by a critical state model, then a large number of intact samples isprobably not necessary since scatter of the data should be minimized by thenormalization process; if more than one potentially unstable stratigraphic layer exist,failure envelopes should be determined for each layer; the procedure for obtaining thefailure envelope would be similar to that used in this dissertation:

(i) determine the normal consolidation line coefficients from consolidationtests, if possible, or determine the coefficients of the critical state line fromintact samples, as in this dissertation (or from remolded samples, if this provessuccessful); these various methods will be discussed in more detail in thefollowing section;

(ii) using these parameters, normalize the shear strength data from triaxialtests on intact samples, on samples which have undergone induced slaking, as inthis dissertation, and possibly on artificially remolded samples;

(iii) plot normalized data and calculate coefficients for the failureenvelope;

(e) once the normalized failure surface is defined, finite element analyses can beperformed on the deposit by assuming constant volume; thus the water content whichhas been determined for each point based on the actual measurements, is assumed toremain constant during shear loading, and is used to normalize the stress state at thatpoint.

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Although such analyses are extremely useful for predicting the material stabilityfollowing changes in the stress state, it should remembered that such changes in thestress state may lead to increased slaking in clay shales ; thus, it is important tocontinue to monitor water contents at critical locations during and after construction;the changes in water content, if any, can then be used in updated finite element analysesto recognize any problems which may be developing. With experience and furtherexperimentation, it may become possible to predict the increased slaking which mightoccur in response to changes in stress states. With such information, iterativecalculations using finite element analyses will allow the long-term stability analysis ofclay shale slopes.

Before applying the procedures described above to actual field conditions, much morepreliminary research is necessary. The intent of this discussion has been to provide ahopeful summary of the potential application of the critical state model to the analysisof a clay shale deposits which have undergone, or are undergoing, softening by slaking.The results presented in this dissertation are, at most, only suggestive of therelationship between the slaking process and the critical state model, and then only forthe Pierre shale. More research is required before such relationships can be establishedfor all clay shales under various conditions. Even then, further research must beconducted before these results are successfully applied to practical problems in thefield. In the following section, the author presents some suggestions for furtherresearch with regard to the effects of slaking, and the application of the critical statemodel.

Recommendations for further studies. This dissertation has presented significantpreliminary results which suggest that the effects of the slaking process can, at leastunder certain conditions, be constrained by the critical state model. However, furtherresearch is needed before these results can be validated and applied to practicalproblems in deposits of the Pierre shale and other clay shales.

As with most naturally-occurring geological materials, clay shale deposits can bestructurally complex. Therefore, the testing and analysis of clay shale deposits entail thesame complications encountered when working with other geological materials,including the scale effects, inhomogeneity, and anisotropy which result from the

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presence of joints and the variation of mineralogy and fabric. While these complicationsare certainly a concern in any geotechnical analyses, they do not invalidate the potentialapplication of the critical state model to practical problems in a progressively slakingclay shale deposit. In fact, the application of the critical state concept to slake-susceptible clay shale deposits, alleviates the complications of inhomogeneity andanisotropy resulting from water content variation. Still, scale effects, inhomogeneity,and anisotropy affect geotechnical analysis in all geological materials to some extent,and should therefore continue to be investigated.

The problems inherent to the present investigation of clay shales can be divided intotwo major groups: (1) defining the normalized failure envelope, and (2) understandingthe process and effects of slaking. Future studies involving the definition of the failureenvelope for slake-susceptible clay shales should proceed along four paths:

(a) determination of the normalization parameters: In critical state soil mechanics,the parameters used to normalize the shear strength envelope are typically determinedfrom the position and slope of either the normal consolidation line or the critical stateline in v:ln P' space. The simplest method for determining these parameters useconsolidation tests or triaxial tests on normally-consolidated, remolded samples. Initialstudies should employ such procedures to determine if parameters derived fromremolded clay shale material can be used successfully to normalize data from naturalclay shale deposits. Using remolded samples to determine the normalization parameterswould not only be the simplest method due to the ease of retrieving and preparingremolded samples, but the use of a standard method would also allow for bettercomparison of results between different clay shale deposits. However, as discussedpreviously, data from Graham and Li (1985), indicates that the normal consolidationlines for artificially remolded and natural clay deposits often differ due to aging effects.In addition, it is questionable whether the normal consolidation line remains linear atthe high preconsolidation stresses often associated with clay shale deposits. Initialstudies need to determine what affects, if any, these factors will have on thenormalization process.

If remolded samples cannot be used to determine the normalization parameters, thenthese parameters must be derived from the normal consolidation line or critical stateline of natural samples. Of the two lines, the normal consolidation line should in general

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be more easily determined. Relative to strength tests, consolidation tests are fairlydirect and require fewer intact samples. If possible, consolidation testing of naturalsamples is probably the best method for obtaining the necessary parameters fornormalization. However, due to the high preconsolidation pressures associated withmany clay shales, the pressures applied during standard consolidation tests areinsufficient to return the clay shale to a normally consolidated stress state, therebypreventing the determination of the normalization parameters. In such cases, a possiblealternative would be to extensively slake the clay shale under low confining pressuresprior to consolidation testing.

As a last resort, parameters which would allow normalization could be derived usingthe ultimate strengths of intact clay shale samples. This is the method used in thepresent dissertation, and which results in defining a "pseudo-critical" state line. Thecritical state line from which the parameters in this dissertation were determined definesa true material state of the Pierre shale, and the use of the term "pseudo" is not meantto imply otherwise. However, because failure in highly overconsolidated clays occursalong narrow failure zones, the water content within the failure zone is generally muchdifferent from the average total water content measured experimentally.

Therefore, the parameters derived from the ultimate strengths of overconsolidatedsamples are likely to vary significantly from those derived from the "true" critical stateline of normally consolidated samples. This results in two complications which areabsent in normalization methods which employ remolded samples: (i) while there isonly one "true" critical state line, there can be many "pseudo-critical" state lines,resulting in nonstandard results which cannot be easily compared between deposits, and(ii) a "pseudo- critical" state line must be determined for samples having the same initialstress-void ratio state prior to testing. Therefore, the samples used for determining thenormalization parameters must have undergone the same degree of slaking andcompression. In the present dissertation, unaltered samples were used, since they hadsimilar values for specific volume and were retrieved from the same horizon. However,in some instances, it may be difficult to retrieve a sufficient number of samples whichhave undergone the same degree of slaking.

If the critical state method is to be of practical use in clay shales, it is essential thatmethods be developed for successfully normalizing the failure envelope. Thus, it is

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suggested that future research concentrate initially on developing and testing thenormalization methods discussed above. Such research should preceed or be carriedout concurrently with the research discussed below.

(b) extending the range of the failure envelope: Although the results presented inthis dissertation indicate a very drastic reduction of strength with a single wetting anddrying cycle, the data also indicated that the strength of even the most softened samplewas still higher than the fully softened strength of the Pierre shale. Within the limitedrange of softening induced in this study, a two-segment failure envelope was exhibited,apparently resulting from the effects of jointing.

It is important to derive the failure envelope for the entire range of softening, fromintact to fully softened conditions. How does the envelope change before interceptingthe critical state line? At what point are the effects of jointing and aging erased? Theanswers to these questions could come from studies on the Pierre shale or other clayshales, which are extensively softened, either naturally or by induced slaking. Asdiscussed previously, more extensive softening and natural remolding might beproduced in the laboratory or field by increasing the number of wetting and dryingcycles, by increasing the amount of drying during each cycle, by allowing more time forthe clay structure to adjust itself after rewetting, and by slaking in the presence of shearstresses. Some success might also be achieved in the laboratory by physically remoldingthe material, although the results of this method should be checked closely againstresults of extensive softening of natural samples.

(c) extending the method to other clay shales: The results presented in these initialstudies of the Pierre shale suggest that the changes in the strength which result fromslaking, are constrained by the critical state model. More studies are needed to confirmthese findings for the Pierre shale, and to extend the range of the failure envelope of thePierre shale. As discussed previously, the Pierre shale is an excellent material for suchstudies, being highly susceptible to slaking, having a wide range of strengths, beingstructurally simple, and being easily sampled. It is therefore suggested that furtherresearch into the relationship between slaking and the critical state model be initiallyconducted using the Colorado Pierre shale as the model material.

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However, similar investigations should be conducted on other clay shales, in order todetermine the extent to which the critical state model can be applied to all slakingargillaceous materials. Other clay shales, because of greater degrees of isotropy orsimpler joint patterns, may prove to be even better research materials than the Pierreshale. Furthermore, if similar results are obtained for the London clay and the Cucarchaclay shales, then the historical records associated with these materials may provideinvaluable information regarding the effects of slaking with time.

