CE3099 Individual Project Bekar Bedir 0707295 MEng

8/2/2019 CE3099 Individual Project Bekar Bedir 0707295 MEng

http://slidepdf.com/reader/full/ce3099-individual-project-bekar-bedir-0707295-meng 1/66

Bedir Bekar - 0707295

Performance-basedstructural design of a RC

high-rise building in

central London with

seismic considerations CE3099 – Individual Project



2

Contents1 Introduction ....................................................................................................................................... 4

1.1 Purpose of project ...................................................................................................................... 4

1.2 Project overview ........................................................................................................................ 4

1.2.1 Site location ....................................................................................................................... 4

1.2.2 A4 tower ............................................................................................................................ 5

2 Design process ................................................................................................................................... 6

2.1 Project literature ........................................................................................................................ 7

2.1.1 Eurocodes .......................................................................................................................... 7

2.1.2 Eurocode 1 - EN1990 - Actions on structures ................................................................... 8

2.1.3 Eurocode 2 - EN 1992: Design of concrete structures..................................................... 10

2.1.4 Eurocode 8 - EN 1998: Design of structures for earthquake resistance .......................... 14

2.2 Design guides .......................................................................................................................... 19

2.2.1 IStructE - Manual for the design of building structures to Eurocode 1 and Basis of

Structural Design ............................................................................................................................. 20

2.2.2 IStructE - Manual for the design of concrete building structures to Eurocode 2 ............ 20

2.2.3 IStructE - Manual for the seismic design of steel and concrete buildings to Eurocode 8

20 3 Scheme Design ................................................................................................................................ 21

3.1 General Arrangement .............................................................................................................. 21

3.2 Occupancy loads ...................................................................................................................... 22

3.3 Dead loads ............................................................................................................................... 23

3.3.1 Slab dead load .................................................................................................................. 23

3.3.2 Cladding perimeter load .................................................................................................. 24

3.4 Load combinations .................................................................................................................. 24

3.5 Structural durability and member sizing.................................................................................. 24

3.5.1 Durability ......................................................................................................................... 24

3.5.2 Member sizing ................................................................................................................. 25

3.6 Wind load ................................................................................................................................ 27

3.7 Serviceability and comfortability criteria for wind loading..................................................... 33 4 Primary design and analysis ............................................................................................................ 34

4.1 FEA Model .............................................................................................................................. 34

4.1.1 Construction and Assumptions ........................................................................................ 35

4.1.2 Loadings .......................................................................................................................... 38

4.1.3 Materials .......................................................................................................................... 38

4.1.4 Load cases and combinations .......................................................................................... 39

4.1.5 Simplification and limitations in modelling .................................................................... 39

5 Secondary design and analysis ........................................................................................................ 40

5.1 Desirable characteristics for desired lateral response and dynamic analysis ........................... 40

5.2 Seismic analysis ....................................................................................................................... 41

5.2.1 Seismicity of site ............................................................................................................. 41

5.3 Structural dynamics ................................................................................................................. 42 5.3.1 Modal analysis outputs .................................................................................................... 43

5.4 Structural response to ground motion ...................................................................................... 46

5.4.1 Desirable characteristics of earthquake resistant buildings ............................................. 46

5.4.2 Damage limitation ........................................................................................................... 46

5.4.3 Response spectrum analysis to EC8 ................................................................................ 47

5.4.4 Multimodal analysis response spectrum analysis using SAP2000 .................................. 51

6 Design Performance ........................................................................................................................ 53

6.1 Comfortability criteria performance ........................................................................................ 61

6.2 Design implications ................................................................................................................. 62

6.2.1 P-Δ effects ....................................................................................................................... 62

6.2.2 Axial shortening .............................................................................................................. 63

7 Conclusions ..................................................................................................................................... 63

8 References ....................................................................................................................................... 65



3

List of FiguresFigure 1 - Site location and layout of Canada Quays A Sites ....................................................................5

Figure 2 - Architectural impression and typical section detail (Courtesy of URS Corp. Ltd) ...................5

Figure 3 – Project process overview ..........................................................................................................6

Figure 4 - Relation of Eurocodes to each other (IStructE manual to EC1) ................................................8

Figure 5 - Wind load distribution on a tall building ...................................................................................9

Figure 6 - Punching shear failure of RC flat slab (Taranath, 2010) .........................................................11

Figure 7 - Typical flat slab section ...........................................................................................................11

Figure 8 - Lateral load deformation of shear wall ....................................................................................12

Figure 9 - Lateral load deformation of rigid frame ..................................................................................12

Figure 10 - Lateral load behaviour of shear wall-frame structure ............................................................13

Figure 11 - Three stages of seismic modelling (Bommer and Stafford, 2009) ........................................14

Figure 12 - High-rise building behaviour during earthquakes (Taranath, 2010) ......................................15

Figure 13 - Forced vibration of mass-spring damper system (Elghazouli, 2009) ....................................17

Figure 14 - Base vibration subjected on a mass-spring-damper system (Elghazuoli, 2009) ....................17

Figure 15 - damping effects on free vibrations (Williams, 2009) ............................................................18

Figure 16 - Mode shapes of a four storey building and shear wall ..........................................................19 Figure 17 - 1st floor general arrangement ................................................................................................21

Figure 18 - Simplified general arrangement of an individual floor. .........................................................22

Figure 19 - Simplified elevation ...............................................................................................................23

Figure 20 - Internal RC slab components .................................................................................................23

Figure 21 - Span/depth ratio for solid flat slabs .......................................................................................25

Figure 22 - Values for Cdir from UK NA to EC1 part 1-4 ........................................................................27

Figure 23 - Aerial view of site location and surrounding terrain (location of A4 in red triangle) ...........28

Figure 24 - obstruction height and upwind spacing (IStructE guide to EC1 pp.108) ..............................28

Figure 25 - Eurocode equivalent to wind loading for buildings up to 200m (IStructE, 2010) .................29

Figure 26 - Key to external pressure coefficients for a rectangular plan building (IStructE, 2010) ........30

Figure 27 - adapted external pressure coefficient key ..............................................................................31

Figure 28 - human perception levels to wind induced acceleration (Hira, 2003) ....................................33 Figure 29 - Typical floor plan and elevation of A4 model in SAP2000 ..................................................36

Figure 30 - extruded view of FEA model with offsets and insertion points of structural elements .........37

Figure 31 - Use of the diaphragm constraint to model a rigid floor slab .................................................37

Figure 32 - Static wind point loads applied on positive wind pressure face ............................................38

Figure 33 - Seismic hazard map of Peak Ground Accelerations on rock (PGA) for 475 year and 2500

year return periods (IStructE manual to EC8) ..........................................................................................41

Figure 34 - Interstorey drift (EC8) ...........................................................................................................47

Figure 35 - Displacement amplification factor curves for an SDOF structure subject to sinusoidal

ground shaking (Williams 2009) ..............................................................................................................48

Figure 36 - Typical time-history accelerogram ........................................................................................48

Figure 37 - Typical response spectra with envelope (left) with EC8 Response spectrum (right) ............49

Figure 38 - Values of horizontal response spectrum parameters recommended in EC8 ..........................50 Figure 39 - Methods of analysis for new buildings permitted by EC8 (IStructE manual to EC8) ...........50

Figure 40 - Response spectrum parameters for EC8 in SAP2000 ............................................................52

Figure 41 - Deformed shapes in FEA model ............................................................................................53

Figure 42 Concept of transferring lateral load and flexural deformation of vertical cantilever ...............54

Figure 43 – Maximum axial forces in shear core .....................................................................................55

Figure 44 - anticipated region of wall-frame interaction..........................................................................56

Figure 45 - Lateral storey drift for Unfavourable loading ........................................................................58

Figure 46 - Lateral storey drift for favourable loading ............................................................................58

Figure 47 - Inter-storey drift for unfavourable loading ............................................................................59

Figure 48 - Inter-storey drift ratio for unfavourable loading ....................................................................59

http://y/CE3099_Individual%20project_Bekar%20Bedir_0707295_MEng.docx%23_Toc288826712
















































































































































4

1 Introduction

1.1 Purpose of project

The intention of this project is to demonstrate the design considerations, technical processes and appliedanalytical methods in the performance based structural design process of a high-rise reinforced concrete

(RC) structure situated in Central London. The report covers the work undertaken throughout the

analytical stages and the corresponding design implications the structural behaviour will have on the

detailed design process to provide a functioning RC frame fit for the planned end-use of the building.

The proposed structural form of the building is based on real-life client specifications for a project

provided by URS Corporation Ltd.

Analysis and design of the structure is to conform to relevant design codes (Eurocodes) and encompass

the use of Finite Element Analysis (FEA) software, in conjunction with applied engineering judgement.

The project intends to outline the fundamental concepts and applications of engineering judgement

required for high-rise building design. Essentially, this can be characterised as:

Structural analysis and design to requirements of Eurocode 2

Calculation of Wind and seismic loading to relevant Eurocodes and corresponding

application to FEA model

Consideration of global structural performance to and secondary seismic design

performance to Eurocode 8

Encompass usage and learning of Finite Element Analysis software (SAP2000)

The intention of the project is not to detail and analyse the complex behaviour of every individual

structural element to the level of final detailed design, but rather to investigate the global behaviour of a

high-rise RC structure under wind and seismic lateral loading. The structure considers typical UK high-

rise building design procedures of design to gravity loading and wind loading, but demonstrates theimplications of seismic design factored into the typical UK design procedure. Although seismic design

is not common practice for the structural design of regular high-rise structures within the larger UK

area, the effects of low seismicity may pose some threat of exceeding performance criteria outlined in

EN standards. The predominant aim of the project is to perform an analysis of the global structural

behaviour under wind and seismic lateral loading, and check the compatibility of the expected structural

performance to serviceability limit state checks such as overall deflection, storey drift and interstorey

drift ratio, as well as considerations into human comfortability criteria.

1.2 Project overview

1.2.1 Site location

The proposed high-rise building is part of a residential-led mixed-use development in the London

Borough of Southwark (LBS), to provide 144 units of new residential accommodation together with

530m2

of retail space. The proposed development comprises a 26 storey residential tower with ground

floor retail space including a full basement for cycle/plant and refuse storage. This 26 storey tower

(named A4 Tower) functions as the feature structure of the entire development, and is the high-rise

structure that this project pertains to. The site is located on Surrey Quays Road; in the LBS. Canada

Water Underground station and Canada Dock are located immediately to the south of the site, with the

River Thames approximately 300m to the north of the site. The A4 site is currently undeveloped and

being used for the site compound, whilst the adjacent residential plots are currently under construction.

A4 Tower is highlighted in the red triangle in figure 1.



5

1.2.2 A4 tower

The design for A4 tower comprises of a 26 storey tower to be formed with a reinforced concrete (RC)frame with flat slabs supported on columns and walls. The ground floor/mezzanine area will be for

retail space, whilst the remaining floors (1-24) upwards are of residential use requirements.

Furthermore, the roof area will provide a congregational terrace with open area seating and planters, for

occupants of the building to enjoy. The height of the building is 88m, with a height to width ratio of ~

4. Figure 2 details the architectural impression and a typical section of the building with corresponding

heights for each area.

For value engineering purposes and simplicity in construction, the design is to ensure that all the

columns are continuous from basement car park to roof level, thus removing the need for costly transfer

structures. The design will adopt a column grid of approx 5.0m2

with 200mm thick RC flat slabs. In

order to support the roof terrace area, the roof slab will be formed with a 250mm thick RC slab to allow

for additional loading. The central stair/lift core which is comprised of RC shear walls will provide the

lateral stability of the structure. It will also act as the main access channels for occupants of the

building. Detailed information regarding the scheme design, general layout and sizing of structuralelements can be found in the Structural Scheme design section of this report.

RC Lift/Stair Core

Ground

floor/Mezzaninelevel (0-8m)

Residential Floors1st-24th

(8-82.64m)

Roof Terrace with

tall parapets (82.64

– 88m)

Figure 1 - Site location and layout of Canada Quays A Sites

Figure 2 - Architectural impression and typical section detail (Courtesy of URS Corp. Ltd)



6

This project will only involve itself with the design and analysis of the superstructure, as an

investigation into the behaviour of a high-rise RC building. The predominant concern regards the

performance of the building under lateral loading, and the implications that the effects of lateral loading

will have on the structure in order to ensure the desired performance. Consideration will not be made

into building substructures (foundations), mechanical services and other building design issues that may

be integrated into the life-cycle of the whole structure.

2 Design processA high-rise/tall building can be defined as a buildings where in terms of structural considerations, its

strength and behaviour in terms of serviceability (deflections) is governed predominantly by lateral loads.

The lateral loads are a cause of wind and/or earthquake action upon the building. Although there is no

specific value that defines a tall building, a commonly acceptable dividing line is where the structural design

moves from the field of statics into the field of dynamics. It is therefore imperative to consider these aspects

in the design of the structure.

A typical characteristic of tall building design is the significance of all three design criteria in producing a

satisfactory structural solution: Strength, Serviceability and Stability. The principal contributing factor is thepresence of lateral loads, which increasingly dominates the structural form with increasing height. For low-

rise buildings, strength of individual components is the governing criteria however for buildings with

increasing height the global behaviour under lateral loading becomes increasingly important.

The project processes for A4 towers‘ analysis and design are as shown in the pert Chart found in figure 3. It

has been broken down into the 5 key stages of scheme design, primary analysis & design, Secondary

analysis & design, analysis interpretation, and technical reporting.