(d) effects of anisotropy and jointing. Theoretical and experimental evidencepresented in this dissertation indicate that anisotropy and jointing can have not only asignificant effect on the initial strength of a clay shales, but may also greatly affect theamount of strength lost during the process of slaking. More extensive experimentalstudies should be carried out on the Pierre shale and other clay shale, in order todetermine complete failure envelopes for a number of sample orientations. Thesestudies would provide important information on the effects that material anisotropymight have on the shape of the failure envelope, and therefore on the rate of strengthloss with increased water content. With a series of failure envelopes for variousorientations, the normalized strength of a clay shale sample could be determined orinterpolated regardless of the orientation of the material anisotropy relative to thepotential failure plane.

In addition to the experimental testing of samples at various orientations, more direct,systematic investigations on the effects of softening along single and multiple jointswould be quite useful. Such investigations might entail experimental testing, as well ascomputer modeling. Questions to be answered include: How much softening is neededto remove the effect of roughness along a joint? How does softening progress into ajoint wall; under what conditions is there a distinct boundary between softened andintact material, and under which is this transition more gradual? What effect do thesedifferent boundaries have on altering the strength of the joint? How is the shape of thefailure zone altered during the softening of various multiple joint systems? How doesthis affect the overall strength of the material? How well can the Patton model forjointed materials be applied to a progressively softening, jointed clay shale? At whatlevels of stress and softening are the effect of joints insignificant?

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In order to apply the critical state model to a progressively slaking clay shale deposit,we must be able to define a normalized failure envelope which accounts for changes inthe water content (or specific volume) occurring during the slaking process. Thedefinition of such a failure envelope requires that suitable parameters be obtained fromthe normal consolidation or critical state lines. It is therefore important that futureinvestigations develop simple and repeatable methods for determining the parametersneeded for successful normalization of clay shale strengths. In addition, the validity ofthe results must be tested over the entire range of the failure envelope and undervarious conditions, including slaking under in the presence of shear stresses, as well asslaking under different confining pressures, temperatures, and wetting and dryingcycles. Additional failure envelopes should be determined and compared for variousorientations and for several different clay shale materials.

In addition to the investigations discussed above regarding the failure envelope, furtherinvestigations into the actual process of slaking are needed in order to predict the rateof softening in a clay shale which is susceptible to slaking. Simple laboratory tests couldprovide valuable information regarding the changes in water content resulting fromslaking under various factors, including for example, changes in the frequency andlength of wetting and drying cycles, changes in the confining pressures and shear stressstates, changes in the chemical composition of the slaking fluid, and changes intemperature. These laboratory studies could be complimented with studies in the fieldwhich attempt to relate lateral and temporal variations in the water content to similarvariations in rainfall, temperature, ground water levels, and ground water chemistry, aswell as to surface and stratigraphic topography.

With extensive research into the nature of the slaking process and its relationship to thecritical state model, it may become possible to not only accurately determine the lateralvariation of shear strength in a given clay shale deposit, but to predict the changes inthe strength over the lifetime of an engineering structure, as well.

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CHAPTER IX

CONCLUSIONS

Clay shale deposits have been a source of much confusion and trouble for engineerswho must deal with these materials. Our present understanding of the engineeringbehavior of clay shales is primitive relative to the current state-of-the-art for othergeological materials. The author feels that most of our past inadequacies in dealing withthese materials have resulted from our lack of appreciation of their transitional nature.This has resulted in inadequate and confusing classification schemes, as well asdangerous and costly slope and foundation designs.

The transitional nature of clay shales is both temporal and physical. Physically, clayshales are transitional between rock and soil, and therefore exhibit properties of both.This has been a source of problem for geotechnical engineers, who traditionally viewgeological materials in terms of rock mechanics or soil mechanics, but rarely in terms ofboth. In addition, clay shales are transitional in time, and tend to transgress from rock-like behavior to soil-like behavior within a relatively short time period. Such rapidchanges in material properties create challenges in classification and in engineeringdesign, both of which are traditionally based on material properties as they exist at thepresent and not on possible future properties.

The engineering problems related to clay shale deposits have been worldwide andinvolve both natural and man-made structures. These have been discussed in somedetail within this paper using examples associated with four very different clay shalegroups: (1) the British clay shales, (2) the vast clay shale deposits of the north-centralplains of North America, (3) the Cucaracha clay shale in the Panama Canal zone, and(4) the clay shales of southern Italy. These materials range in their initial characteristicsfrom the "simple" stiff fissured clays of England, to the "soapy" clays of Panama, to thescaley, tectonically-sheared clays of Italy, and to the rock-like shales of North America.Still, they all share the common tendency to rapidly lose shear resistance with time.

Engineering designs based on traditional methods for analyzing geological materialshave been dangerously inadequate. From the discussions presented in this paper, two

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conclusions become apparent: (1) any analysis which assumes only soil-like behavior,or only rock-like behavior, will fail to adequately determine the engineering behavior ofclay shales, and (2) any successful analysis must account not only for the materialproperties of the clay shale as can be measured at the present, but must also account forchanges in these properties as the clay shale deteriorates into a more soil-like state.

The evidence presented in this dissertation suggest the importance of fissuredeterioration as a viable model for explaining much of the long-term reductions ofstrength exhibited by clay shales in the field. Furthermore, much discussion in theliterature has alluded to the importance of fissure deterioration in reducing the strengthof clay shale deposits. However, surprising few, if any, theoretical or experimental,investigations regarding fissure deterioration have been reported in the geotechnicalliterature.

Similarly, the susceptibility and importance of slaking within clay shales has often beenrecognized by engineers concerned with clay shale deposits. Yet, the author has foundthe literature essentially void of studies regarding the slaking of materials underconditions of confinement which occur in the field. Similarly, there is a great need forsystematic investigations into the effects of slaking on the strength and stress-strainresponse of clay shales. For these reasons, the author has carried out theoretical andlaboratory investigations regarding fissure deterioration in clay shales, with particularemphasis on the effects of slaking and swelling within fissures.

In Chapter IV, the author presented a simple model for the deterioration of clay shales,and considered the theoretical implications of a jointed material, with progressivesoftening of the fissure walls. The clay shale was depicted as progressing through fourstages of alteration, including (a) a rock-like mass in which the strength was controlledprimarily by the orientation of and strength along the intact fissures, (2) a partially-softened rock-like mass with the strength controlled primarily by the strength andorientation of soft, filled fissures, (3) a highly-softened mass consisting of a matrix ofsoft clay surrounding stiff, intact cores, and finally (d) a fully-softened, remolded clay.Thus, the analysis of a clay shale undergoing progressive softening along fissures iscomplex, and generally requires some understanding of principles from both soil androck mechanics.

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In Chapter V, the author introduced the idea that the effects of slaking in clay shalesmight be incorporated into the important and useful critical state soil mechanicsconcept. It was proposed that the process of slaking may simply provide an alternativepath for altering the void ratio of the clay shale within the constraints of the criticalstate model. Thus, by increasing the void ratio of the clay shale, slaking may act torejuvenate the clay shale, thereby reversing the effects of consolidation and aging whichmight have occurred in the past.

Triaxial compression tests were carried out on unaltered Pierre shale samples, and onsamples which were subjected to various degrees of slaking under confinement. Thepurpose of these tests were to provide initial data on the effects of slaking on thestrength and stress-strain response of a clay shale, and to assess the potential forincorporating the effect of slaking into the critical state concept.

The drying and rewetting of the Pierre shale is accompanied by significant excessswelling well beyond that required to return the shale to its original state. This excessswell results from the destruction of bonds in response to drying-induced slaking. Forone sample which was dried for 28 days prior to rewetting, the void ratio increasedfrom an initial value of .35 to a final value of .51.

The drying and rewetting data indicate the Pierre shale is transversely isotropic.Shrinkage and swelling strains are 2 to 3 times greater in the direction perpendicular tothe major plane of fissuration than those coincident to this plane. These results implythat the values for E and ν may vary with orientation.

The stress-strain response of the Pierre shale is characterized by an initial closing offissures, followed by a two-segment loading curve up to an abrupt failure. After peakfailure, strain softening reduces the strength of Pierre shale to less than 20% of thepeak strength. In addition, the slaking of the Pierre shale significantly reduced thevalues of Youngs modulus, while Poissons ratio changed little.

The strength of unaltered Pierre shale is rather high, with a peak strength of about 750psi for confining pressures of 30 psi. However, this strength was drastically reduced bya single cycle of partial drying followed by rewetting under confinement. For the

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sample which had dried for 28 days, the peak strength exhibited an 80% reduction tovalues very near the ultimate strengths of unaltered samples.