The scheme design process entails the interpretation of the provided architectural specifications. This

information will aid in producing a design scheme where the structural form and preliminary sizes of

structural members will be determined from. Furthermore the scheme design will require the determination

of the structural loading. Load cases and combinations will be determined through usage of Eurocode 1, and

will help determine self weights of construction materials, live occupancy loads and wind loading. Upon

completion of the scheme design, the creation of the model in the Finite Element Analysis software will

follow. This stage will also detail the assumptions and application of engineering judgement made in the

FEA design process.

Once the functional layout of the structure has been determined through the scheme design process, the

preliminary analysis process will follow. The predominant mode of analysis will be via the FEA model (in

Figure 3 – Project process overview



7

SAP2000). Furthermore the wind-loads will be applied to the FEA model, and used in conjunction to

determine the preliminary effects on strength and serviceability that they will incur.

Secondary analysis entails the determination of the effects of wind and seismic loading on the stability of the

structure. The effects of the wind loads from EC1 will form part of a more rigorous final analysis together

with an unconventional consideration into seismic effects in the UK. Then, checks on deflections and global

effects will be performed. This will typically involve investigation of the gravity loading effects, lateral

deflections and member forces. The seismic loading will then be determined from EC8, and a dynamic

analysis will be required to determine the induced lateral deflections. Hence, in the high-rise design process

a thorough knowledge of the structures modes of behaviour is a prerequisite to determine the global

behaviour of the finalised load bearing scheme. Therefore an investigation into the modal behaviour of the

structure will be used in conjunction with adequate wind and seismic analysis techniques.

The analysis interpretation process is essentially an ongoing function of the primary and secondary

stages of analysis. As the nature of the analysis becomes more complex with each differing applications

of load types, the process will necessitate iteration. The effects of loading may call for structural

alterations and even substantial rearrangement of members which in turn will necessitate a complete

review of the design. It is therefore anticipated that the various preliminary stages of analysis will berepeated a number of times before acceptable performance is achieved. Although the architectural

layout of the structure is not being modified for the purpose of this project, in reality, changes to the

buildings layout will be required as client‘s and architect‘s ideas of the building evolve. This in turn,

prompts the engineer to modify and reiterate the design process. It is therefore necessary to form an

appreciation of this stage.

The purpose of this report is invariably classed as the technical reporting stage of the project process.

Essentially this provides the client with the engineer‘s design intent, and justification of method.

2.1 Project literatureDue to the design focused nature of this project, the literature is largely based on structural design codes

and corresponding guidance literature. Involvement with these forms of literature is common practice

for all structural designers (within the public or private sector), and in the case of the required national

engineering practices are deemed as legal requirements that must be satisfied to produce safe and

structurally sound designs for human end-use. The literature that has aided this project process is as

follows.

2.1.1 Eurocodes

Eurocodes are a new set of European design codes for building and civil engineering projects. They

have been developed through combined engineering expertise from the European Union‘s member states over the past 30 years, with the intention of providing a framework for the standards used in

assessing structural products for CE marking. Eurocodes are intended to be the mandatory design

guidance for all European public works, and likely to become de-facto standards for the private sectorsof engineering and construction. Eurocodes, act to ensure uniform levels of safety for construction in

Europe. They also form a common and transparent basis for fair competition in the civil engineering

business sectors. Furthermore, they facilitate the exchange of construction services and broaden the use

of materials and structural components within the EU (Eurocodes Expert, 2007).

Primarily, the Eurocodes are intended for the structural engineer and have to be included in the design

and calculation process of buildings and all other types of structures. Facets of consideration within a

typical project can include geotechnical aspects, structural fire design, seismic design, construction,

temporary structures and much more. However, due to the involvement of other parties in the Civil

Engineering project lifecycle (e.g. geotechnical engineer, project architect, building services engineer,

quantity surveyors), Eurocode guidance will also play an important role in their design considerations in

the project. The Eurocodes are a series of 10 European standards, each of many parts, and related asshown in figure 4.



8

EC0 - EN 1990: Basis of structural design

EC1 - EN 1991: Actions on structures

EC2 - EN 1992: Design of concrete structures EC3 - EN 1993: Design of steel structures

EC4 - EN 1994: Design of composite steel

and concrete structures

EC5 - EN 1995: Design of timber structures

EC6 - EN 1996: Design of masonry

structures

EC7 - EN 1997: Geotechnical design EC8 - EN 1998: Design of structures for

earthquake resistance

EC9 - EN 1999: Design of aluminium

structures

Furthermore, the codes are utilised with the National Annex (NA) of the concerned member state of the

projects location. The National Annexes permits each member state to take into account its own local

differences concerning geography, climate and traditional building practices. The safety level however

remains the responsibility of the government of each member state and differs within each state.

Whenever the EN Eurocodes are used for a structure the National Annex of the state in which the

structure is built, must be utilised. Countries adopting the EN standard are responsible for issuing a setof National Annexes (NAs). This defines the national decision on the situations within each Eurocode

where national choice is permitted. The NA therefore gives the national values for nationally defined

parameters (NDP‘s), which country specific data such as seismicity maps and the national decisions on

whether or not informative annexes to the Eurocode may be used. References may also be made to non-

conflicting, complementary information issued to support the application of the Eurocode.

A prerequisite to the use of design codes is the engineers understanding of structural engineering

fundamentals and methods of analysis, to supplement the guidance of the project. The codes utilised in

this project are as follows.

2.1.2 Eurocode 1 - EN1990 - Actions on structures

The intention of any structure is to provide a load-bearing system which safely and serviceably providesa form fit for end-use by its intended occupants. A fundamental stage of the design process is

determining the loads that must be resisted by the structure, and Eurocode 1 provides comprehensive

information on all actions that should normally be considered in the design of buildings and civil

engineering works. EC1 comprises of ten different parts which detail general and specific actions

induced by all forms of loading in civil and structural works. However, for the requirements of this

project, the parts of Eurocode 1 that will be utilised are:

Part 1-1: General actions – Densities, self-weight and imposed loads

Part 1-3: General actions – Snow loads

Part 1-4: General actions – Wind actions

Part 1-1 is used to determine the dead loads of construction materials and finishes and imposed

occupancy loads on the structure. The load schemes will be made apparent in the Scheme design stages.

Figure 4 - Relation of Eurocodes to each other (IStructE manual to EC1)



9

The snow load from part 1-3 will also be considered here as part of the roof gravity loading.

Part 1-4 will be a major factor relating to the behaviour of the whole structure, as it will aid in

calculation of the applied wind loads which are a major concern for lateral stability of tall buildings.

EC1 covers guidance for structures in the UK of height up to 200m, and it is from this code that the

wind loading effects on the structure will be determined.

Distribution of wind on a tall structure is as shown in figure 5. Forces due to wind are generated on the

exterior of the building based on its height, local terrain roughness and the square of the wind velocity.

The weight of the building has no effect on wind load design (unlike seismic), but it is helpful in

resisting uplift forces that might be generated. Given that the structure has no large openings; all the

wind loads will be transmitted to the exterior surfaces of the structure. As shown in figure 5, the

formation of a positive pressure will occur on the windward face of the building, whilst negative suction

pressure will be exerted on the leeward walls and roof, with corresponding change in magnitude along

the width of the building. The lower positive wind pressure near the structures base on the windward

elevation is due to the frictional effects of the surface on the flow of the wind, which diminish as

effective height of the building increases from base. Positive wind pressures acts inward on the

windward side of a building and suction forces act outward on most other sides and most roof surfacesdue to openings (e.g. windows/doors). Special concentrations of outward force, due to aerodynamic lift

occur at building corners and roof edges, particularly overhangs or parapets (BRE, 1994). Therefore the

overall structure is designed for the sum of all lateral and uplift pressures and the individual parts to

resist the outward and inward pressure concentrations. Furthermore, an internal pressure (wi) will be

formed in the building due to any opening, whether they are intended openings or an occupant leaves an

external door or window open for a prolonged period of time. Therefore, the net pressure of the wind

load acting on the structure will consider the sum of the external (we) and internal (w i) pressures acting

on the surface of the structural frame.

The roughness of the earth‘s surface creates drag which converts some of the wind‘s energy intomechanical turbulence. Since turbulence is generated at the surface, surface wind speed is much less

than wind speed at high levels. Hence the wind load distribution shown in figure 5. For strong winds,

the shape of wind speed profile depends mainly on the degree of surface roughness, caused by the

overall drag effect of buildings, trees, and other projections that impede flow of wind at the surface

level (Taranath, 2010), and thus the exhibited wind pressure profile. It is also the reason why EC1 lists

terrain categories with corresponding values for terrain roughness. Frictional effects on velocity profilewill be larger in ‗town terrain‘ due to surface interference from the built environment, whilst in

Wind Wind

PLAN

ELEVATION

Figure 5 - Wind load distribution on a tall building



10

countryside areas the wind speeds will be greater due to lack of a built environment. Furthermore, the

distance of the structure from sea shoreline is considered, as the winds of highest speed will occur at or

close to sea due to the minimal frictional interference of the ocean surfaces with the wind flow. This

will diminish the further inland the structure is sited.

For tall flexible buildings subject to wind load, assessment of oscillatory movements due to fluctuating

winds is imperative. Such movements can induce responses in humans ranging from mild discomfort to

acute nausea or motion sickness. There is no code specific guidance or regulations stating acceptable

limits for comfort criteria, yet the unsatisfactory performance of a large investment such as a tall

building can direly affect the occupants/owners of the structure. The principal factor governing the

degree of human comfort is the acceleration of the building. An accurate method to determine

exceedance of certain levels of comfort is through wind-tunnel testing of the model; however a

preliminary manual investigation into exceedance of comfort criteria will be made.

2.1.3 Eurocode 2 - EN 1992: Design of concrete structures

The principal design code that will be used for project purposes is Eurocode 2 (EC2). EC2 applies to the

design of buildings and civil engineering works in plain, reinforced and pre-stressed concrete. It

complies with the principles and requirements for the safety and serviceability of structures, the basis of their design and verification that are given in Eurocode 0 EN 1990: Basis of structural design. EC2 is

only concerned with the requirements for resistance, serviceability, durability and fire resistance of

concrete structures. Other requirements, e.g. concerning thermal or sound insulation, are not considered.

Only ‗Part 1-1: General – Common rules for building and civil engineering structures‘ is utilised for this project. This will provide the general structural design basis of the RC frame, and will be used in

conjunction with other design and analysis methods for tall building design.

Structural performance of tall buildings

The structural form adopted is one in which the lateral stability will be provided predominantly by the

central lift/stair core comprised of RC shear wall, with the flat slabs supported on each floor byconnection to the inner core and interior & edge columns. Traditionally, the majority of lateral load-

resisting systems can be classed into three basic categories. Shear wall systems, frame systems and

shear wall-frame systems (a hybrid of the previous two). A4 tower will act as a shear-wall frame

system; however the shear wall is intended to resist the majority (if not all) of the horizontal loads in the

system, with the external columns providing supplementary support as external rigid frame members.

A prerequisite for desirable response of the structure to horizontal loads is to interconnect all lateral-

force-resisting components with a relatively rigid diaphragm surface. This is achieved with the use of

floor and roof systems, which generally possess large in-plane stiffness. This diaphragm action will be

provided by the RC flat slabs in the case of this structure, whilst the vertical elements (Shear wall and

columns) will contribute to the total lateral force resistance (in proportion to their own stiffness). The

function of the diaphragm is to transmit the inertia forces generated by the wind/earthquake

accelerations of the floor mass at a given level to the horizontal load resisting members (Taranath,

2010). The column members interacting with shear walls will provide the necessary resistance to lateral

forces, while each member carries its appropriate share of the gravity load. The lateral stiffness of the

structure may be further enhanced by adjusting the orientation of certain vertical members with the

direction of their major sectional axis in that of the horizontal load direction.

The flat-slab method system (figure 7) has no beam elements except for at the balcony edges, and so to

resist the moment couple, the diaphragm must act as a deep plate resisting both bending and shear

forces. Either type of the diaphragm behaviour requires effective transfer of bending and shears forces,

necessitating careful detailing of connections between the diaphragm and the lateral support systems.



11

The flat slab structure is essentially one of the simplest structural forms that can be adopted for high-rise design. The flat soffit of the slab and column connection requires uncomplicated framework, and

means that the soffit can be used as the ceiling as due the requirement for minimal ceiling finishing.

However at the detailed design stage, considerable thought must be given to punching shear capacity of

the slab/column connection. Punching shear occurs in a two-way concrete flat slab due to the lack of

beams in the slabs connection to the columns. It is the tendency of the slab to drop as a unit around the

column, as shown in figure 6. The column essentially ―punches‖ through the slab, with crackingappearing on the top surface of the slab due to the high stresses. Avoidance of this collapse mechanism

is imperative, so as to not to affect the load distribution of the structure and diaphragm action that the

affected would exhibit whilst functioning normally. Punching shear can only be avoided through

adequate reinforcement detailing and the considerations for this would be outlined in the later design

stages of a project.

For the adopted shear wall-frame system, we can essentially consider the behaviour of the structure

under lateral load as a corresponding hybrid of the two components. As the core walls are

interconnected in a regular tube-like form with three axes of symmetry (in the simplified model), the

behaviour will be similar to that of a thin-walled beam section cantilevering off the base of the

structure. The expected deformation due to resistance of horizontal loads for a simplified planar shear

wall structure of 7 storeys is as shown in figure 8. The shear wall is continuous down to the base to

which it is rigidly attached, thus forming a vertical cantilever. Shear walls offer high in-plane stiffness

and strength, whilst also carrying gravity loads from the floor slabs at the interconnected storey levels.

It is usual to locate the gravity loads at a position on plan so that they attract dead loading sufficient to

suppress the maximum tensile bending stresses.