The strength results of all tests exhibit a significant amount of scatter, similar to thatobserved in the field. Using concepts taken from critical state soil mechanics, these datawere normalized using an exponential function of the specific volume. The normalizedstrengths strongly defined a two-segment Hvorslev failure surface, with friction anglesindicative of a rock-like material. The effective cohesion decreased from a value of69 psi to zero. Thus the softening of Pierre shale in response to slaking is accompaniedby a decrease in both the effective cohesion and effective friction angle.

The definition of a single failure envelope for both unaltered and slaked samples, ishighly significant. With such a tool, the variation of strength observed in the field canbe accounted for and mapped using simple and inexpensive measurements of the watercontent. In addition, by monitoring changes in the water content with time we shouldbe better able to assess the reduction of strength in the clay shale, and might be able toreasonably predict the rate of softening in the future.

The results from the present test program indicate that the slaking and naturalremolding capable of occurring in the field, may be much more intense than that whichhas been achieved in the laboratory by a single cycle of drying and wetting. Furtherresearch is needed to more completely define the failure surface for the Pierre shale,over the range of conditions observed in the field. Such studies should investigate thestrength of samples which have been intensely remolded and softened, either artificiallyor naturally, and should account for the effects of fissure orientation.

Finally, the results presented in this dissertation are very encouraging in that theysuggest a simple method which might alleviate many of the difficulties presentlyencountered during geotechnical analyses of clay shales. Still, the investigationsreported in this dissertation were only initial studies into a complex engineeringproblem. It is the author's hope that the results presented here will encourage continuedresearch into an interesting and important geotechnical problem.

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119. Schuster, R.L. (1965). Minors structures in the London clay, Thesis, ImperialCollege, London.

120. Scully, J. (1973). Landslides in the Pierre shale in Central South Dakota,South Dakota Dept. Highways Study No. 635(67), Final Report, pp. 707.

121. Serafim, J.L. and Lopez, J.J.B. (1961). In-situ shear tests and triaxial tests offoundation rocks of concrete dams, Proc. 5th Cong. Soil Mech Found. Eng.,Paris, v.1, 533-539.

122. Shamburger, J.H., Patrick, D.M., and Lutten, R.J. (1975). Survey of problemareas and current practices, Design and Construction of Compacted ShaleEmbankments, v. 1, Federal Highway Administration, FHWA-RD-75-61,Washington, D.C..

123. Sinclair, S.R. and Brooker, E.W. (1967). The shear strength of Edmontonshale, Proc. Geotech. Conf. Oslo, v.1, 295-299.

4. 125. Singh, R., Henkel, D.J., and Snagrey, D.A. (1973). Shear and Ko

swelling of overconsolidated clay, Proc. 8th Intl. Conf. Soil Mech. Found. Eng.,Moscow, v. 1.2, 367-376.

126. Skempton, A.W. (1942). The failure of a retaining wall at Kensal Green, BRSSoil Mechanics Record No. 3, Loc. 110.

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127. Skempton, A.W. (1948). The rate of softening in stiff fissured clays, withspecial reference to London clay, Proc. 2nd Intl. Conf. Soil Mech., v. 2, 50-53.

128. Skempton, A.W. (1953). Soil mechanics in relation to geology, Proc. of tehYorkshire Geol. Soc., v.29:1, 33-62.

129. Skempton ,A.W. (1961). Horizontal stresses in an overconsolidated Eoceneclay, Proc. 5th Intl. Conf. Soil Mech. Found. Eng., v.1, p.351.

130. Skempton, A.W. (1964). Long-term stability of clay slopes, Fourth RankineLecture, Gèotechnique, v. 14, p. 77.

131. Skempton, A.W. (1977). Slope stability of cuttings in brown London clay,Proc. 9th Intl. Conf. Soil Mech. Found. Eng., Tokyo, v. 3, 261-270.

132. Skempton, A.W., and Delory, F.A. (1957). Stability of natural slopes inLondon clay, Proc. 4th Inst. Conf. Soil Mech., London, v. 2, 378-381.

133. Skempton, A.M., and Hutchinson, J.N. (1969). Stability of natural slopes andembankment foundations, Proc. 7th Intl. Conf. Soil Mech. Found. Eng., State-of-the-Art Volume, Mexico City, p. 291.

134. Skempton, A.W. and La Rochelle, P. (1965). The Bradwell slip; a short-termfailure in London clay, Géotechnique, v.15, no.3, p.221.

135. Skempton, A.W. and Northerly, R.D. (1952). The sensitivity of clays,Géotechnique, v.3:1, 100-106.

5. 136. Stimpson, B. and Walton, G. (1970). Clay mylonites in English coalmeasures, Proc. 1st Cong. Intl. Assoc. Eng. Geol., Paris, v.2, 1388-1393.

136. Sture, S. and Ko, H.Y. (1978). Strain-softening of brittle geologic materials,Intl. J. Num. Analyt. Meth. Geomech., v.2, 237-253.

137. Sture, S. (1976). Strain softening behavior of geologic materials and its effectson structural response, Ph.D. dissertation, University of Colorado, Boulder,Colorado.

138. Tavenas, F. and Leroueil, S. (1977). Effects of stresses and time on yielding ofclays, Proc. 9th Intl. Conf. Soil Mech. Found. Eng., Tokyo, v.1, 319-326.

139. Terzaghi, K. (1936). Stability of slopes of natural clay, Proc. Intl. Conf. SoilMech., Harvard, v. 1, 161-165.

140. Terzaghi, K. (1961). Discussion on "Horizontal stresses on anoverconsolidated Eocene clay, Proc. 5th Intl. Conf. Soil Mech. Found. Eng.,v.3, 144-145.

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141. Terzaghi, K., and Peck, R.B. (1967). Soil Mechanics in Engineering Practice,2nd ed., John Wiley & Sons, NY.

142. Thomson, S., and Hayley, D.W. (1975). The Little Smoky Landslide, Can.Geot. J., v. 12, 379-392.

143. Tourtelot, H.A. (1962). Preliminary investigation of the geological setting andchemical composition of the Pierre shale, Great Plains Region, USGS Prof.Paper 390, pp. 72.

144. Tulinov, R. and Molokov, L. (1976). Role of joint filling material in shearstrength of rocks, Symp. Intl. Soc. Rock Mech., Nancy, paper II-24.

145. Twenhofel, W.H. (1950). Principles of Sedimentation, McGraw-Hill, NewYork.

146. Underwood, L.B. (1967). Classification and identification of shales, J. SoilMech. Found. Div., ASCE, v. 93, 97-116.

147. Uriel, S. and Serrano, A.A. (1973). Geotechnical properties of two collapsiblevolcanic soils of low bulk density at the site of two dams in Canary Islands(Spain), Proc. 8th Intl. Conf. Soil Mech. Found. Eng., Moscow, v.2.2, 257-264.

148. Vallejo, L.E. (1979). An explanation for mud- flows, Géotechnique, v.29,no.3, 351-354.

149. Vallejo, L.E. (1980). Mechanics of mudflow mobilization in low-angled clayslopes, Eng. Geol., v.16, 63-70.

150. Ward, W.H., Samuels, S.G., and Butler, M.E. (1959). Further studies of theproperties of London clay, Géotechnique, v.9, 33-58.

151. Weaver, C.E. (1967). The significant of clay minerals in sediments,Fundamental Aspects of Petroleum Geochemistry, B. Nagy and U. Colombo(ed), New York, Elsevier Pub. Co., 37-76.

152. Weisner, W.R. (1969). residual strength of over- consolidated clay shales,M.Sc. Thesis, Dept. Civil Eng., Univ. of Saskatchewan, Saskatoon.

153. Wentworth, C.K. (1922). Geology, W.C. Putnam, Oxford University Press,Inc., NY.

154. White, W.A. (1961). Colloid phenomena in sedimentation of argillaceousrocks, J. Sed. Petrology, v.31:4, 560-570.

155. Widger, R.A. and Frelund, D.G. (1979). Stability of swelling clayembankments, Can. J. Geotech., v.16, 140-151.

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APPENDIX A

CLASSIFICATION SCHEMES

The general characteristics of clay shales include (1) prestressed (i.e. highlyoverconsolidated), (2) commonly fissured with slickensides often present, (3) strongdiagenetic bonding, (4) tendency to slake when rewetted after drying, and (5) highswelling pressures in the presence of water. They have been referred to in the literatureas "stiff", "fissile", "intact", "compacted", or "brittle" clays, as well as "soil-like shale".Beyond this general description of clay shales, the classification of these materials hasbecome complicated and confusing. Numerous classification schemes for argillaceousmaterials have been proposed, and most are reviewed by Shamburger, Patrick, andCutten (1975) and Deen (1981), as well as in the following section.

Many of these classifications are geological and depend on such properties as quartzcontent, grain size, color, and the "degree of compaction". Although these provideimportant information regarding geological history of these materials, suchclassifications can be highly deceptive when concerned with engineering behavior. Thisis particularly evident when evaluating the behavior of clay shales. For example, Pierreshale can be described as a grey-black, highly-compacted shale with no visible tendencytoward fissility on a fine scale, whereas the argille varicolori might be described as a redor green, highly-tectonized, scaly clay. Yet these two materials have similarities in theirengineering behavior which would classify them together as clay shales.

Most engineering classification schemes for argillaceous materials employ such materialproperties as grain size, shear strength derived from simple tests, overconsolidationratio, and Atterberg limits. These classification schemes have proved useful fornormally consolidated and slightly overconsolidated clays, but have generally not beenadequate for transitional materials between highly overconsolidated clays and shales.As will be discussed below, the major deficiency of most engineering classificationschemes of argillaceous materials is the absence of an adequate time factor. Fortransitional materials, such as clay shales, it is important to not only consider how thesematerials behave at the present, but also how they will behave within a reasonableengineering time frame.

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Geological Classification

Most classification schemes of crustal materials are either geological or engineeringoriented, depending on the intended application. Although geological classificationshave proven useful for many practical problems the major objective of these schemes isthe determination of the geological history of the deposits. With this in mind, thevarious geological classification schemes for argillaceous materials are reviewed below.

Initial geological classifications were developed primarily on the basis of grain size. Thescheme of Wentworth (1922) arbitrarily set 0.0625mm as the boundary betweenargillaceous materials (shale or mudstone) and the remaining clastic, or fragmental,sedimentary rocks. Ingram (1953) subdivided clayey materials based on percentages ofsilt and clay components, and on their breaking characteristics (Table A.1). In thisscheme, the presence of fissility distinguishes "shales" from "stone", while the prefixes"clay", "silt", or "mud" are derived from the relative percentages of the grain sizecomponents. Therefore, such terms as claystone, siltstone, and clay shale began to beentrenced in the literature.

Folk (1968) clarified Ingram's scheme by refining "mudstone" as an argillaceousmaterials with subequal amounts of clay and silt (Table A.2). The classification schemeof Gamble (1972) is essentially the same as those of Folk and Ingram, except that theterms "clay shale" and "silt shale" have been changed to "clayey shale" and "silty shale".Although this change may seem insignificant, the term "clayey shale" does help todistinguish a clay-rich shale from a "clay shale" which, in engineering usage, impliescertain engineering behavior and not simply a fissile rock which is rich in clay content.

The classification of Underwood (1967) is the first attempt to divide shales into "soil-like" shale and "rock-like" shale (Fig. A.1). Although this scheme is essentiallygeological, it begins to approach the concept of a soil-like and rock-like behavior.

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Table A.1. Geological classification of mudrocks by Ingram (1953).

Table A.2. Geological classification of mudstone (Folk, 1968).

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However, the division between these two groups is poorly defined. Similarly,Skempton and Hutchinson (1969) attempt to crudely relate geological origin ofmaterials to their potential engineering behavior (Table A.3). The usefulness of thisscheme for purposes other than for providing a general understanding of possiblerelationships is quite limited.

Figure A.1. Classification scheme of Underwood (1967).

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Table A.3. Classification scheme of Skempton and Hutchinson (1969).

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Although the classification schemes above can provide some useful information forengineers, they are generally inadequate for assessing potential engineering behavior ofargillaceous materials. Regardless, the above review traces the use of the term "clayshale" in the geological literature to describe a fissile rock, rich in clay-sizedcomponents. As will be discussed below, this use of the term "clay shale" is notsynonymous with its use in engineering literature.

Engineering Classification

The objective of an engineering classification scheme is to categorize geologicalmaterials according to their potential engineering behavior. In this regard, anengineering classification is often oriented toward specific applications. This tends tocause some confusion among investigators when schemes employed for one applicationare considered valid for all applications.

Classification of argillaceous materials for engineering purposes has been particularlydifficult. Many of these difficulties have resulted because of the transitional nature ofsome of these materials. As will be discussed further, these transitional materials createfurther confusion for many geotechnical engineers who are accustom to viewing amaterial as either a rock or a soil, but not as a material that can have proper ties ofboth. In addition, few engineering classification schemes account for the potentialchanges in material behavior which can occur in a relatively short time in many of thesedeposits.

Terzaghi (1936) divided clays based on stiffness and the presence or absence offissures: (a) soft clays free from joints and fissures, (b) stiff clays also free from joints and fissures, (c) stiff fissured clays.Bjerrum (1967) proposed an overlapping three-fold classification, based on bondstrength and extending up to shale materials: (a) overconsolidated clays (i.e. overconsolidated clays with weak or no bonds), (b) clay shales (i.e. overconsolidated plastic clays with well-developed diagenetic

bonds), and (c) shales (i.e. overconsolidated plastic clays with strongly developed diagenetic

bonds).

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These two schemes have significant, but poorly distinguished overlap between themcreating some confusion of terms. Further confusion has developed from the use of theBritish Standard Institute classification, which uses similar terms based on consistencyor strength:

Consistency Field indication StrengthVery stiff Brittle or very tough >150 kN/m2

Stiff Cannot be molded in fingers 75 - 150Firm Molded in fingers by firm pressure 40 - 75Soft Easily molded in fingers 20 - 40Very soft Extrudes between fingers <20 kN/m2

Since the introduction of these classifications, more ambiguity in terminology hasresulted from the liberal use of the terms "overconsolidated shale" (Johnson, 1969;Fleming et al, 1970), and "stiff, fissured clay" (Chandler, 1970) to indicate a weakly-bonded shale. This inconsistency in terminology has been most pronounced and mostconfusing for the argillaceous materials which are transitional between normallyconsolidated clays and intact shales. Most of these difficulties result from inability’s toadequately account for potential changes in material behavior with time. Someinvestigators have attempted to account for this effect by including a "durability", orslaking factor in their classification schemes.

Gamble (1971) carried out extensive investigations on the durability of shales of allages, location, and consistency. Based on correlation’s of material properties, such aswater content, liquid limit, dry density, etc., he determined that these materials couldbest be grouped according to a relationship between a two- cycle slake durability indexand their plastic index (Fig. A.2). A clay shale is generally considered as having highplasticity and a low slake durability index. However, Gamble does not attempt to relatehis classification scheme to established terminology, and concludes that more work isnecessary to correlate laboratory results with field behavior.

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Similarly realizing the importance of shale deterioration, Deo (1972) proposed ascheme based entirely on shale durability. Three tests, all of which measure shaledurability (i.e. slaking, slake durability, and sulfate soundness), were performed onvarious shales from Paleozoic deposits in Indiana. Using indices derived from thesethree tests, Deo categorized Paleozoic argillaceous materials into soil-like shales, twotypes of intermediate shales, and rock-like soils (Fig. A.3). The most significant aspectof this scheme is the classification of the materials according to the their susceptibilityto deterioration rather than the initial state of the material.

Morgenstern and Eigenbrod (1974) were the first to attempt to combine earlierclassification schemes based on initial properties with those schemes based ondurability. They present two classification schemes, one based entirely on the slakingcharacteristics (i.e. the rate of slaking versus the amount of slaking), and a moresignificant scheme based on the undrained shear strength, strength loss after softening,changes of water content after softening, and the time of softening

Figure A.2. Classification scheme of Gamble (1971), based on the relationshipbetween slaking durability and plastic index.

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(Fig. A.4). The major importance of this scheme is the emphasis on changes of strengthand water content with softening.

Unfortunately, this scheme first divides argillaceous material into either soil or rockbased on three potentially conflicting properties, the undrained shear strength, thedegree of strength loss after softening, and the degree of change in water content aftersoftening. Only after this division are slaking characteristics used to determine if any ofthe "soil-like" materials are clay shales. According to this scheme, Pierre shale could beclassified as "rock-like" according to its initial strength characteristics, or as "soil-like"based on its response to softening. In addition, the Pierre shale, although rock-like ininitial strength, slakes completely to a soft mud with only one cycle of the slake-durability test.

The Transitional Nature of Clay Shale

The term "clay shale" has become widely adopted in engineering practice. Yet, thediscussion above illustrates that the term has also become poorly defined and nebulous.In its original meaning, the term "clay shale" did not necessarily imply that the materialwas particularly troublesome. However, the term "clay shale" has

Figure A.3. Classification scheme of Deo (1972), based entirely on resistanceto slaking.

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Figure A.4. Two part classification scheme of Morgenstern and Eigenbrod(1974).

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become increasing used within engineering practice to imply a stiff to rock-likeargillaceous material that is susceptible to significant detrimental changes in itsengineering behavior as a result of its interactions with water.