However, the name ―shear wall‖ can be misleading as it may imply that the walls deform predominantlyin shear. Shear deformation occurs due to the fact that large fractions of the lateral load shear forces are

carried in the shear wall structure and therefore exhibit the shear deformation shown in figure 8b. This

is not necessarily the case, as shear walls predominantly deform in the flexural mode shown in figure 8c

due to the behaviour of it like cantilevering beam in which the bending resistance is governed by the

large in-plane stiffness of the entire wall structure. Furthermore, the flexural stiffness can be increased

by the arrangement of shear walls in L, T, I, U or closed shaped sections (such as the core of A4‘ssimplified model).

Figure 6 - Punching shear failure of RC flat slab

(Taranath, 2010)

Flat slab

Figure 7 - Typical flat slab section

Column



12

Taking the rigid frame component of the structural system, the behaviour of a rigid frame under lateral

loading can be demonstrated as shown in figure 9 for a 7-storey 3-bay 2D frame with infinitely rigid

diaphragms. The horizontal stiffness of a rigid frame is governed by the bending resistance of thecolumns and their connections. In a sufficiently tall frame, lateral bending resistance is provided by the

axial rigidity of the columns. As shown in figure 9b, the applied lateral loads induce storey shear forces

which are resisted by the columns in that storey. This shear leads to a double curvature deformation of

the columns in that storey, with a point of contraflexure approximately at the mid-height of the

columns. The global deformation produces a sway shape of the structure. However, it is not represented

in scale for fig. 9b that the largest sway deformation of each storey occurs is at the base, and tends to a

minimum at the uppermost storey. This is due to is the shear forces at an individual storey being equal

to sum of the lateral forces imposed on the structure above the storey that is being considered.

Alternatively, the overall moment of the lateral load is resisted at each storey by the couple resulting

from the axial and compressive forces in the columns on opposing sides of the structure (figure 9c).

These axial forces lead to extension (on tensile side) and shortening (on compressive side) of thecolumns, which results in the overall bending and associated horizontal displacements. As a result of

the cumulative rotation up with the height of the structure the storey drift increases with height (as

opposed to that in figure 9b). The contribution to total storey drift from overall bending may in the

highest stories exceed that of the shear contribution to drift, however the contribution to overall drift of

the structure from overall bending will typically not exceed around 10% of that of the sway motion

(except in very tall and slender frames). Hence the predominant mode of deformation for rigid frames is

typically that of figure 9b (Smith, 1991).

a) Non-deformed shape b) Shear deformation shape c) Flexural deformation of

shear wall

TensionSide

NA

b) Shear deformation – Sway mode

c) Flexural deformation

Vcol

Tension Compression

NA

Figure 9 - Lateral load deformation of rigid frame

a) Non-deformed shape

Figure 8 - Lateral load deformation of shear wall



13

Combining the two systems to demonstrate the expected behaviour of the structural mechanism in use

for Tower A4, the expected behaviour of the shear-wall frame structure under horizontal loading is as

shown in figure 10. This is demonstrated by considering the predominant deformed shapes of each type

and superimposing their behaviour with each other. When the structure is loaded laterally the two

different deflected forms of the wall and the frame are made to interact horizontally through the slabs at

each storey level. The form shown in figure 10c is the deflected form of the two acting together in a

symmetric wall-frame system. As a consequence of this, the distribution of lateral loads on the wall-

frame system may very well be different from the distribution of the external lateral loads due to the

different predominant forms of resistance each individual system component offers.

It is common practice for such systems to be designed so that it is assumed that all horizontal loading

will be resisted by the shear core only, whilst the rigid frame component of the RC structure will only

take gravity loading. It must be noted that wall-frame systems offer advantages in relation to the amount

of horizontal loading they are subjected to. The previous design assumption can prove costly compared

to considering the interaction of the two systems together. This assumption can lead to overly largeshear wall system with gross reinforcement requirements, and thus it is worth considering a more

rational approach to detailed final design. The advantages on the preliminary design of considering the

interaction of the two are as such that:

The estimated storey drift may be less than if the walls were considered to be the only

horizontal load resisting members.

The estimated bending moments in the walls/cores will be less if than if only wall action is

considered alone.

The columns of the frames can be designed as fully braced, limiting requirements to design for

considerable secondary effects (i.e. PΔ effect) The estimated shear in the frame will likely be uniform throughout the height, and as a

consequence of this can be designed and constructed repetitively and economically.

Furthermore, if the structure is subjected to twist, the torsional stiffness of the core can be a significant

component of the total torsional resistance of the entire building. Therefore as the core twists, the

original plane sections of the core warp. Due to the relative proportions of the height, length and

thickness of the entire walls, the torsional behaviour of this structure is similar to that a thin-walled

beam which is cantilevering vertically from its base connection (Smith, 1991). As warping at the base

is prevented by the connection to the foundation, the warping will induce vertical warping stresses and

strains throughout the height of the core walls, and would typically be considered in the detailed design

stages of the project. However, considering the regular and tube-like nature of the structure Torsional

stiffness is assumed as adequate. Hence, the project will not be considering the twist action of the

structural form due to project time-constraints.

Point of

Contraflexure

Flexural

Shape

Shear

shape

a) Flexural Deformation

of Shear wall

b) Shear deformation of

rigid framec) Combined deformation

Figure 10 - Lateral load behaviour of shear wall-frame structure



14

2.1.4 Eurocode 8 - EN 1998: Design of structures for earthquake resistance

Complementary to Eurocodes 1-7 and 9. EC8 states additional provisions for the structural design of

buildings and civil engineering works to be constructed in seismic regions where risk to life and/or risk

of structural damage are required to be reduced. ‗Part 1-1: General rules seismic actions and rules for

buildings‘ will be utilised for the purpose of this project, where general requirements and rules for

assessment of seismic actions and combinations with other actions will be detailed.

EC8 and supplementary texts state that seismic design of a structure is a subtle interplay between

balancing the seismic capacity of structures, with the expected seismic demand to which they may be

subjected. Essentially it is the mitigation of seismic risk. In areas of moderate to high seismicity these

risks could be defined as the possibility of losses of human life, social welfare or economical, due to the

effects of future earthquakes. Whilst for an area of low seismicity this could be minor damage and

gradual deterioration of the structure which could lead to restrained use, increased maintenance and a

shortening of the intended design life.

Earthquakes can exhibit a whole range of damaging effects on a particular location, including things

such as soil liquefaction, landslides, surface rupture, tsunami‘s and more. It is however shown that the

most common cause of damage on a global scale is via earthquake induced ground shaking (Bird andBommer, 2004). This form of this ground shaking hazard is predominantly described as the probability

of exceeding a specific level of ground shaking within a given time. This is known as a probabilistic

seismic hazard analysis (PSHA).

The assessment of ground shaking hazard due to potential future earthquakes involves three stages. The

first is the development of a seismicity model for the location and size of the future earthquakes in the

investigated region. The second is the development of a model defining ground-motion, which predicts

the expected level of shaking at a given site as a result of the investigated earthquake scenarios. Finally

the third stage is the integration of these two models, into a model for the expected levels of shaking at

the site of interest (Bommer and Stafford, 2009). Figure 11 provides a schematic overview of seismic

hazard analysis. Where the seismicity model defines earthquake scenarios of magnitude M, at a distance

R from the site of interest, whilst the ground-motion model probabilistically determines the shakingparameter of interest for the particular M-R combination employed for design. The results are then

expressed in terms of acceleration response spectra as shown in the integrated model.

The study and determination of seismicity models is a complex realm, which in itself could fill the

pages of numerous texts. In essence the concept behind the determination of the model is through

historical measurement and calculation of the earthquakes epicentre. The majority of earthquakes occur

due to sudden rupture of geological faults which release strain energy stored in the surrounding crust of

the earth‘s tectonic plates. This released energy radiates from the fault rupture in the form of seismic

waves, where the location of the earthquake is specified by the location of the focus (hypocentre). The

Epicentre is simply the projection of the hypocentre on the Earth‘s surface, detailed in geographical

Figure 11 - Three stages of seismic modelling (Bommer and Stafford, 2009)



15

coordinates. The common practice of measuring these earthquakes over historical occurrence is down to

usage of sensitive measuring equipment, and although the readings are used to pinpoint the earthquake

to a specific location, it must be noted that earthquakes in reality can occur from very large sources. The

magnitude of the earthquake is simply a measure of the energy released in the form of seismic waves

measured in a variety of different scales but typically using the Richter scale, ML.

The seismicity model then needs to specify the expected location and probability of future

earthquakes of differing magnitudes, and is often compiled from regional earthquake catalogues. In the

case of EC8 and the corresponding NA‘s this will be from catalogues of earthquakes in the UK through

instrumental and historical recordings. The integrated model will aid in estimating the ground motions

that will be induced by the earthquake, and the related inertial loads that will be exhibited upon the

structure.

Response of the structure to ground motion

The behaviour of a building during an earthquake is a vibration problem. Seismic ground motion does

not damage the structure by impact (through external pressures, such as wind), but rather through

internally generated inertial forces caused by the vibration of the structure. Increasing the mass of the

structure will have undesirable effects in earthquake design. Firstly this will result in an increase of the

force the structure is subjected to, and secondly it can attribute to buckling and crushing of columns andwalls when the mass of each floor pushes downwards, or moved to an eccentric position through

shaking. This secondary effect is known as PΔ and is proportional to the movement of each floor/mass.Tall buildings respond to seismic forces differently to low-rise buildings. The magnitude of the inertia

forces induced during earthquakes depends on the building mass, ground acceleration, the nature of the

foundation and soils, and the dynamic behaviour of the structure. If the building and its foundations

where infinitely rigid, then it would exhibit the same acceleration as the ground, resulting in an inertia

force of F = mẍg for a given ground acceleration ẍg. However tall buildings exert a degree of flexibility

in their behaviour to seismic loading, and therefore this inertial force tends to be less than the product of

F = mẍg. Due to this flexibility, high-rise buildings experience much lower accelerations than low-rise

buildings, but if a high-rise building is subjected to ground accelerations for a prolonged duration, then

it may experience much larger forces if its natural period of vibration is closer to the period of the

ground acceleration. This phenomenon is known as resonance, and it is this that must inherently beavoided. Hence, the magnitude of lateral force exerted on the structure is not only a function of the

acceleration of the ground, but also influenced to a great extent by the type of response of the structure

itself and its foundation as well. This interrelationship of building behaviour and seismic ground motion

also depends on the building period determined through the modal analysis, and is formulated in the

response spectrum method detailed further on.

The intensity of the ground motion essentially reduces with increase in distance from the epicentre of

the quake. This reduction, known as attenuation occurs at a faster rate for higher-frequency components

Figure 12 - High-rise building behaviour during earthquakes (Taranath, 2010)



16

(than for lower frequency components). Although the cause of the change in rate is too complex to be

fully detailed, its existence is certain in the seismic design world (Taranath 2010). This is a significant

factor in the design of tall buildings, as even if the structure is located farther from the epicentre of the

quake than a low-rise building, it may experience greater seismic loads due to short-frequency

components of the structure not being attenuated as fast as the short-period components. As the building

vibrates due to ground motion, resonance will occur if the period of the building coincides with the

period of waves transmitted through the soil. The natural periods of the soil change with soil type and

characteristics, but typically range from 0.5-1.0s. EC8 makes provision in response spectrum analysis as

to the effects of different soil types upon the structural response. The obvious design strategy is to

ensure that buildings have a natural period different from that of the expected ground vibration, so as to

prevent amplification of the structures response.

Basic Dynamic properties of a structure

For linear dynamic analysis, a structure can be defined in terms of its stiffness, k , its mass, m, and

damping, c. The oscillatory behaviour of the combination of mass and stiffness needs consideration to

understand the behaviour under dynamic loading. Under application of a force, k is a constant of

proportionality between the force and the displacement x (such as in Hooke‘s law). For a simple single

degree of freedom (SDOF) system the behaviour of a structure displaced from equilibrium will generatea restoring force equal to stiffness x displacement (kx). This force accelerates the structure back towards

its equilibrium position with an acquired momentum of mass multiplied by velocity. The oscillatory

nature of this restoring forces causes the structure to overshoot and so the sign of the process reverses

and is repeated in the opposite direction, causing the structure to oscillate about its equilibrium position.

It must be noted that the process can be defined in terms of energy, where the vibrations induce a

repeated transfer of strain energy into kinetic energy as the structure oscillates around its unrestrained

position.

However, this energy will be gradually dissipated by the motion of the structure via an array of internal

mechanisms. This dissipation of energy is characterised as a grouping of the internal mechanisms which

result in this, known as damping which is denoted by the coefficient, c. Damping, c, is the coefficient of

proportionality between force and velocity, however it is best termed as a ratio ξ, where;

The damping of a building is dependent on the construction materials, the types of connections and the

influence of non-structural elements on the stiffness characteristics of the building. Damping is

attributable to internal and external influence sources. Chief among them are:

External viscous damping due to surrounding air of the building. However this is often

negligible due to the low viscosity of air.

Internal viscous damping due to material viscosities. This is proportional to velocity of the

structure under excitation.

Frictional damping occurring at connections and support points of the structure. This is a

frictional constant, irrespective of the velocity or displacement.

Damping due to a large part of the energy absorbed due to ductility of members within the

structure (Hysteretic damping).

The combined behaviour of these mechanisms are grouped together into the dimensionless parameter ξ.Damping c, is an onerous value that is very difficult to determine. Values are traditionally based on

experience and empirical knowledge of structural behaviour. However, within Civil Engineering the

values of ξ can range from 0.01 to 0.10, with the low-end values typically used for wind, while those of

the upper end are for seismic design (Taranath, 2010). Furthermore, the analysis methods are based on

the assumption of viscous damping.

Each possible displacement of the structure is a degree of freedom, and a real structure with distributed

mass and stiffness has an infinite number of degrees of freedom. It is therefore the concept of



17

considering the dynamic behaviour of the structure using ‗lumped parameter models‘. This is where thedistributed mass, stiffness and damping are modelled as a number of discrete masses, m, connected by a

series of springs with stiffness, k, and with dashpots representing damping. For purpose of detailing the

theory, the simple SDOF structure is shown in figure 13.

From Newton‘s second law of motion, it is evident that the resultant force, F, is equal to mass , m,

multiplied by acceleration , ẍ. Therefore for a forced vibration varying with time, t, can be equated to:

F or inertia force + damping force + stiffness force = external force

Where ẋ is velocity and x is displacement of the system. In the case of representing a seismic action, thestructure is not forced directly, but rather the motion of the ground beneath it is subjected to a

predominantly horizontal time varying motion, as in figure 14.

Due to the absence of external forces, and considering the relative displacement of the structure to the

ground from a fixed datum, the previous relationship will be defined as

.

And by considering the relative displacement between the ground and the structure as y = x - xg , this

can be further expressed as

.

However, the free-vibration behaviour of the structure must be considered for so as to understand the

fundamental modal behaviour under earthquake excitation. Considering a SDOF system without

damping included and no external force (free vibration), the equation of motion is simply .

Displacing this mass from equilibrium position, it will undergo free vibrations at a rate known as the

natural frequency. The solution for non-damped free of a SDOF system is given as:

ωn is the circular frequency measured in (radians/s), and denotes the cyclic frequency of the structure as

that of the angular speed of an equivalent circular motion. More easily visualised, and of more

importance to the structural engineer is the visualization of this parameter as the natural frequency of

the structure, f n (measured in Hz), and the natural period Tn (time taken to complete one full cycle of oscillation and measured in seconds). Therefore

Figure 13 - Forced vibration of mass-spring damper system (Elghazouli, 2009)

Figure 14 - Base vibration subjected on a mass-spring-damper system (Elghazuoli, 2009)



18

Considering with damping (where ), the solution for natural period, it is evident that

the behaviour of the system will be dependent on the relative magnitudes of c, k and m. Assuming the

system is critically damped, factoring in the damping ratio, then the displacement can be denoted as:

The response of a SDOF structure with T n =1, under free-vibration with release from an initial unit

displacement over time that is damped (i.e. ξ = 0.05), and undamped (ξ = 0) differs as shown in the

graphs in figure 15.

The oscillations are multiplied by a decay term, where the greater the damping, the quicker the

oscillations will die away. Furthermore the natural frequency is altered by the factor , however

practical values of this are often close to unity and so it is acceptable to neglect damping when

calculating natural frequencies (Williams, 2009). By considering the relationships between of ωn, ξn, m,

c, and k, the following equation is gained

However, not all structures can be modelled as SDOF systems, as civil engineering structures in their

complex nature are multi degree-of-freedom (MDOF) systems. This is as the distributed masses and

stiffness within the structure may undergo significant deformations in several modes of vibration. The

solution of MDOF systems warrants usage of computer methods due to the complexity of their nature in

a typical design scenario. The equation of motion for a system of N degrees of freedom can be given as:

Where [m], [c] and [k] are the mass, damping and stiffness matrices with dimensions N x N, y is the

displacement vector and is an N x 1 influence vector containing vectors corresponding to the DOF‘s inthe direction of the earthquake load and zeroes elsewhere. The stiffness matrix can be determined asfor a static analysis, and is a banded matrix. The mass matrix m, is typically determined through

division of the mass of each element between the nodes of the model, giving a lumped mass matrix

which contains only diagonal terms (CSI, 2009).

Modal behaviour

A prerequisite to understanding the behaviour of a structure under dynamic lateral excitation (whether

wind or seismic) is the determination of the structures mode shapes. Modal analysis is used to

determine the free-vibration modes of a structure. The free vibration modes are determined by taking

into account the overall mass and stiffness of the structure, to find various periods at which the structure

will naturally resonate at. These natural modes of vibration provide an excellent insight into the

behaviour of the structure, and are used for the basis of response spectrum analysis.

Figure 15 - damping effects on free vibrations (Williams, 2009)



19

Considering the undamped free vibration of an SDOF system, we know that , whilst the

solution to this equation is

SAP2000 was utilised to perform the Modal analysis of the structure, thus determining the fundamental

periods and other corresponding information of each mode of vibration for A4. SAP2000 was used to

determine the undamped free-vibration modes of the structure using Eigenvector analysis. Eigen vector

analysis involves the solution of the generalized eigenvalue problem for a MDOF system:

Where k is the stiffness matrix, m is the diagonal mass matrix, ω2is the diagonal matrix of eigenvalues

to be solved for natural frequencies, and Ф is the matrix of corresponding eigenvectors (mode shapes).

Each eigenvalue pair is called a natural vibration mode of the structure, and the modes are identified by

numbers from 1 to N in order of which they are determined from the program (CSI, 2009). In solving

this, we are given N number of circular frequencies corresponding to an associated mode shape. Each

mode of the structure has a distinct deformed shape with a particular natural frequency (or period) at

which it occurs. The modes of vibration of a structure are a function solely of position within the

structure, and independent of external loading. Typical horizontal modes of vibration for a four storeyframe and a shear core structure (vertical cantilever) are as shown in figure 16.

The eigenvalue is the square of the circular frequency, ω, for that particular mode. The cyclic

frequency, f, and period T of the mode are related to ω by

2.2 Design guidesDesign manuals are often published by national engineering institutions who act on behalf of the

Engineering profession within their regions. The intentions of the design guides are to provide

engineers with literature and guidance in familiarisation and prompt utilisation of the latest engineering

codes. The following design manuals provided by the Institution of Structural Engineers (IStructE) have

been used as aids in this project.

Figure 16 - Mode shapes of a four storey building and shear wall

Mode 1 Mode 2 Mode 3



20

2.2.1 IStructE - Manual for the design of building structures to Eurocode 1 andBasis of Structural Design

This Manual provides guidance on the design of building structures which rely on actions determined

from all parts of EC1 and EC0. Given the complexity of loadings (actions) and load combinations in the

it is intended that computer analysis through the FEA package will determine the load effects of the

structure. The manual will be used to facilitate the generation of load cases and load combinations forthe structure through hand analysis methods. These will then be used to aid in the verification of the

computer outputs.

2.2.2 IStructE - Manual for the design of concrete building structures to Eurocode2

The intention of this manual is to provide guidance on the design of reinforced and pre-stressed

concrete building structures that do not rely on bending in the columns for their resistance to horizontal

forces and are also non-sway. Therefore it is immediately notable that there is only so much that

guidance this manual will offer. As the nature of the project structure is tall and will require

considerable design into resisting horizontal loads, this manual will only be used for the design of slab

members for gravity loading, and preliminary calculation into the behaviour of column and core

members under gravity loads. The guide is not intended to be a substitute for the greater range of EC2‘sparts. It is laid out for hand calculation of structural behaviour, but the procedures are equally

applicable to computer application of structural design methods.

Among many assumptions the scope of the manual covers, the key ones are that:

The Manual has been drafted to comply with BS EN 1992-1-11 (EC2 Part 1-1) and

BS EN 1992-1-22 (EC2 Part 1-2) together with the UK National Annexes.

The assumed design working life of the structure is 50 years

The concrete used is of Normal weight, and up to characteristic cylinder strength (f ck, cyl =

50N/mm2)

2.2.3 IStructE - Manual for the seismic design of steel and concrete buildings toEurocode 8

The intention of this Manual is to provide seismic design application rules for the majority of low to

medium rise steel and concrete buildings falling within the scope of EC8 part1-1, for all levels of

seismicity. However, it is stated that the manual may still be useful for some aspects of the preliminary

design of a wider range of buildings, for example high rise buildings or buildings required for a post-

earthquake emergency. However is states that for detailed design, engineers will need to refer to EC8.

In reality the integrity of a building under seismic loading depends not only on the performance of the

superstructure itself, but also very strongly on the behaviour of its foundations and supporting soils.

Hence, suitably qualified and experienced geotechnical engineers would be working together with the

structural engineer(s) to ensure successful project delivery. As this project only involves itself with the

design of the superstructure irrelevant of the geotechnical and foundation design aspects, certainassumptions and modifications have been made to allow for the exclusion of these parts of the project

process. All assumptions and modification are detailed in the corresponding design stages.



21

3 Scheme Design

3.1 General ArrangementThe Structure is to be a reinforced concrete frame adopting a flat slab layout. The flat slabs will be

connected to columns with plain soffits to remove the presence of intrusive ceiling details for theresidential portions of the structure, as well as provide ease in construction due to there being no

requirement for complicated shuttering or casting techniques. The flat slabs will be of 200mm depth for

each floor, except for the 250mm roof terrace floor. The typical first floor arrangement is as shown in

figure 17.

The central stair/lift core acts as the access route for all floors from Ground to Roof. The Layout of each

floor is of hexagonal form, with the smaller edges of the hexagon being where the balcony areas of each

apartment will be. The longer edges of each hexagonal floor are 22.5m lengths, with edge columns

spaced at 4.5m bay width. The smaller balcony edges are 6.25m in span, and are supported by beams

between the outermost edge columns. Furthermore, the interior spans are supported by three internal

columns arranged in the formation of a regular triangle centred at the center-point of the grid, with edge

lengths of 15.75m. The largest spanning areas of the flat slab between columns to shear wall are

approximately 6m in span. The internal columns from ground to 1st floor underside are circular

sections, whilst the columns for all residential floors (1-24th) have been sized as blade columns to fit

within internal / external walls detailing so as to provide non-intrusive architectural detailing due to the

residential use. However, simplifications have been made to the original architectural layout (see figure

18).

Figure 17 - 1st floor general arrangement

200mm Deep

RC flat slab

Internal blade

columns

Edge column

Supportingbeam at balcony

edge.

Lift/Stair Core



22

Due to project time constraints and the highly detailed form of elements such as the shear wall, the

structural form has been simplified for the project purposes. Therefore the simplified scheme in figure

18 has been adopted for a representation of the uniformly expected global behaviour of the project.

Detailed design to the original structural form requires lengthy consideration and would be worked on

by a team of numerous engineers with balanced time constraints and in-depth project management.

The simplified scheme adopts the same geometric form, but the only changes are the simplified profile

of the core and that the blade columns from 1-24th

have been equated into circular sections of similar

cross-sectional area. The simplified core is of a similar cross-sectional area to that of the original

complex form. However, the intention of the simplified and regular form is to provide symmetry and

regularity in plan. This regular floor form will simplify the analysis, particularly for horizontal loading

in which it should exhibit uniform behaviour in-plane to any three of the axis of symmetry it provides.

Furthermore, the simplified design does not account for the proposed basement floor. Instead the

building is to be designed without basement floor, and the ground floor is instead connected at base of

columns to foundations. As stated previously, the foundations not be included in the analysis. Rather, it

will be assumed that it‘s bending and membrane stiffness is infinite and that any settlement can be

ignored. Sizes of structural members are detailed in the member sizing section of this document.

3.2 Occupancy loadsFrom EC1 ―Part 1-1: General actions — Densities, self-weight, imposed loads for buildings‖, for theproposed different uses of the building, the following occupancy loads in table 1 have been determined

from the categories in table ‗6.1 Categories of use‘ in section ‗6.3 Characteristic values of Imposed

Loads‘. It must be noted that the simplified profile of the building, means that the occupancy loads for

the retail area will in fact be imposed upon the ground floor. As the ground floor is not part of the

superstructure, but rather that of the foundation/ground floor slab system, the retail section is therefore

not accounted for in this in analysis. The retail loads will be taken as part of the slab gravity loads.

Table 1 Lists the live occupancy loads for each floor type., whilst figure 19 details the simplified

elevation.

Simplified

central core

Internal

Circular

Column

200mm DeepRC flat slab

Supportingbeam at balcony

edge.

Edge column

Figure 18 - Simplified general arrangement of an individual floor.



23

Table 1 - Imposed Occupancy Loads

Floors Storey(s) Category of Use Occupancy

load (kN/m2)

Retail Ground D1: Areas in general retail shops 4.0

Residential 1st – 24

th A: Areas for domestic and residential activities 1.5

Roof

Terrace

Roof C3: Areas without obstacles for moving people, e.g.

areas in museums, exhibition rooms, etc. and access

areas in public and administration buildings, hotels,

hospitals, railway station forecourts.

4.0

3.3 Dead loads

3.3.1 Slab dead load

Eurocode 2 specifies the bulk density of concrete is to be taken as 25kN/m3. This will be the

predominant contributor to the weight of the structure. However, other structural elements and finishes

must be considered in the load combinations used in design. Unit floor dead loads will be taken as

shown in table for a typical internal RC slab section as shown in figure 20.

Floor Component Weight x Thickness Dead Load (kN/m2)

40mm Screed 22 kN/m3

x 0.05m 1.1

200mm thick RC Slab 25 kN/m3x 0.2m 5

Mechanical Services and Ceiling Finishes - 0.15

Internal Partitions - 1.0

Total Dead Load 7.25

This dead load value of 7.25kN/m2

load will be included in the ULS gravity design loads for all slabs.

As there are no finishes on the core wall or columns to be considered, the automatically computed

values for the concrete weight of 25kN/m3 will be taken from the FEA package. Further load cases andcombinations are detailed in the preliminary design stage of this report.