For example, Fleming et al (1970) refer to clay shales as clayey or silty,overconsolidated, sedimentary materials, which may or may not be slightly cemented.They continue by describing clay shale materials as tending to slake when exposed tocyclic wetting and drying. Materials cemented to the extent that they do not slake whenexposed to cyclic wetting and drying are termed siltstone or claystone ....

Furthermore the relationship of deterioration to the clay shales of Great Britain isillustrated by Attewell and Farmer (1976):

In a freshly excavated state, the Lias would be considered a true clay shale. Where itoutcrops ... it tends to degrade very readily into a clay/mud flow.... Such dessicationfractures, together with any jointing and laminations ... can rapidly reduce most clayshales ... to an aggregate in a few months, the actual length of time being a functionof wet-dry (slaking) cyclic frequency together with the compositional mineralogy andfabric of the rock.

It must be remembered that the purpose of any engineering classification scheme is toprovide terms which aid the user in distinguishing materials which have similarengineering properties. The more recent classification schemes for argillaceousmaterials have attempted to account for the important role of durability by includingslaking factors. However, these schemes have failed to recognize a unique property ofclay shales.

Other engineering materials are classified according to the engineering properties thatthey presently exhibit. Yet, a "clay shale" is unique not in its present properties, butrather in its potential for significant deterioration of these properties as a result ofinteractions with water. Classification schemes to date have failed to recognize that, asthe term now implies, a "clay shale" can encompass a stiff clay, such as the Londonclay, or a clayey shale, such as the Pierre shale. The terms "stiff clay" and "clayey shale"define these materials according to their present engineering properties. However, the

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London clay and Pierre shale can be further classified as "clay shales" based on theirpotential for significant reduction of strength upon wetting, or wetting and drying.

A clay shale is therefore defined by the present author as a stiff clay, or shale, whichcan undergo significant deterioration of its engineering properties as a result ofinteractions with water. This definition recognizes that the term "clay shale" implies atransitional material rather than a material with well-defined initial properties.

The classification scheme proposed here in Fig. A.5 employs previous classificationschemes, while placing clay shales within the proper perspective. According to thisscheme, the Pierre shale is an argillaceous material, a mudstone, and a clayey shale, aswell as a clay shale. Similarly, the London clay is a stiff, fissured clay because of itspresent engineering properties, and a clay shale because of its potential for rapiddeterioration.

Of course, an adequate factor is still needed to determine whether a stiff clay or clayeyshale could "significantly deteriorate as a result of interaction with water". Likeprevious schemes, this factor should most likely be a measure of the slaking potential ofthe material. However, as discussed in the main body of this dissertation, the author

Figure A.5. Modified classification scheme of the present author, showing clayshale as a unique class of argillaceous materials.

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questions whether the present practices of measuring slaking potential are adequate fordetermining the realistic response of these materials to slaking conditions in the field.

Summary

The term "clay shale" has become widely adopted, but increasingly nebulous inmeaning. This is partly the result of confusion between the use of "clay shale" in thegeological sense of a shale with a predominant clay fraction, and the use of the term inthe engineering sense of a clayey material that is intermediate between a firm clay andschemes to date have proved inadequate in establishing a widely accepted distinctionbetween clay shales and other argillaceous materials.

The term "clay shales" has become increasingly used in the engineering practice toimply a stiff to rock-like argillaceous material that is susceptible to relatively rapiddegradation with time. The most recent schemes have recognized the need for anadequate temporal factor in the classification of clay shales. This factor is typicallysome measure of the slaking characteristics of the material. As discussed within thispaper, there are some inadequacies in the present procedures for determining the slakedurability. Still, even considering these inadequacies, slake durability provides areasonable temporal factor for the classification of clay shales.

However, confusion has remained because the classification schemes to date have madefutile attempts to distinguish clay shales from stiff clays or clayey shales. Instead, itmust be recognized that a clay shale can be a "stiff clay" or "clayey shale" based on itspresent properties, and still be a "clay shale" based on its tendency to rapidly degradewith time. The author has proposed a modified engineering classification scheme forargillaceous materials, which places clay shales within this proper perspective.

The slaking tests employed in the present classification schemes of argillaceousmaterials include (1) the modified jar slaking test, (2) the slake durability test, (3) the"one-dimensional free swell test", and (4) the sulfate soundness test. In the modified jarslaking test (Moriwaki, 1974), an undisturbed or dried sample is placed on a wire mesh,which is then lowered into a jar of water. The slaking potential is indicated by the lossof weight as the material crumples and falls through the mesh. The slake durability testof Franklin and Chandra (1972) is performed by rotating six 40 to 60 gram samples in awire mesh drum, which is partly immersed in water. The weight percentage of material

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remaining in the drum after one cycle (200 revolutions in ten minutes) defines the"slake durability index". The sulfate soundness test is similar to the above tests, exceptthat the previously dried samples are wetted in a sodium or magnesium sulfate solution.The percentage of material retained on a 3/8" sieve indicates the sulfate soundnessindex.

The slaking factor of Morgenstern and Eigenbrod (1974) is unique in that it is based onthe one-dimensional free swell of a laterally confined sample. In these tests, the changeof height, and therefore the change of water content, were measured as a function ofwetting and drying cycles. Increased swelling is assumed to indicate progressive slakingwithin the specimen.

The modified jar slaking, slake durability, and sulfate soundness tests described above,all provide very useful information on the relative "weatherability" of variousargillaceous materials. However, the extent to which these tests significantly modelconditions existing in the field is limited. All of these tests measure the amount ofslaking under conditions of zero confining pressure. The review of slaking processes inChapter IV, discusses several different mechanisms of slaking, each caused by variousmagnitudes of disruptive stresses. It is conceivable that slaking forces exhibited bysome materials may be balanced by very low confining pressures. In such materials,slaking would only affect the very top surface of deposits in the field. In addition, theslake durability test employs a tumbling factor, which generally does not exist in thefield.

More important, however, is the fact that none of these tests measure the relativereduction of shear strength that occurs as these materials undergo deterioration in thefield. Although these tests do measure the important tendency of a material to degrade,they do not necessarily measure the relative reduction of strength due to degradation.

Present evidence regarding the relative importance of weathering of clay shales in thefield is inconclusive. There is debate as to whether the long term behavior of clay shalesis controlled predominantly by material deterioration or by the reduction of negativepore pressures with time. The modes by which weathering affects the engineeringbehavior of clay shales is also not well understood. The purpose of this dissertation isto investigate the role of slaking in altering the strength and stress-strain response of

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clay shales. Such research is necessary before a classification factor can be developedwhich will adequately distinguish the unique and troublesome clay shales.

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APPENDIX B

PROCESSES ACTING DURING THE FORMATIONOF CLAY SHALES

The unique characteristics of clay shales can be attributed to a combination of severalprocesses occurring during deposition and during subsequent unloading. Several of thefactors responsible for the present behavior of clay shales, such as lithology,compaction, and bonding, were developed prior to the unloading of the originalmaterial. This appendix therefore examines the geological development of a clay shaledeposit. It is important to develop a firm understanding of the physico-chemicalprocesses acting on clay shale before attempting to account for the unique behavior ofclay shale deposits.

Formation of Clay Shales

The behavior of clay shales is closely related to their lithology, which includes themineralogy and the amount of clay fraction present, as well as the degree ofcompaction, and the extent and strength of internal bonding. An increase in the clayfraction and in the percentage of montmorillonite results in increases in the plasticityand swelling potential and a decrease in the residual strength of clay shales. Themineralogy and grain size distribution of any deposit is originally controlled by themineralogy and climatic conditions existing in the source area from which the materialis eroded and transported, and by the energy regime of the transporting agent.

Most clay shales were deposited in a marine environment, although some are lacustrine(i.e. lake) deposits. Most of the present clay shales can further be characterized byrelatively high amounts of montmorillonite. Several studies presented by Blatt et al(1972), indicate that the relative amounts of montmorillonite, illite, and kaolinite inmodern marine deposits, can be directly related to the climatic environment of adjacentshores. As temperature and rainfall increase, illite is converted to montmorillonite, withgreater rainfall and temperature, montmorillonite might be degraded further tokaolinite. Therefore, kaolinite is abundant where tropical rivers empty into the ocean,whereas an abundance of montmorillonite reflects more moderate climatic conditions.

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However, the relative amount of montmorillonite in an area can be greatly increased bynearby volcanic activity, and most deposits rich in montmorillonite can be attributed tosuch a source. In particular, highly plastic bentonitic layers, such as those found in theclay shales of the upper Missouri River basin, are the result of in-situ alteration ofvolcanic ash.