Ground floor mezzanine

hstorey = 8m

Residential Floors 1-24

(8.11-82.64m)

h = 3.11m

Roof Terrace (82.64 – 90.64)

Steel frame and storey height

excluded from sim lified model

200mm thick

RC flat Slab

50mm floor

Screed

Figure 20 - Internal RC slab components

Ceiling finishes

Figure 19 - Simplified elevation



24

3.3.2 Cladding perimeter load

The external wall construction will typically consist of an aluminium rain screen cladding system with

brittle glass panel windows. In reality, the system will be designed to transmit the horizontal wind

pressures acting on the cladding panels to the RC frame. A 6kN/m perimeter load has been taken toaccount for the cladding loads on the perimeter of the RC slabs. The deflections of the structure will

need to be to a minimum so that no crushing or stressing of the cladding members occurs under

deformation of the frame.

3.4 Load combinationsFrom EC0, the following load combinations for favourable and unfavourable action will be used in the

analysis. Where dead loads will be deemed as permanent action Gk , and live loads (occupancy, wind

and seismic) will be deemed as variable actions Qk . As the concern of the project is to determine the

overall performance of the structure to lateral loading, the predominant combination of actions will be

for the serviceability limit state performance of the structure. Serviceability limit state concerns the

functioning of the structural members under use and the comfort of people using the structure. The

strength considerations of the structure will be a larger concern in the detailed design stage for ultimatelimit strength (ULS) considerations. The verification of serviceability limit states should be based on:

1. deformations that:

affect the comfort of users

affect the functioning of the structure (including the functioning of machines or services), or

cause damage to finishes or non-structural members.

2. vibrations that:

cause discomfort to people

limit the functional effectiveness of the structure, or

cause damage to finishes or non-structural members.

3. damage that is likely to adversely affect:

the appearance

the durability, or

the functioning of the structure.

The following load combinations for lateral behaviour under favourable and unfavourable lateral load

will be used in the analysis to determine the deformation of the structure under ULS limit state loading

(1.35Gk + 1.5Qk ).

Favourable: 1.35Gk + 1.5Qk, occupancy + 1.0Qk, wind/seismic

Unfavourable: 1.35Gk + 1.5Qk, occupancy + 1.0Qk, wind/seismic

3.5 Structural durability and member sizing

3.5.1 Durability

From client specification, exposure conditions are taken as XC1 (dry or permanently wet). Therefore

minimum cover for reinforcement of the concrete is required as 25mm (Cobb, 2009). Minimum

required concrete type for XC1 conditions is C20/25 (f yk,cyl =20N/mm2), however the design will adopt

C50/60 (f yk,cyl =50N/mm2) concrete. C50/60 concrete was chosen for its higher compressive yield

strength than the required minimum, as due to the tall nature of the building the self-weight of thestructure will be a significant component of the compressive axial forces taken in the core and column



25

members. Furthermore, it is imperative that axial load capacities of the core/columns under gravity

loading are such as not to reduce moment capacity of the members when horizontal loading is also

induced.

Due to close proximity of A4 to the exit of Canada Water station, the bottom retail area of the structure

provides 2 hour fire resistance (R120) to enable adequate time and structural strength for evacuation.

For continuity, this level of fire resistance will initially be provided for the entire structure. To provide

the required R120 fire resistance, from table 4.1-5 in EC2 minimum dimensions for axis distance ‗a‘ and minimum depth ‗d‘ for R120 fire resistance are stated as;

amin = 35mm and dmin = 300mm for columns fully exposed to fire (internal columns)

amin = 25mm and dmin = 175mm for columns partly exposed to fire (edge columns)

amin = 35mm and dmin = 200mm for continuous flat slabs

amin = 35mm and dmin = 160mm for shear walls

amin = 30mm and wmin = 200mm for edge beams

Subsequent dimensions are accounted for in the analysis, design and detailing of structural members.

Furthermore, it is required that the design life of the structure be 50 years. Factors affecting the designlife of a concrete structure are the upper limit for the water/cement ratio, the lower limit to the cement

content, the nominal cover to reinforcement and adequate compaction, curing & detailing of the

concrete. For an XC1 Design life of 50 years, all cement types are applicable in construction; however

the strength class for concrete must be a minimum of C20/25. A maximum w/c ratio = 0.70 must also

be provided, whilst minimum cement content must be 240kg/m3. Furthermore, the nominal cover to

reinforcement cnom must equal 15mm + Δc, (typically Δc = 10mm from Eurocode 2 and BS8500: Part

1). Henceforth, all members within in the detailed design process should comply with theses

requirement to provide a 50 year design life and should provide R120 fire resistance for XC1 exposure.

3.5.2 Member sizing

Preliminary member sizing is an essential stage of the scheme design process, particularly if the

structural engineer is working with the Architect in the early stages of developing the clients‘ desired

structural form. Member sizing exercises provide invaluable information, as they are the starting point

of the FEA models members, and imperative for the later stages of design, as the architect can be made

aware of the typical member sizes to expect when determining the optimal layout for occupancy usage.

Member sizes were determined with usage of span/depth ratios provided from IStructE ‗Manual for the

design of concrete building structures to Eurocode 2‘. To determine member sizes, the layout of the

building, and constituents of structural components must be determined. Sizing is as follows:

Slabs sizing

Using figure 21 for ‗Table 5.8 Span/effective depth ratios for slabs‘ from Span/Depth ratio for the slab,

assuming reinforcement distribution of Asteel /bd < 0.35% for the entire composition of the slab gives aspan-depth ratio of 36 (IStructE 2010).

Figure 21 - Span/depth ratio for solid flat slabs



26

Longest span of slab is to interior column from edge column, where L=6.62m, acceptable depth, d =

6.62/36 = 0.18m

Therefore, flat slabs of 200mm will be adopted for all residential floors with live loading of 1.5kN/m2.

However, as the roof will take a larger imposed load of 4.0 kN/m2

plus the dead weight of a proposed

steel frame shelter, tall parapet walls and planters, a larger thickness of 250mm thick will be adopted for

the roof RC flat slab.

Columns:

Columns for two different heights must be considered. This is as all residential floors will have a clear

column height of 3.11m, whilst the larger column height of 8.11m will require an increased depth to

ensure that buckling does not occur due to the greater height.

Shorter height columns:Columns should typically be no less than 200mm in breadth, according to IStructE guidance. In

compliance with section 4.84 for ‗sizes of reinforcement and columns‘ in IStructE ‗Manual for thedesign of concrete structures to Eurocode 2‘, best general practice is stated as providing ―stocky

columns‖ (IStructE, 2006). Stocky columns will avoid the necessity of designing for the effects of slenderness. The value for of effective height to least lateral breadth (l o) ratio does not exceed 15. As a

preliminary measure effective height can be taken as heff = 0.85h. Given that individual column height is

3.11m, this provides an effective height of 2.635m. Dividing by the ratio, therefore lo = 0.176m. The

use of braced stocky columns will hopefully minimise the requirement at the detailed design stage to

secondary buckling effects likelihood of buckling failure as well the PΔ effect.

Longer height ColumnsFollowing the same procedure above, but for h = 8.11m, heff = 6.885. Therefore lo = 0.459m.

However, due to the large height of the overall structure, the columns will be sized to those larger than

the required lo for each different floor height. This is as the self-weight of the concrete will play a

significant part of axial loads induced in the structure and therefore will require an adequately largecross sectional area to resist the high anticipated axial loads. Furthermore, columns will need to be able

to resist the applied lateral loads and so a large enough cross section will be crucial in bending about

minor or major axes. Should these sizes prove inadequate during analysis then design iterations will

outline new column sizes. The adopted column sizes are as shown in table 2.

Column Name Dimensions Column Type Floors

C1 750mm ø - Circular Internal Ground Floor Mezzanine

C2 650x500mm -Rectangular Edge Ground Floor Mezzanine

C3 500mm ø - Circular Internal Residential 1st-24

th

C4 650x250mm - Rectangular Edge Residential 1st-24

th

Core Shear Walls:

The preliminary determination of the wall sizing will be based on span-depth ratios from typical values

from engineering experience rather than the more detailed methods required in EC2. This is as the

project will not consider extensively detailed design of the shear walls, but rather the contribution that

the core wall will give on the stability of the whole structure. Taking a height to thickness ratio of 30 for

continuous walls (Cobb, 2009), the following minimum depths for preliminary sizing are determined.

For storey h = 3.11m, t = 0.11m. For storey h = 8m, t = 0.27m.

Therefore, the overall adopted shear wall thickness will be 350mm, to ensure a larger size than the

preliminary minimum.



27

3.6 Wind loadEurocode 1 part 1-4 has been used in conjunction with the ‗IStructE - Manual for the design of building

structures to Eurocode 1 and Basis of Structural Design‘ to determine the wind load effects on the

structure. It is worth noting that due to the complex structural form that is traditionally outside of the

detail of EC1‘s scope, EC1 part 1-4 allows properly conducted and validated wind tunnel testing of a

model to provide a more accurate representation of the wind loads acting on the structure. However,due to lack of resources and to integrate usage of code in to the project, a more conservative approach

has been utilised using EC1.

The wind is assumed to comprise a fluctuating wind speed component that is superimposed on the mean

wind speed. This is used to determine peak velocity pressure qp(z) by considering the effects of the two.

Peak velocity pressure is the main parameter through which the wind forces will be determined;

however the nature of wind loading on a structure is affected by many factors, including national wind

climate, local terrain roughness, orogprahy, site altitude and directional seasonal effects.

(EC1 Clause 4.1)First, the basic wind velocity vb is determined, which is a function of the wind direction and time of

year at 10m above ground for terrain category II (Country terrain), and is given by the UK nationalAnnex as vb = cdir . cseason . cprob . vb,0 , where cdir is the directional factor given, cseason is the season factor,

cprob is the probability factor for the general case of an annual probability of exceedance of 0.02 (50 year

mean return period), and vb,0 is the fundamental value for basic wind velocity given by vb,0 = vb,map. calt.

(Clause 4.2)Altitude factor calt is determined by considering the site‘s altitude from sea level, as calt = 1 + 0.001 A for

sites with structural height above ground, z < 10m. However this formula can also be used for

structures with z > 10m as a conservative measure. The altitude of the site from sea level A = 7m,

therefore calt = 1.007. From EC1 NA part 1-4, vb,map = 22m/s for London. Therefore, vb,0 = 22.15m/s.

Due to the structures highly regular form, the calculation will only consider the worst-case direction

wind loading out of all the values of cdir as shown in figure 22. It is evident that the worst case wind load

will come from the direction of 240°, assuming the orientation of the structure to North is as shown in

figure 22. This adopted orientation is approximately representative to proposed site conditions.

However, a more accurate approach would need to consider the true orientation of the structure and

interpolate for values of Cdir.

N

30°

60°

90°

120°

150°180°210°

240°

270°

300°

330°

0.78

0.73

0.73

0.73

0.73

0.800.850.93

1.00

0.99

0.91

0.82

Figure 22 - Values for Cdir from UK NA to EC1 part 1-4



28

Coincidentally, the orientation of the structure as assumed in figure 22 means that the worst case wind

from 240° is applied perpendicular to the largest face of the structure and therefore will exert the largest

wind pressure on that available face. Cdir =1.00 for worst case direction.

Cseason = 1.00 for the worst weather winter period of October-March. Lower values of Cseason need only

be considered for temporary structures during construction.

Therefore;

vb = 1 x 1 x 1x1(22.16m/s) = 22.15m/s

(Clause 4.3)Having determined the basic wind speed, the effects of the local terrain on the wind load and the

variation with height must be considered. The site is in central London, surrounded by buildings and

developments. Under EC1 table 4.1 can be classified as in terrain category IV; an area in which at least

15% of the surface is covered with buildings and their average height exceeds 15m. The terrain

category is used to determine the terrain roughness characteristics which exert frictional effects on the

wind flow.

Furthermore, In Town terrain, closely spaced surrounding buildings may provide shelter which can

cause the wind to behave as if the ground level was raised to a displacement height (h dis) as shown in

figure 23. This lifts the profile of the peak velocity pressure. The displacement height should be

subtracted from the actual height (z) of the structure to give a reduced effective height (z – hdis). This

consideration is useful for determining a more accurate representation of the wind loads, however the

determination of the average height of the surrounding buildings needs to be found for all directions

considered, and in lieu of this topographical information a value of have can be determined through

assuming a storey height of 3m for surrounding buildings, whilst hdis = 0 for country terrain. However,

due to lack of detailed terrain information, the calculation for wind load effects will be conservative andnot factor hdis. Furthermore, orography is deemed insignificant for the site due to no significant changes

in topography (i.e. large hills, excavated areas, quarries etc).

Figure 23 - Aerial view of site location and surrounding terrain (location of A4 in red triangle)

Figure 24 - obstruction height and upwind spacing (IStructE guide to EC1 pp.108)



29

EC1 offers a simplified procedure for determining the peak velocity pressure for buildings where

orography is not significant. For sites of insignificant orography in Town terrain, peak velocity pressure

at a given altitude is given as .

Where qb is the basic velocity pressure ( , ce(z) is the exposure factor of the building,and cet is the exposure correction factor for sites in town terrain. The exposure factor c e(z) considers the

distance of the site from sea and the effective height for which the ce(z) is to be determined.

(Clause 4.5)To determine these effects, first the basic velocity pressure must be determined;

where ρ is the density, which in the UK is taken as 1.226kg/m3, and vb is the determined basic wind

velocity. As vb = cdir . cseason . cprob . vb,0, for 240° vb is;

Once qb is determined, then peak velocity pressure at height can be determined

by finding ce(z) and ce,T from EC1. However, before this is done the effective heights for which the

wind loads will be calculated need to be determined. EC1 recommends the applied distribution in figure

25 for buildings where height h is greater than two times of the windward breadth b. This considers the

structure as multiple parts, comprising: a lower part extending upwards from the ground by a height

equal to b; an upper part extending downwards from the top by a height equal to b and a middle region,

whilst between the upper and lower parts, the structure may be divided into horizontal strips with a

height hstrip.