In addition, chemical analyses on thousands of rocks in the U.S. indicate that therelative abundance’s of the illite, kaolinite, and smectite (i.e. montmorillonite) vary withthe age of the rock (Weaver, 1967). Such variations can be attributed to changes in thepaleoenvironment or to diagenetic processes which act to change the material afterdeposition. As shown in Fig. B.1, expandable clays have been relatively abundant from100 million year ago to the present. This may be a major factor accounting for the factthat most of the clay shales in the U.S. are Upper Cretaceous or younger.

An abundance of clay sized particles in a deposit can result from a similar abundance ofclay particles in the source area or from deposition in an environment far from high-energy streams (i.e. farther offshore). In a marine environment, sand and silt-sizedparticles are generally deposited nearer to shore, whereas clay sized particles can betemporarily suspended and carried farther offshore. In addition, kaolinite has beenshown to be more abundant closer to shore, whereas montmorillonite becomes moreabundant farther offshore (Blatt et al, 1972). This can be attributed to the small size ofmontmorillonite particles. Even though montmorillonite is highly reactive and tends toreadily flocculate in the marine environment, these floccules normally do not exceed thesize of kaolinite particles.In addition to the lithological effects discussed above, the behavior of clay shales isrelated to the degree of consolidation and the degree of unloading to which they havebeen exposed. Consolidation can be defined as the progressive decrease in the volumeof voids in a soil in response to loading, and in saturated materials is accompanied byexpulsion of water from the voids. Overconsolidation is a condition where the presentoverburden load is less than any previous load which the material has experienced. Thehistory of consolidation and unloading in clay shales can be very complex, involving thecyclic deposition and erosion of overlying materials, the advance and retreat of glaciers,cyclic wetting and drying, and loading and unloading by tectonism.

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The probable stress history and resulting variation of water content, of the Bearpawclay shale is graphically illustrated in Fig. B.2. Consolidation loads are generallyestimated from measurements in laboratory consolidation tests. However, Kenney et al(1967) have shown that the measured consolidation load can be greatly dependent onthe composition of pore fluid and cementing compounds, and can therefore beerroneous.

Still it appears that the Bearpaw, Claggett, Pierre, and Colorado clay shales in thenorthern plains of North America were consolidated by loads of a hundred tons/sq ft(about 10 MPa) or more (Fleming et al, 1970). As indicated by Table B.1, the loads

Figure B.1. Relative abundance of major groups of clay minerals inPhanerozoic mudrocks (Blatt et al, 1972; Weaver, 1967).

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Figure B.2. Probable stress history of the Bearpaw sediments (Scott andBrooker, 1968).

Table B.1 Estimated maximum preconsolidation loads on NorthAmerica clay shales units (Fleming et al, 1970).

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from glacial ice were apparently less than those from earlier overlying sediments. TheLondon clay is estimated to have been covered with 500 to 700 ft of overburden(Henkel, 1957), resulting in consolidation loads of about 150 tons/sq ft, or 15 MPa.Consolidation loads of 13 to 20 MPa have been reported by Banks (1971) for theCucaracha clay shale in Panama, while those estimated for the Italian clay shales rangefrom 0.3 to 5.0 MPa (Cancelli, 1981; Esu and Grisolia, 1977; A.G.I., 1977; Fenelli etal, 1982). The stress history of the Italian clay shales has in general been greatlycomplicated by additional tectonic shearing forces.

The strength of clay shales and their subsequent softening depend on the extent andstrength of the bonds that form during and after consolidation. The physico- chemicalprocesses that are believed to be responsible for this bonding are discussed in greatdepth throughout the book of Mitchell (1976), and will not be reviewed extensively inthis dissertation. The nature of bonding is very complex, involving interlayer forcesacting within clay particles, as well as those forces acting between individual clayparticles. In some clay minerals, such as illite and kaolinite, the interlayer bonding maybe very strong; in contrast, the interlayer bonds within particles of montmorillonite aremoderate, readily allowing separation between layers and thus swelling of the clayparticle.

Bonding between clay particles involves the complex interplay of attractive van DerWaals forces and the repulsive forces acting between the tightly held viscous layers ofwater that surround the particles. In the schematics of Fig. B.3 is illustrated the natureof these forces as they are believed to exist for different conditions of chemicalenvironment and mineralogy. It can be seen that under some conditions, there is aninitial repulsion between particles as these particles are forced together. However, ifthis "threshold" of repulsion is exceeded, the forces become attractive and thereforeprovide a degree of bonding between particles. In addition to the forces discussedabove, very strong bonds can be developed between particles by the introduction ofchemical bonds, or cementation, between these particles. All of these bonds act toinfluence the nature of the fabric during deposition and compaction, and subsequentlyinfluence the susceptibility of the material to loss of strength during unloading andweathering.

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The geological history of overconsolidated clays and clay shales was discussed bySkempton (1964), and is illustrated in the schematic of Fig. B.4, presented by Fleminget al (1970). Point (a) represents a unit of clay immediately after deposition. The clayat this stage has zero horizontal and vertical stresses acting on it and is characterized bya very high water content and essentially no strength. As the effective vertical stress isincreased by the continual deposition of overlying sediment, as at point (b), the watercontent of that clay unit decreases, the shear strength increases, and the horizontalstresses increase such that the Ko ratio of horizontal to vertical stress is constant but

less than one. This process continues to the stage represented by point (c), at whichtime further deposition is halted. At the points (a), (b), and (c), the clay is "normally-consolidated" since it has never been exposed to overburden loads greater than the loadexisting at that stage.

Figure B.3. Energies of repulsion, attraction, and net curves of interaction forparallel flat plates (Mitchell, 1976).

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For clay shales, the consolidation stage described above is followed by a period ofvertical stress release, resulting for example from the erosion of overlying sediments.The clay unit thus "rebounds" in response to this unloading by regaining some of thewater lost, and losing some of the strength gained during the consolidation process.The clay is now "overconsolidated" since the overburden stresses are less than they hadbeen in the past. The rebound effect exhibited by the clay unit does not typically resultin total recovery of water content or total loss of strength. The effective vertical stressacting on the overconsolidated clay unit at point (d) in Fig. B.4, is the same as thatexperienced by the normally consolidated clay at point (b). However, in theoverconsolidated stage, the clay has a lower water content and is stronger than thenormally consolidated clay at the same effective confining pressure. Upon reloading ofthe overconsolidated clay, the void ratio response is approximated by that shown inFig. B.5. The clay would initially exhibit low compressibility until the stress reaches thehighest load experienced by the sample, or the "preconsolidation" load, after which thevoid ratio curve again follows the curve defined by the normally consolidated clay.

The time period between the process of deposition and the beginning of unloading canbe considerable. Large overburden stresses acting over this long period of time canresult in significant reduction of the void ratio, a rearrangement of soil particles, and asubsequent increase in the number and strength of interparticle bonds. This process hasbeen referred to as "secondary compression" or "aging", and its occurrence duringconsolidation tests in the lab has been reported by Leonards and Rahmiah (1960). Theresults showed in Fig. B.6, illustrate the effects of a sustained load on decreasing thevoid ratio and compressibility of a normally consolidated clay. Upon initial reloading,the clay specimen acts as though it was an overconsolidated clay which has experienceda preconsolidation pressure higher than that which has been applied.

The effect of sustained loading over geological time was shown by Bjerrum (1972) tobe similar. As illustrated in Fig. B.7, Bjerrum has suggested that the rate of decrease ofthe void ratio is roughly proportional to the logarithm of time. Therefore, if a clayundergoes secondary compression for a period of 10,000 years, about 50% of thecompression will have occurred in the first year and about 80% after 100 years. Theclay which sustained a load of po for 10,000 years in Fig. B.7, will act upon reloading

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Figure B.4. Schematic of the stress history of an overconsolidated clay overgeological time (Skempton, 1964; Fleming et al, 1970).

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as though it were an overconsolidated clay which has experienced a greaterpreconsolidation load of pc'.

This behavior can be attributed to increased bonding arising from recrystallization ofparticles, adhesion between particles, or precipitation of cementing agents betweenparticles. Bjerrum (1967) has presented a scheme for the geological history ofoverconsolidated clays and clay shales similar to that presented by Skempton (1964)and discussed above. However, the scheme of Bjerrum accounts for the effects ofdiagenetic bonds formed during or after the consolidation process. Referring back to

Figure B.5. Schematic showing void ratio response to reloading after rebound.

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Figure B.6. Aging effects observed in the laboratory for a normallyconsolidated clay (Leonards and Rahmiah, 1960).

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Figure B.7. Schematic illustrating aging effects resulting from sustained loading overgeological time (Bjerrum, 1972).

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the schematic in Fig. B.4, if the clay unit which has been consolidated to the point (c)then undergoes extensive secondary compression with time, the properties of the claywould then be represented by the point (c)'. Water content would be reduced, bondingwould be increased, and as a result, the strength would be increased. This bonding canbecome so strong that the material can then be classified as a rock.