For A4, h ≈ 86m (accounting for excluded parapet roof area) and b ≈ 25m, therefore the velocity profile

in figure 25 is used for the wind load forces. In the case of A4 h strip is equated to 3.11m , the height of

one individual residential storey, whilst the top and bottom strips have been taken as the closest storey

slab level to the width of b. The values of q p for the mid-section strips will be linearly interpolated

between the maximum values at the top and minimum value at the bottom. The effective heights for

each strip, along with corresponding values of ce(z), are as shown in table 2.

Figure 25 - Eurocode equivalent to wind loading for buildings up to 200m (IStructE, 2010)



30

Table 2 - Peak wind pressure values for different effective heights

Storey number z (m) ce(z)q (z) (N/m

2)

1 - 6 23.55 2.85 857.91 Lower strip

7 26.66 2.93 880.49

8 29.77 3.00 903.06

9 32.88 3.08 925.64

10 35.99 3.15 948.22

11 39.10 3.23 970.79

12 42.21 3.30 993.37

13 45.32 3.38 1015.95

14 48.43 3.45 1038.52

15 51.54 3.53 1061.10

16 54.65 3.60 1083.68

17 57.76 3.68 1106.25

18-25 + Roof 85.75 3.75 1128.83 Top strip

(Clause 5.3)To determine external and internal surface pressures, EC1 provides pressure coefficients for the

different areas of that must be considered to determine the exerted forces. Where

, .

To account for the external positive pressures and suction pressures, EC1 converts the external pressure

to the external surface pressure using the pressure coefficients in figure 26.

Figure 26 - Key to external pressure coefficients for a rectangular plan building (IStructE, 2010)



31

However EC1 provides detailed guidance for structures of simplified shape (i.e. rectangular plan

buildings) and therefore these external pressure coefficients have been conservatively adapted to the

form of A4, as shown in figure 27. Note; the same side lengths for the region for A and B (where d =

25m = A+B), were applied where LA = e/5 = 5m and LB = d-e/5 = 20m. The values highlighted in red

are those taken for A4. Values for cpe, depends on the size of the loaded area. cpe,1 values are given for

loaded external areas of

1m

2and cpe,10 values for loaded areas of > 10m

2. EC1 Part 1-4 gives a

recommended procedures for determining the cpe value for loaded areas between 1m2 and 10m2.

However under the UK NA, this recommended procedure should not be used in the UK. In the UK the

cpe,1 values should be used for all areas > 1m2

and the cpe,10 values should be used for all areas >1m2.

The cpe,1 values should be used for small cladding elements and fixings and the cpe,10 values for larger

cladding elements and for overall structural loads (such as in the case of A4).

Internal pressures coefficient cpi may be taken as the more onerous of +0.2 or -0.3 where the

permeability of the external surfaces and number of openings cannot be fully determined. The reference

height zi for the internal pressures should be taken as the effective height z e used for the external

pressures on the faces which contribute through their openings to the internal pressure. However, if

there are several openings EC1 recommends that the largest value of z e should be used to determine a

conservative value of zi and this is the approach that will be taken for determining cpi.

Furthermore, the values of net surface pressure for D and E can however be determined using Net

pressure coefficients of 1.3 to be applied to the corresponding values for w e to determine the netpressure on those faces.

Therefore, we only need to apply the value of wpi for faces A and B, which due to the values W pe of A

and B being negative pressure; the most onerous case will be the addition of a negative internal

pressure. Hence, taking Table 3 shows the values for external pressures for all faces using the corresponding values of c pe in

figure 26.

D

E

A

B

A B

Figure 27 - adapted external pressure coefficient key



32

Table 3 - External wind pressures on structure

W e (N/m2)

Storey

numberz (m)

qp(z)

(N/m2)D E A B

1 - 6 23.55 857.91 686.30 -600.54 -1029.50 -686.33

7 26.66 880.49 704.36 -616.33 -1056.50 -704.39

8 29.77 903.06 722.43 -632.12 -1083.50 -722.45

9 32.88 925.64 740.49 -647.91 -1110.51 -740.51

10 35.99 948.22 758.55 -663.70 -1137.51 -758.57

11 39.10 970.79 776.62 -679.49 -1164.51 -776.63

12 42.21 993.37 794.68 -695.28 -1191.51 -794.69

13 45.32 1015.95 812.75 -711.07 -1218.51 -812.76

14 48.43 1038.52 830.81 -726.86 -1245.51 -830.82

15 51.54 1061.10 848.87 -742.65 -1272.52 -848.88

16 54.65 1083.68 866.94 -758.44 -1299.52 -866.94

17 57.76 1106.25 885.00 -774.23 -1326.52 -885.00

18-25 +

Roof 85.75 1128.83 903.06 -790.02 -1353.52 -903.06

Summing internal and external pressures for faces A and B and multiplying the external pressures for D

and E by the net pressure coefficient of 1.3, Table 4 gives the net pressures for worst-case wind load

from 240° along the entire building height that will be applied to the corresponding faces.

Table 4 – Net wind pressures on structure

Storey

numberNet pressures (kN/m

2)

z (m)qp(z)

(N/m2)D E A B

1 - 6 23.55 857.91 0.89 -0.78 -1.20 -0.89

7 26.66 880.49 0.91 -0.81 -1.23 -0.91

8 29.77 903.06 0.94 -0.83 -1.26 -0.93

9 32.88 925.64 0.96 -0.85 -1.29 -0.94

10 35.99 948.22 0.98 -0.87 -1.32 -0.96

11 39.10 970.79 1.01 -0.89 -1.35 -0.98

12 42.21 993.37 1.03 -0.91 -1.38 -1.00

13 45.32 1015.95 1.05 -0.93 -1.41 -1.02

14 48.43 1038.52 1.08 -0.96 -1.44 -1.04

15 51.54 1061.10 1.10 -0.98 -1.47 -1.06

16 54.65 1083.68 1.12 -1.00 -1.50 -1.07

17 57.76 1106.25 1.15 -1.02 -1.53 -1.0918-25 &

Roof 85.75 1128.83 1.17 -1.03 -1.56 -1.11

The net surface pressures will then be resolved into simple static point forces applied on the slab

column connections on the corresponding outer faces of the FEA model. This is to ensure the rigid-

diaphragm assumption made in modelling in on SAP2000. The suction pressure on the roof has not

been accounted for in the analysis due to the simple flat nature of the roof. Therefore only snow load

will be included as part of the gravity loading for the roof, where a minimum value of 0.6kN/m2

has

been taken.



33

3.7 Serviceability and comfortability criteria for wind loading

Comfortability criteria

As previously mentioned, tall flexible buildings subject to wind load require an assessment of

oscillatory movements due to fluctuating winds. This is so as to inform the structural engineer of the

human responses which can be induced by acceleration of the structure. These can range from milddiscomfort to acute nausea or motion sickness.

The principal factor governing the degree of human comfort is the acceleration of the building.

Although no regionally applicable code specific guidance or regulations stating acceptable limits are

available, the following empirically derived effects of human perception will be used as preliminary

factors for assessment.

An accurate method to determine exceedance of certain levels of comfort is through wind-tunnel testingof the model; however a preliminary investigation into exceedance of comfort criteria will be made

through manual calculations later on.

Serviceability limits

The maximum applicable lateral deformation based performance objectives regarding the design of

high-rise buildings using Eurocodes for the characteristic combination (expression 6.14b in EC0) are as

summarised as:

Overall top deflection, ∆top < H/500

Interstorey drift, h/100

Figure 28 - human perception levels to wind induced acceleration (Hira, 2003)



34

4 Primary design and analysis

4.1 FEA ModelUpon completion of the scheme design stage, there is sufficient understanding of the structural

mechanisms to begin creation of a finite element analysis model (FEA) for detailed analysis. Finiteelement analysis offers a powerful tool for the structural designer as it allows the analysis and design of

complex structural forms within a short time span (relative in comparison to manual calculation). It has

proved an invaluable tool in the engineering professions, and is largely responsible to the vast speed and

confidence with which modern complex structures are now designed and constructed.

FEA (also finite element method, FEM) is in essence, a numerical analysis method for obtaining

approximate solutions of boundary value problems in engineering structures. The variables of these

problems are ranged and may differ with consideration of the problem at hand, but may include

physical displacement, temperatures, velocities and accelerations of structural components in

investigation or other dependent variables relative to the type of system being analyzed. These are all

problems whose solutions are bounded within the FEA model, by analysis of a computer model of a

simulated material or structure which is stressed and analyzed for specific results. In case of structural

mechanisms or failures, FEA may be used to help determine the design modifications required to be

made to the simulated system in order to meet the required design performance of the design codes or

engineers needs.

Engineering analysis of structural and mechanical systems has been addressed through derivation of

numerous differential equations, relating the variables of the problems through simple physical

principles such as equilibrium, Newton‘s laws of motion, conservation of energy, the laws of

thermodynamics etc... However, upon formulation of these models, the resulting mathematical models

can often prove impossible to solve through sheer manual calculation. Only solution of simple models

of 2D statically determinate structures of regular shape and solid form prove tractable (Lin, 2005). The

computational procedures within structural FEA involve finding approximate solutions for these partialdifferential equations and integral equations.

The basic concept of FEM concerns the process of converting continuous models and equations into

discrete counterparts (discretization). This entails subdividing the mathematical model into non-

overlapping components of simple geometry called finite elements. Each element is then expressed in

terms of a finite number of degrees of freedom, characterized as the value of unknown function(s) at a

set of nodal points. It is from this that the response of the model is approximated numerically. The

majority of commercial computer FEA analysis packages are based on the stiffness method, with it

forming the core code. The stiffness method is particularly suited for computer-automated analysis of

complex, statically indeterminate structures. The application of the method entails modelling the

structural system as idealised elements connected at nodes. The material stiffness properties of the

element are then compiled into a single matrix equation which governs the behaviour of the entireidealized structure. Through this, the unknown variables of the problem, such as displacement and

forces can then be determined through solving these equations.

FEA programs traditionally come with element libraries, and can often be developed over repeated

usage of the package. Typical elements types are (but not limited to:

Cable/Tendon elements

Frame/Beam elements

Shell/Area elements

Solid Elements

Spring Elements

Mass Elements

These elements are connected at the nodes of the model, and restraints and/or constraints are applied in

order to simulate the structural conditions intended to be achieved.



35

In area/shell elements, the nodes form a grid called a mesh. This mesh is programmed to contain the

material and structural properties which define how the structure will react to certain loading

conditions. Nodes are assigned at a certain density throughout the material depending on the anticipated

stress levels of a particular area. It is typical to apply a fine mesh to areas of anticipated high stress and

complex distribution, whilst areas of lower and less complex stress will receive coarser meshing. The

mesh acts like a spider web in that from each node, there extends a mesh element to each of the adjacent

nodes. This web of vectors is what carries the material properties to the object, creating numerous

elements within that section. These elements must have section properties relating to geometry and

composition assigned to them, and fundamentally a material type assigned. FEA packages will typically

offer the capability to use multiple materials within the structure and even determination of composite

actions. These will allow isotropic and orthotropic material definitions, and input options for all forms

of Material characteristics ranging from Modulus of Elasticity, Shear Modulus, yield stresses and

densities to behaviour under different temperature ranges.

Furthermore, the structure must be loaded in the program to simulate and determine its behaviour in

use. FEA packages will offer multiple loading conditions, ranging from Static and Dynamic point,

pressures and gravity loads, to thermal loads, enforced displacements heat flux and convection etc...

There are generally two types of analysis employed in industry: 2-D modelling, and 3-D modelling.

While 2-D modelling conserves simplicity and allows the analysis to be with relatively minor

computational requirements, however it tends to yield less accurate results. 3-D modelling, however,

produces more accurate results but with far more complex computational requirements. Within each of

these modelling schemes, the user can insert numerous algorithms (functions) which may make the

system behave linearly or non-linearly. Linear systems are far less complex and generally do not take

into account plastic deformation, whilst non-linear systems do account for plastic deformation, and

many also are capable of testing a material all the way to fracture (Widas, 1997).

The Structural analysis and Design process within an FEA package can be generalised within the

following steps:

1.

Create or modify a model that numerically defines the geometry, properties, loading, andanalysis parameters for the structure

2. Perform an analysis of the model

3. Review the results of the analysis

4. Check and optimize the design of the structure

This is usually an iterative process that may involve several cycles of the above sequence of steps.

The technical aspects, mathematics and concepts of FEA are regions so complex and broad that in their

selves could fill libraries with the accumulated knowledge and research developed over the years in the

fields‘ existence. Thus, it is outlined that the intention of this project is not in detailing the limitations

and complex mathematics of FEA, but rather the utilisation of it as a tool in structural design common

in engineering practices throughout the world. It is however imperative that the engineer is highlyfamiliar with the assumptions of their utilised FEA package and aware of its limitations. The economic

advantages of FEA in the structural design project outweighs that of traditionally time intensive manual

analysis, but must be approached with a fundamental understanding of structural behaviour and

complemented with manual approximation and checking.

4.1.1 Construction and Assumptions

The FEA model is only as valid as the assumptions made during its construction. Otherwise, given the

incorrect structural assumptions applied within the construction of the model, the model will not exhibit

the behaviour expected of the real-life structure. The utilised FEA package for this project is SAP2000.

The model was constructed befitting to the structural scheme determined in the scheme design stage.

Element types corresponding to the frame members with the material properties outlined from scheme

design are as shown in the table 6.



36

Table 5 - Structural members and corresponding element types

Member type Member name Dimensions FEA element type

Column C1 C1 - 750mm ø - Circular Frame

Column C2 650x500mm -Rectangular Frame

Column C3 500mm ø - Circular Frame

Column C4 650x250mm - Rectangular Frame

Beam Edge beam 600x300mm deep - Rectangular Frame

Flat Slab RC flat slab 200mm deep Area (shell with 4x4 mesh)

Flat Slab RC roof slab 250mm deep Area (shell with 4x4 mesh)

Shear Wall Shear Core 350mm thick Area (shell with 4x4 mesh)

The structure has been modelled with frame elements as column members, with flat slabs modelled as

shell area elements sitting atop the columns. The typical floor plan shows that 20 individual area

elements of the same RC flat slab properties for each floor make up one individual floor slab. To ensure

that each individual area behaves as a whole floor diaphragm, the connections between each slab edge

is rigidly fixed to the adjoining slab edge. All connections between frame elements and area elements

(slab and shell) are rigidly fixed, so as to mimic the continuous nature of the RC structure, and the‗vertical cantilever‘ mechanism of the high-rise building. The base connections to the foundation are

pinned, and allow no rotation or translational behaviour of the columns at base connection. Figure 29

outlines the model components and the Global coordinate system adopted in the model.

Column and

beam frame

elements

Core area

elements

Slab area

elements

Pinned base

connections

Y

X

Figure 29 - Typical floor plan and elevation of A4 model in SAP2000



37

To ensure the continuity between columns on all floors, the columns and beams members have been

offset at insertion points to nodes to distances so that external faces of the columns are flush with the

slab edges and that all columns sit directly atop of each other.

In order for the lateral load analysis to exhibit the behaviour of the structural mechanisms outlined inthe earlier chapters, the structure will be modelled with diaphragm constraints to model the assumed

rigid diaphragm at each storey. The diaphragm constraint will enable all constrained joints to move

together as a planar diaphragm that is rigid against membrane deformation. This assumption allows the

horizontal plane displacements of all vertical elements to be definable in terms of the horizontal plane

rigid body rotation and translations of the RC slabs. Hence, the number of unknown displacements to be

determined in the analysis is greatly reduced. When applied to the shell elements used for the RC slabs,

this will constraint will not permit the rotation of the slab members. The rigid diaphragm assumption

should prove adequate for modelling the high in-plane stiffness exhibited by concrete floors. Figure 31

demonstrates the constraining mechanism used in SAP2000 to create the diaphragm constraint.

Figure 31 - Use of the diaphragm constraint to model a rigid floor slab

Figure 30 - extruded view of FEA model with offsets and insertion points of structural elements



38

The use of the Diaphragm Constraint for building structures is very useful in the lateral dynamic

analysis of buildings, as it results in a significant reduction in the size of the eigenvalue problem to be

solved in Modal analysis (CSI, 2009).

4.1.2 Loadings

Gravity loadings (dead and live) have been applied as area loads (kN/m2) onto the corresponding slab

elements. The perimeter dead load for the cladding however, has been applied as a line load (kN/m) on

frame elements of no defined section property, interconnected at the edge nodes joining slab edges and

exterior columns. This will mimic the line load and apply it directly onto the edge of slab members

which in turn will transmit these to the columns. This is as SAP2000 does not permit line loading on

area elements, and so this method has been used to mimic this form of loading.

Due to the regularity of the structural form (both in plan and elevation), the lateral loads will only be

applied in a single direction. This is in the Y-direction of the model‘s global coordinates, and for wind

loads results in the wind load being applied on the largest area of frontal projection. Wind loads will be

applied as static point loads on the column-slab nodal joints. The dynamic loads for earthquakes will be

applied via SAP2000 as a result of a simulated ground motion, and the forces are therefore determinedautomatically in SAP2000.

4.1.3 Materials

The behaviour of the structural materials and components have been taken as linearly elastic. This

assumption allows the superposition of actions and deflections, and thus the use of linear methods of

analysis. It is due to the development of linear methods and their computationally less tasking

procedures that it is possible to analyse large complex, statically indeterminate structures.

From the scheme design process, it was determined that C50/C60 concrete will be used as the concrete

type for the entire RC frame. Furthermore, EC2 permits only the usage of Grade 500 steel for

reinforcement bars, whose characteristic strength f yk is to be taken as 500MPa.

Figure 32 - Static wind point loads applied on positive wind pressure face



39

The material properties input to SAP2000 are as shown in table 6, and comply with BS EN 206-1:2000

―Concrete – Part 1: Specification, performance, production and conformity‖, and should thereforeensure conformity of the design materials to Eurocode requirements (BSI, 2006)

Table 6 - Material properties in SAP2000

Material Type Bulkdensity

(kN/m3)

Modulus

of Elasticity,

E (GPa)

ShearModulus, G

(GPa)

Poissonsratio, ν

Yield strength(N/mm2)

C50/60 Concrete 25 37 15.41 0.2

f c,k = 50 (Cyl

compressive

crushingstrength)

Grade 500

Reinforcement

bars

Steel 78 210 80 0.3

f y,k = 500

(characteristic

yield strength)

4.1.4 Load cases and combinations

The following load Combinations corresponding to those mentioned in section 3.44 comprised of thefollowing individual loads with the relative partial factors for permanent (Gk ) and variable (Qk ) action

type applied where utilised in the FEA analysis. Each individual load case component of the load

combinations can also be used to determine the structural behaviour under each individual component

as well.

Load cases

Load combination

name

Dead (inc self-

weight of structure

and Perimeter load)

Live

occupancy

Wind load (all

directions

ABCDE)

Seismic – Response

spectrum 0.18g

Seismic – Response

spectrum 0.04g

ULS gravity load &

unfavourable wind1.35Gk 1.5Qk 1.5Qk - -

ULS gravity loadfavourable & wind 1.35Gk 1.5Qk 1.0Qk - -

ULS gravity load &

unfavourable seismic

PGA 0.18g

1.35Gk 1.5Qk - 1.5Qk -

ULS gravity load &

favourable seismic

PGA 0.18g

1.35Gk 1.5Qk - 1.0Qk -

ULS gravity load &unfavourable seismic

PGA 0.04g

1.35Gk 1.5Qk - 1.5Qk 1.5

ULS gravity load &

favourable seismic

PGA 0.04g

1.35Gk 1.5Qk - 1.0Qk 1.35

4.1.5 Simplification and limitations in modelling

The major simplification made in FEA modelling of the structure is the exclusion of openings in the

shear wall. As the shear walls structural use as a non-load resisting member is to house the stairwell and

elevator access routes, openings to ensure exit and entry of the occupants will be required for that

region. This would typically be done by including openings in the architecturally specified locations of

the door ways. The lack of openings in the FEA modelling should not affect the output for the overall

structural behaviour, as in detailed design, the location of openings would ensure adequate stiffening of

the edges of the openings through sufficient reinforcement detailing so as to deal with the increased

stress concentrations at the perforations of the shear core. It is assumed that post-performance analysis,

wherever openings be included in the detailed design they will be adequately reinforced so as not to

affect the uniform behaviour of the shear core in resisting lateral load.



40

5 Secondary design and analysis

5.1 Desirable characteristics for desired lateral response anddynamic analysis

For both wind and seismic response, there are certain principles in the structural form that can aid inminimising the effects of lateral loading. It may be observed that these principles are essentially

qualitative in nature, and largely concern the scheme design stage of A4; however they are factors that

will inherently affect the later stages of design, e.g. RC detailing of the structure, construction issues,

quality of materials, workmanship and the correct establishment of seismic response data at the site.

The basic principles as followed are:

Structural simplicity

Uniformity, symmetry and redundancy

Bi-directional resistance and stiffness

Torsional resistance and stiffness about the vertical axis, θz

Adequate performance of the floor slabs acting as diaphragms to distribute lateral loads

Adequate foundations

For the purpose of structural simplicity, the simplified design of A4 has provided continuity from

ground floor to roof. Furthermore, the adopted structural plan is provides three axis of rotational

symmetry. Hence, for analysis purposes, the application of the lateral loads (both wind and seismic) in

only one direction should prove adequate. This chosen direction for seismic loading will be in the Y-

direction of the models global co-ordinates, which coincides with the 240° worst case wind-load

direction, and will therefore provide a good basis for accurate comparison of behaviour of the two load

types.

Uniformity in both elevation and plan has been provided to some extent. There are no setbacks or

protruding members in elevation, and the members in plan continue from ground to roof, with no

transfer mechanisms or eccentrically spaced column members. However, the larger floor height of thebottom most storey may attribute some soft-storey features.

The continuation of the core from ground to roof in a uniform and thin-walled tube manner will result

in significant torsional stiffness, and therefore no analytical consideration will be made into the

torsional behaviour of the structure. Torsional stiffness is assumed adequate.

The rigid diaphragm assumption will mimic the high-in plane stiffness of the RC slabs, and therefore

this criterion will be satisfied for analysis purposes. Lateral forces are assumed to be transmitted

adequately. Finally, as the project does not detail itself with the substructure, it is assumed to be

adequate in that all gravity and lateral loads will be transmitted to the foundations from the

superstructure, and that all ground motion will be adequately transferred to the superstructure to

determine behaviour to seismic excitation.

Non-regular configurations are expressly discouraged in EC8 Part 1 Clause 2.2.4.1 which states: ―Tothe extent possible, structures should have simple and regular forms both in plan and elevation‖. It maybe observed that buildings classified as non-regular are permitted by EC8, but lead to more onerous

design requirements. Compliance with the conditions for regularity or moderate irregularity be

considered as satisfying many of the principles for good conceptual design listed in Section 5.1 above.

In particular, regular or moderately irregular buildings should generally possess a satisfactory level of:

The majority of the desired characteristics have been presented in the design of A4 and therefore should

provide satisfactory preliminary behaviour to seismic loading (from a qualitative aspect).



41

5.2 Seismic analysis

5.2.1 Seismicity of site

Eurocode 8 defines the entire territory of the UK as an area of very low seismicity, for which in routine

practice the application of EC8 to structural design is not required. It is not a country in anywayassociated with high seismic activity such as that of Japan or Western parts of America (e.g. California).

Nevertheless, provisions are made in the UK national annex that certain types of structures through

reason of their local geography, function or form may warrant seismic consideration in their structure.

These forms of structures are defined in the EC8 UK NA supporting document ‗PD6698:2009 -

Recommendations for the design of structures for earthquake resistance to BS EN 1998‘, where in some

cases the function of a structure is such that failure due to very low probability events, including

earthquakes, might need to be considered. Four example categories from the document are:

Structures whose failure poses a large threat of death or injury to the population (e.g. Nuclear

Power Plants, Natural Gas storage tanks, high pressure pipelines etc...)

Structures which form a part of the national infrastructure of which the loss would have large

economic consequences (Major transportation bridges, Dams etc..)

Structures whose failure impedes the regional and national ability to deal with a disaster causedby a major damaging earthquake.

Strengthening or upgrading historic buildings that form an important part of the natural

heritage.

Certain structures could fall into more than one of these categories. It would not be unexpected in a

design situation for a structure such as A4 to be regarded as seismically insignificant in its function,

location and form to warrant seismic consideration.

However, the UK cannot be disregarded for occasional minor seismicity and it is known for some

structures to undergo structural damage from these minor seismic events. Although seismicity induced

damage has predominantly been on older structures and largely results in effects such as cracking of

walls and collapse of protruding members such as chimneys, the potential for earthquakes that may

exceed the predicted (and historical) events is indeed there (Musson, 2003). The EC8 seismic hazard

zoning map denoting PGA for the UK is as shown in figure 33.

Figure 33 - Seismic hazard map of Peak Ground Accelerations on rock (PGA) for 475 year and 2500 yearreturn periods (IStructE manual to EC8)



42

The information for PGA‘s in EC8 has been determined through seismicity models from historical and

instrumental readings for the UK, and equated to the ground motion model as shown (Musson and

Sargeant, 2007). The region of greatest seismicity is the North Wales region, and historically the largest

earthquakes (corresponding to the 2500 year return period) have been of magnitudes ranging from

approximately 2-5.5 ML (BGS, 2009). London exhibits a very low PGA (0.02-0.04g) however,

geological predictions are stating possible exceedance of these values within London, for a possible

seismic event of up to a magnitude of 5.5 M L (Musson, 2003). It is stated that although modernbuildings will likely exhibit no major structural damage, the deformations induced may lead to damage

of non-structural elements such as cladding, surface finishes, services etc. The possibility of minor

damage to structural elements need not be ruled out however. Seismically induced lateral deflections

should be limited to prevent distress in structural members and architectural components. Non load-

bearing in-fills, external wall panels, and window glazing should be designed with sufficient clearance

or with flexible supports to accommodate the anticipated movements.

Furthermore, recent global seismic events have prompted worldwide reconsideration and review of

seismicity modelling and hazard zoning, which will in turn affect the integrated models for response

detailed in the relevant regions. Events such as the 2011 Christchurch, New Zealand earthquake (ML

6.3) and the more recent 2011 Tohoku earthquake in Japan (ML ~9.0) were both events of which

significantly exceeded seismic effects considered in typical structural design projects for theirrespective regions.

With the above justifications, a consideration into the minor seismic performance of the structure within

London will be carried out, contrary to usual UK practice. These will be for two PGA values of 0.04g

(2500yr return) and 0.18g, taken as a theoretical maximum taken for the largest UK PGA measured in

Wales. It is expected that seismic loading will not be the predominant cause of lateral deformation, and

shall prove minor in comparison to wind. However consideration into the multi-hazard performance

with wind may pose an issue.