During unloading, the clay has a tendency to swell and to pull in more water. However,the stronger the bonds formed during consolidation, the more the clay will be inhibitedfrom swelling and taking on more water. Therefore, some of the strain energy gainedduring consolidation may be retained by the clay after unloading, depending on thestrength of the bonds. Clays which have weak bonds will release all of their strainenergy immediately upon unloading. In contrast, clays or claystones with strong bondsmay retain some strain energy upon unloading, but will release this energy slowly inresponse to weathering processes. According to the definition advanced in AppendixA, such materials would be classified as clay shales. Clays and claystones which haveformed what Bjerrum called permanent bonds, may never release their stored strainenergy.

During the loading of clay deposits, the horizontal stresses increase in direct proportionto the vertical stress, so that the ratio of horizontal to vertical stresses, Ko, is constant

and less than one. Upon unloading, the clay is able to expand vertically more easilythan in the horizontal direction. Therefore, vertical stresses are released to a greaterextent than horizontal resulting in a relative increase in the horizontal stresses and anincrease in the ratio, K. Because clays with weak bonds tend to undergo extensiveexpansion upon unloading, the horizontal stresses in these materials can be very highrelative to the vertical overburden. Similar results have been observed for clay shaleswhich have expanded in response to weathering near the surface.

Results obtained by Brooker (1967) suggest that the susceptibility to disintegration byslaking is a function of the stored strain energy and that the amount of absorbed strainenergy is dependent on the mineralogy. Plastic clays containing significant percentagesof illite, and in particular, montmorillonite, exhibit a high tendency to store largeamounts of strain energy, and to release this energy in response to weatheringprocesses. This is consistent with the relationship between these minerals and theresponse of clay shales observed in the field.

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It can be summarized that the formation of a clay shale requires one or more cycles ofextensive loading followed by a period of unloading, during which strong but notpermanent, bonds are formed. The tendency of a clay deposit to develop into a clayshale is greatly enhanced by the presence of montmorillonite, and to a lesser degree,illite. Although a clay shale may appear to be much stronger than otheroverconsolidated clays as a result of stronger bonds, these bonds are readily destroyedby the processes of weathering occurring near the surface of the earth, resulting in arapid and drastic loss of strength.

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APPENDIX C

THE MECHANICS OF SINGLE FISSURES

Clean joints. The shear strength along a perfectly straight and smooth joint iscontrolled simply by the residual strength of the wall material. However, as illustratedin Fig. C.1, joint surfaces may be undulated or stepped, as well as planar, and on asmaller scale, these surfaces may be rough, smooth, or very polished. The effect ofnon-planar, or rough, joint surfaces is to create a certain degree of interlockingbetween the two joint surfaces, as they try to slip past each other. Therefore, in orderfor slip to occur, these rough peaks or "asperites" must either slide over one another,resulting in dilation normal to the joint, or the asperities must themselves break byshear or tensile failure.

This behavior was emphasized by Patton (1966), who developed a model for the shearstrength along a joint based on simple geometric relationships and the assumption ofzero cohesion. From the geometric relationships shown in Fig. C.2a for an inclined,planar joint, the normal and shear stresses acting on the surface are given by

Γi = Γcos2i + σ sin i cos i, (A.1)

σi = σ cos2i + Γ sin i cos i, (A.2)

and assuming zero cohesion, its shear strength by

Γi = σ itan φ. (A.3)

By combining these equations, one obtains the shear strength in terms of normal andshear stresses acting along the primary joint direction

Γ = σ tan (φ + i). (A.4)

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Figure C.1. Schematic illustrating various shapes and roughnesses that arepossible in joints.

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Patton demonstrated in a series of tests on models with regular surface projections, thatat low normal stresses this equation likewise gives the strength of the analogous jointmodel shown in Fig. C.2b. At higher normal stresses, however, the shear strength ofthe wall material is exceeded, and failure begins to occur through the inclinations, orasperities. In accordance with Patton's model, the shear strength at higher stresses, istherefore given by the shear strength of the wall material, which in terms of Mohr-coulomb theory, is

Figure C.2. Schematic of the joint models used by Patton (1966).

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ΓI = cI + σ tan φI. (A.5)

Ladanyi and Archambault (1970) noted that the transition from dilation to shearingthrough asperities would be smooth rather than abrupt, and based on theory andexperimentation, proposed that the strength of a joint could be defined by the equation

I = (1-as)(v + tan φ) + as ΓI (A.6)

1 - (1 - as)v tan φ

where as is the proportion of the discontinuity surface which is sheared through intactrock material, v is the dilation rate at peak shear strength, and I is the shear strength of

intact material. Ladanyi and Archambault suggested that the shear strength of intactmaterial next to the joint be given by the parabolic equation of Fairhurst (1964):

ΓI = σj √1+n - 1 (1 + n σ )1/2 (A.7)

n σj

where σj is the uniaxial compressive strength of the rock material adjacent to the

discontinuity, and n is the ratio of uniaxial compressive strength to uniaxial tensilestrength of the adjacent rock material. It should be noted that, duet to weathering orloosening, σj may be lower than the uniaxial strength of the material within intact

blocks.

The plot in Fig. C.3 compares the normalized shear strength of a joint as predicted bythe equation of Ladanyi and Archambault, to that predicted by the Patton equation.According to Patton's analysis, the strength envelope would change abruptly once theshear stresses exceeded the shear strength of the intact material. In contrast, theenvelope of Ladanyi and Archambault exhibits a more gradual transition from thestrength as predicted by Patton to the intact material strength, and demonstrates thatthe strength may be overestimated by Patton's equation. However, problems do existwith the equation of Ladanyi and Archambault, particularly with regard todetermination of the values for as and v. These parameters must generally be

determined empirically, leading to rather complex equations.

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An alternative approach has been proposed by Barton (1973), who developed the JointRoughness Coefficient, JRC. Based on careful tests on artificially produced "joints",Barton derived the following empirical equation for the shear strength of a joint:

Γ = σ tan [ φ + JRC.log10(JCS/σ)] , (A.8)

where JCS is the Joint Wall Compressive Strength, which is equivalent to σj in the

equations of Ladanyi and Archambault. Thus the effects of joint roughness are moredirectly seen in the equation of Barton. The value of the JRC is determined from plotssuch as in Fig. C.4 and is dependent on the length of the joint over which shear occurs(Pratt et al, 1974; Bandis et al, 1981). This dependency on scale, and the effects ofdifferent values of the JRC, are illustrated in Fig. C.5. Therefore, even though a large-scale fault may be extremely rough, its JRC may be very low due to the extreme

Figure C.3. Comparison of normalized shear strength of a joint as predicted byequations of Ladanyi and Archambault and the model of Patton.

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Figure C.4. Roughness profiles and corresponding ranges of JRC valuesassociated with each.

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length over which slip occurs. Similarly, a joint which has a low JRC value in the field,may have a high JRC value with regard to the size of samples tested in the laboratory.

All of these models emphasize the importance of the degree of roughness along thejoints. The dependency on the scale of roughness relative to the magnitude ofdisplacement has likewise been emphasized. However, no methods have beendeveloped to adequately measure and quantify the roughness of joint surfaces at allscales, although the profilometric and analytical procedures presented by Farrington(1983) are promising.

The few models reviewed above primarily attempt to define the failure envelope for thepeak strength of single joints, and therefore do not describe the stress strain

Figure C.5. Plot showing the dependency of shear strength on scale and JRC.

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behavior before or after failure. At least three models have also been proposed toaccount for this behavior of joints based on elasto plastic concepts. The model of

Agbabian, as presented by Ghaboussi et al (1973), and the Goodman model (Goodman,1966, 1974; Goodman and Dubois, 1972; Goodman et al, 1968) adequately modelsome behavioral aspects of fissured materials, but have been shown by Roberds andEinstein (1978) to be somewhat limited.

Roberds and Einstein (1978) proposed a comprehensive model for the behavior of rockjoints based on elasto-plastic principles and the concept of critical state mechanics. Asshown in Fig. C.6, the model accounts for the complex behavior of joints by allowingfor four yield surfaces: (a) the intact rock yield surface, which acts in nonjointed rocksor when the normal stress is very high, (b) the discontinuity, or joint, yield surface,which governs yield at low stresses once a joint has formed, (c) the discontinuityresidual yield surface, and (d) the ultimate residual yield surface, given by the CS lineand determined apparently as the strength of a planar, perfectly smooth joint. In

Figure C.6. Critical state model of Roberds and Einstein (1978) for thebehavior of rock joints.

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Fig. C.6, is illustrated the stress path corresponding to the loading of an intact rockuntil fracture is initiated, followed by unloading, then reloading to the strength givennow by the discontinuity yield surface, and finally by the discontinuity residual yieldsurface.

Within this scheme, the ultimate residual strength, or critical state condition, can beobtained if the joint is sheared at sufficiently high stresses to again mobilize the intactrock strength. Evidence for two "residual" yield surfaces has been presented by Krahnet al (1979), Insley et al (1977), Weisner (1969), and Thomson and Hayley (1975),who report that the strength mobilized in Canadian clay shales was found to be equal tothe residual strength of remolded, precut samples, rather than the residual strength ofundisturbed samples.

Roberd and Einstein have demonstrated that the previous models of Patton, Barton,and Ladanyi and Archambault can be incorporated into the comprehensive critical statemodel. Although much more investigation is required to define the yield surfaces ofjointed rock, the critical state model of Roberds and Einstein, provides a usefulframework within which further advances in analyzing the behavior of clean rock jointsmight be made.

Filled joints. Joints in nature are often filled with materials with different mechanicalproperties than the joint wall material. These can consist of materials which have beencarried into the joint by water, or material derived from the joint walls by weathering orby crumbling during shear. Several studies have investigated the strength of filledjoints, although most have been concerned with a specific problems encountered duringconstruction projects. The studies of most interest to this dissertation include thosedealing with clay filling (Brekke, 1965; Brekke and Selmer-Olsen, 1965; Brekke andHoward, 1972; Sinclair and Brooker, 1967; Leussink and Muller-Kirchenbauer, 1967;and Stimpson and Walton, 1970), and those dealing with the effects of wall rockalteration (Serafim and Lopez, 1961; Rocha, 1964; Patton, 1966; and Deere andPatton, 1971).

More systematic studies have been performed by Goodman (1970), Tulinov andMolokov (1971), and Goodman et al (1972). An extensive review of the studies prior

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to 1974 on filled joints was presented by Barton (1974). The findings of these, andlater, studies can be summarized, as per Ladanyi and Archambault (1977), as follows:

(1) For most filled joints, the failure envelope is located between that for thefilling and that for a similar clean joint.

(2) As illustrated in Fig. C.7, the strength and stiffness of filled joints decreasesgradually as the thickness of the filling increases relative to the amplitude of theasperites; although the strength and stiffness approach that of the filling, theystill remain significantly higher than the strength of the filling alone, even at100% filling thickness.

Figure C.7. Shear strength of a rough joint as a function of the joint-fill thickness.

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(3) The stress-strain curves of filled joints often display two segments, onereflecting the deformability of the filling before rock to rock contact, and theother reflecting contact between asperities, after the filling has been sheared.

(4) The dilation rate at failure decreases with increasing normal pressure andeventually becomes negative (i.e. contractive) at high normal pressures.

(5) The strength of filled joints does not always depend on the thickness of thefilling. If the joint walls are flat and smooth, or covered by a coating with avery low coefficient of friction, the shear plane is always located along thecontact between the wall and the filling.

(6) One of the most dangerous fill, or gouge, materials is a swelling clay,because of its loss of strength due to swelling and the high pressures created ifswelling is prevented.

Thus, the thickness of filling required to significantly reduce the strength of a jointdepends on the roughness of the joint walls. For joints which are smooth and planar,the strength of the joint can be rapidly reduced to the strength of the fill material. Inaddition, if the strength of the fill material is much lower than the shear strength alongthe clean joint, the reduction of strength resulting from the presence of a filling can berather drastic. Finally, a filling of swelling clay is particularly troublesome due to theloss of strength and high pressures associated with swelling.

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APPENDIX D

THE STRENGTH OF A CLAY MASS CONSISTINGOF INTACT CORES SURROUNDED

BY A MUD MATRIX

At some stage during the softening of clay shale, the deposit might be considered as anassemblage of stiff, unaltered clay cores, surrounded by a soft, remolded clay, or mud,matrix. In evaluating the strength of such a clay mass, we will consider three cases: (a)the intact cores, or grains if you will, are totally separated by the clay matrix and thereis essentially no interaction between the cores, (b) the cores are in contact, but there isno interlocking, and (c) the cores are closely packed, resulting in particle interlock. Forthe first case, the strength of the mass is provided entirely by the soft, clay matrix.Therefore it is important to note that it is not necessary to soften the entire mass of aclay shale before it will exhibit the "fully softened", or remolded, strength. It is onlynecessary that the concentration of intact clay cores is sufficiently low as to inhibitsignificant interaction between intact cores.

The second and third cases have been analyzed by Rowe (1962) using Mohr-Coulombtheory. The second case involving contact of the cores, but no interlocking, isillustrated in Figs. D.1a. From the analogous block model in Fig. D.1b, the pressure Prequired to slide the block is equal to the interparticle cohesive force,C, and the interparticle frictional force, Q.tan µ , sothat

P = C + Q.tan µ. (D.1)

However, if adjacent intact cores are interlocked as in Fig. D.1c, the analogous blockmodel in Fig. D.1d indicates that force required for sliding is now equal to

P = {C / [cos β (1-tanβtanµ)]} + Q.tan(µ+β) . (D.2)

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Thus, the effective friction angle now includes an interlocking, as well as aninterparticle, component,

φe = β + µ ,

(D.3)

and the effective cohesion, likewise, includes both interlocking and interparticlecomponents, Ce = C / [cos β (1-tanβtanµ)] . (D.4)

These two cases are essentially the same as for loosely, and densely, packed granularsoils, respectively.

Figure D.1. Two dimensional friction model for granular materials (Rowe,1962).

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Rodine (1974) performed special cone penetration tests on various mixtures of sand-sized particles within a clay-water slurry in order to assess the effects of particleconcentration on the strength of the mixtures. He determined that the concentration ofparticles of a single size, could be 45 to 65% of the total volume before significantinterlocking occurred, and up to 64% for particles of two select sizes. However,theoretical analysis of mixture of multi-sized particles suggests that interlocking wouldbe negligible at particle concentrations as high as 89 to 95%.

Once initiated, mudflows consisting of hard clay fragments surrounded by a mudmatrix, are characterized by very low strengths and the ability to transport largefragments, and even boulders, over long distances on shallow slopes (Rodine andJohnson, 1976; Hampton, 1979). The least slope angle at which these mudflows aremobilized in the field is always smaller than the slope angles predicted by the method ofSkempton and DeLory (1957) based on residual strength conditions (Vallejo, 1979;1980). Vallejo, as well as Hampton (1979), have shown that these very low strengthscan be accounted for using the particle "dispersion" theory, proposed by Bagnold(1954; 1956).

The present author has observed the presence of very low-strength mudflows whichflow for long distances on slopes of only a few degrees, at the village of Bisacchia,Italy. These highly fluid mudflows originate from rather steep, and apparentlymetastable, slopes composed of "argille varicolori". The transition of this clay shaleinto a mudflow is surprisingly drastic, with the mudflows originating from within deep"collapse channels" which are lined by near vertical walls. These collapse channels arevery similar in appearance to those often associated with the failure of "quick clays".Although there does appear to be some softening occurring along many of the fissures,the wall material itself has not been softened to any great extent. The proportion ofmaterial which has been significantly softened certainly does not exceed five percent.Such mudflows may originate in the manner of the flows discussed above, and suggestagain that very little softening is required to develop a highly fluid clay mass.

In summary, if the concentration of intact cores is low, the strength of the mass will beequal to the strength of the matrix material. The strength in this case would bedependent on the water content of the matrix material and might be analyzed inaccordance to the critical state model. If the concentration of intact clay cores ishigher, such that there is interference between cores, then the material should be

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modeled as a granular soil whose strength is controlled by the interparticle andinterlocking components of strength acting between the intact cores, as well as thedegree of packing of the cores. However, two factors make the analysis of highly-altered clay shale more complicated than that for typical granular materials: (a) unlikepore water which essentially has no resistance to shear, the interparticle "fluid" in analtered clay shale has a shear strength of its own, as well as a component of strengthacting at the boundaries between the fluid and cores, and (b) the "grains", or cores, maythemselves be quite soft, and therefore cannot be assumed to be infinitely stiff.

The interlocking component of strength, which is certainly of great importance infissured clay shales, could in theory be eliminated if only 5% of the clay shale materialis softened. In reality, the percentage of softening required to eliminate interlocking isprobably somewhat higher, but may still be surprisingly low. The percentage of clayshale material which must be softened before the "fully softened" strength is obtainedhas not been determined. However, it is not necessary to soften the entire clay shalemass, nor even a large percentage of the mass, before mobilizing "fully softened"strength.

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APPENDIX E

PRESENTATION OF DATA

THE FOLLOWING PAGES PRESENT DATA IN THE FORM OFGRAPHS, SKETCHES, AND NOTES FOR EACH TEST.

ALL GRAPHS SHOULD BE SIMILAR IN SCALE FORCOMPARISON.

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