5.3 Structural dynamicsDynamic analysis of a structure is essentially a two-part process. First it is necessary to determine the

basic dynamic properties and behaviour of a structure (natural frequencies and mode shapes) through



43

analysis in the absence of loading, with the second being the usage of this information to determine the

response of the structure. It is noted that earthquakes often induce non-linear response of structures,

however most practical dynamic analysis (typically in seismic design) continues to be of the linear.

Although both wind and seismic forces are essentially dynamic in nature of loading, there is a key

difference in the manner in which they are induced upon a structure. Wind loads, applied as external

loads, are characteristically proportional to the exposed surface of a structure, while the earthquake

forces are principally internal forces resulting from the distortion produced by the inertial resistance of

the structure to earthquake motions. Essentially, the magnitude of earthquake forces induced within a

structure is a function of the mass of the structure rather than its exposed surface. Simply put, in wind

design, one would feel greater assurance about the safety of a structure composed of weighty sections,

whilst in seismic design; this does typically does not produce a safer design.

5.3.1 Modal analysis outputs

From SAP2000, the following Mode shapes with the modal characteristics in table 7 were determined

through computer analysis. In any structure, there are horizontal, vertical and combined modes of

behaviour, for instance, the model of A4 exerted vertical and twisting modes as well as horizontal

modes. The vertical modes have not been considered in analysis for earthquake and wind loading, andalthough the twisting modes of the structure in the ϴz is shown, it will not be considered in the analysis

of seismic loading. This is as torsional stiffness is assumed adequate and so torsional effects will be

excluded due to time constraints.



44





45


Table 7 - Modal informationMode

Number

Direction Natural

Period,

Tn

(Sec)

Natural

Frequency, f n

(Hz)

Natural

Circular

Freq, ωn

(rad/s)

Eigenvalue

(Rad2 /s

2)

MPR in

direction X

(%)

MPR in

direction Y

(%)

1 Y 2.634 0.380 2.386 5.692 0.487 68.324

2 X 2.604 0.384 2.413 5.824 68.223 0.487

3 ϴz 0.858 1.166 7.323 53.623 1.09E-05 2.069E-06

4 Y 0.528 1.895 11.905 141.730 0.151 19.282

5 X 0.522 1.914 12.027 144.640 19.352 0.152

6 ϴz 0.289 3.465 21.773 474.070 2.25E-06 1.454E-06

7 Y 0.218 4.594 28.864 833.150 0.053 4.724

8 X 0.216 4.620 29.029 842.690 5.182 0.053

9 ϴz 0.176 5.675 35.658 1271.500 7.99E-05 1.531E-06

The mass participation ratios (MPR) are an indication of how much of the mass is contributing in the X

and Y directions for each particular mode. MPR values for the Z direction have been excluded. It can be

seen that for the first horizontal modes for X and Y, the MPR is in favour of the direction, ~68% for

respective direction of X and Y modes. The twisting modes exhibit negligible mass participation due to

the turning of the mass about the centroidal axis of the structure. In essence, the mass participation

factor gives a representation of how much of the structure is contributing to the eigenmode, which in

turn means how much vibrational energy is dissipated through the eigenmode. It is evident that the first

three modes show significant MPR‘s, and that is why in common seismic design, the first few



46

fundamental modes of the structure are considered due to the largest participation of the mass of the

structure in these modes (and hence the largest inertial forces).

5.4 Structural response to ground motion

5.4.1 Desirable characteristics of earthquake resistant buildings

As previously mentioned, there are certain desirable characteristics of a building for lateral actions.

With regards to seismicity. EC8 Part 1 Clause 4.2.3 sets out quantified criteria for assessing structural

regularity, complementing the qualitative advice on symmetry and uniformity given in clause 4.2.1, and

these criteria are elaborated in the code.

In seismic design situations, it may be observed that concrete buildings are designed to provide

ductility, in order to ensure energy dissipation (damping) by plastic deformation. As previously

mentioned, the damping characteristics of the building can be provided by ductility within the structural

members. Overall ductile behaviour is ensured if the ductility demand involves dissipation of seismic

energy to a globally large volume of the structure. This entails the spread of seismic energy to different

elements and structural members, in order to achieve plastic deformation. However, as the analysis islinearly elastic and no plastic deformation is being considered. This will not be in the assessment.

EC8 sets out three ductility classes for seismic design criteria. Three classes of damping exist within

EC8, Ductility class low (DCL), ductility class medium (DCM) and ductility class high (DCH).

Structures which fall into the category of DCM and DCH structures are to be designed as dissipative.

This however is not the case for A4, as due to the UK being an area of very seismicity; A4 only requires

to be designed as DCL structure. A concrete DCL structure needs must conform to EC2 part 1-1 in

design and provide ductility class B or C reinforcement in the detailing of the structure, from Table C.1

of EC part 1-1. The required reinforcement characteristics must:

have a characteristic yield strength 400MPa ≤ f y,k ≤ 600MPa,

a minimum bar size 8 mm or greater Make use of ribbed bars so as to enable adequate bonding between concrete and steel

Including the standard detailing requirements stated by EC2 part1-1, no further consideration into

seismic capacity, detailing and design provisions are required for DCL structures. Hence, the scheme

design and material properties of A4 conform to DCL requirements. The final criteria that must be met

is that the DCL structure meets the relevant serviceability criteria.

Should DCM or DCH be required, the provisions that must be made in seismic design are inclusive of,

but not limited to:

Specific seismic detailing requirements of primary seismic components (main lateral load

resisting members, shear wall, rigid frame etc...)

Application of partial factors to material properties for ULS design.

Member sizing performed to requirements of EC8, not just EC2 part 1-1

5.4.2 Damage limitation

The main serviceability performance criterion for seismic loading that the building will be assessed by

is the interstorey drift limit, dr for each storey of height, h. The interstorey drift is the relative

displacement of the observed storey, with respect to the storey below. Figure 34 demonstrates the

concept of storey drift in a lumped mass model for EC8.



47

The limitation of interstorey drift is fundamental to damage limitation in cases of lateral seismic

loading. By meeting the criteria for dr, the damage limitation criterion may generally be deemed assatisfied. Section 4.3 of EC8 part 1 denotes the equation as shown in figure 34. The reduction factor ν =

0.5 for buildings of class I and II importance. Therefore the inter-storey drift ratio for the two floor

heights are as follows:

The above values for dr will be used to evaluate satisfactory seismic performance to EC8.

5.4.3 Response spectrum analysis to EC8

Earthquake ground motion is complex and the variation of it is erratic. However, a simplification can bemade to illustrate the main characteristics of response, as in the example of assuming sinusoidally

varying ground motion with circular frequency, ω, with a period of Tg = 2π/ω as .

Typically there are three forms of structural response to seismic loading.

1. For a structure where the motion of the ground is much slower than the rate of the structures

natural oscillations (Tg /Tn < 1), the structure will simply move with the ground. Therefore the

displacement of the structure is approximately equal to that of the ground. Thus no major

internal deformation occurs within the structure.

2. For a structure where the ground motion frequency and the natural frequency of the structure

are similar or the same (Tg /Tn = 1), resonance occurs and this results in a large dynamic

amplification of the structures motion. This is where the inertia forces and stiffness forces are

approximately equal and opposite so that the main resistance to the motion is provided by thedamping of the structure. As the main dissipation to the energy is provided by the damping,

which is provided by the structural components and materials, resonance can have a

deteriorating effect on the structural components.

3. However, if the ground motion is much faster than the natural oscillations of the structure

(Tg /Tn > 1) than the mass of the structure undergoes less motion than the ground, with its

stiffness and damper acting as vibration absorbers.

By considering a simplified SDOF system, the effects of loading rate will be very different for varying

damping levels of the structure. If we consider the acceleration of the structure at these three different

stages of ground motion periods, figure 35 for the sinusoidal ground motion with period Tg, we get the

following graph for the different ratios of damping.

For building having non-structural elements of brittle

materials attached (such as A4‘s cladding):

Where:

h, is storey height

dr, is the design inter-storey drift

ν, is the reduction factor which accounts the lower

return period of the seismic action associated with

the damage limitation requirement.

Figure 34 - Interstorey drift (EC8)



48

On the left hand side of the graph is the response regime of T g /Tn < 1, and the large resonant

amplification increases around Tg /Tn = 1. When Tg /Tn is unity, then pure resonance occurs and the

amplification factor increases to where the ratio X/Xg is approximately equal to 1/2ξ. The peak displacement at resonance is very sensitive to damping, and hence amplification is infinite when ξ = 0.This is the essential concept of the behaviour of dynamic response, but it must be noted that dynamic

amplification under realistic earthquake loading is much different than as shown in figure 35. This is as

Earthquakes are not sinusoidal in time-history, and tend to have finite and short durations of occurrence.

There are a whole range of complex scientific methods to determine the time-history response of the

structure to earthquakes; however this is graphically presented as shown in figure 36 as what is known

as a time-history graph. This plots acceleration of the ground motion against time during the earthquake

event, and is a representation of the earthquake ground motion variation with respect to threeorthogonal directions (typically N-S, E-W and vertical). It is evident in figure 36 that the acceleration

over time changes, and is thus a complex ground motion which cannot be simplified as sinusoidal.

Although the evaluation of such a case is complex, the behaviour under general dynamic load can be

quite easily understood by comparison with a single-frequency sinusoidal load case as discussed before.

The structures natural frequency will often lie within the range of periods exhibited in the loading, and

will tend to pick up and amplify the components of the ground loading that are close to its natural

period, just as it would within the sinusoidal case. Thus, the response of the structure will be dominated

by vibration at or close to the natural period of the structure. However, due to the lack of constant

amplitude in the response and due to the fact that the duration is finite and short in nature, the

amplifications achieved will therefore be smaller than that of in the sinusoidal nature.

Figure 35 - Displacement amplification factor curves for an SDOF structure subject to sinusoidal ground

shaking (Williams 2009)

Figure 36 - Typical time-history accelerogram



49

The response of a structure to a particular earthquake can be summarised using a response spectrum

analysis. This is where time-domain response of numerous SDOF systems of different natural periods is

computed, and the maximum absolute displacement, acceleration or velocity of the structure is achieved

through plotting it as a function of the SDOF systems period. Essentially, the response spectrum

approach of analysis shows the peak response of a SDOF structure to a particular earthquake, as a

function of the natural period and damping ratio of the structure. The response spectrum method offers

an advantage in design, as it is useful in dealing with a future earthquakes, whose precise nature is

unknown. The concept of determining response spectra for design purposes is to compute spectra for

several different earthquakes (i.e. of that local region/continent), and then to envelope and smooth them,

resulting in a single curve that encapsulates the dynamic characteristics of a large number of possible

accelerograms.

EC8 defines a range of elastic response spectra, which can be divided into two categories, Type 1 andType 2. Type 1 spectra are for areas of high seismicity (Ms > 5.5), whilst Type two is for areas of

moderate to low seismicity (Ms ≤ 5.5). Within each category, spectra are given five different soil types ;

A-Rock, B-very dense sand or gavel, or very stiff clay, C – dense sand or gravel or stiff clay, D – loose-

to-medium cohesionless soil or soft-to-firm cohesive soil and E – soil profiles with a surface layer of

alluvium thickness 5-20m. The vertical axis of the EC8 spectra denotes spectral acceleration, S e, which

is normalised by the design peak ground acceleration ag, whilst the horizontal axis denotes the periods of

the envelope (Elghazouli, 2009).

Within the graph, Se (T) is the elastic response spectrum, T is the vibration period of a linear SDOF

system, ag is the ground acceleration on rock, TB is the lower limit of the period of the constant

acceleration branch (Tg /Tn ≈1), Tc is the upper limit of the period of the constant spectral acceleration

branch, TD is the value defining the beginning of the constant displacement response range of thespectrum, S is the soil factor and η is the damping coefficient factor determined by .

As with the simple harmonic load case shown earlier on, it is evident those similar regimes of response

exist. Very stiff, short period structures will move with the ground, whilst at intermediate period the

dynamic amplification of the ground motion occurs, and at longer periods the structure moves less than

the ground beneath it. The region of spectra TB - TC the acceleration is constant, whilst between TC - TD

is the state of constant velocity, whilst beyond TD the region of constant displacement of the response

occurs.

For EC8, for the location of A4 and the ground conditions, the type two spectrum has been chosen for

soil type D (for an assumed ground conditions of London clay, due to lack of geotechnical data).

Figure 37 - Typical response spectra with envelope (left) with EC8 Response spectrum (right)



50

In the use of response spectrum analysis, the intention is to determine the force and maximum

displacement to which the structure is subjected to. This is achieved by determining the natural periods

of the structure, and the damping ratio, and then running a multi-modal analysis. Then the peak spectral

acceleration Se that is experienced by the structure can be determined from the response spectrum

curve. However, before this can be done, certain characteristics with regards to the structural

characteristics must be determined.

As A4 is to be designed as a DCL structure, a behaviour factor, q is included in EC8. The behaviour

factor q is a structure-dependent parameter used to reduce seismic design forces (but not design seismic

deflections) below those corresponding to elastic response. It is a function of the ductility of the

structure (i.e. its ability to sustain repeated deformations into the inelastic range without significant

degradation of stiffness and strength) and the ratio of ultimate lateral strength to lateral strength at

effective plastic yield a value q of up to 1.5 can be taken for analysis of a concrete frame that is regular

in plan and elevation, and so for simplicity, this has been taken as the preliminary value. The methods

of analysis permitted by EC8 are as shown in figure 39, with the method taken highlighted in red.

Once q has been determined, the application of the response spectrum analysis can be used in one of the

two possible equivalent linear analysis methods. The lateral force method equates the spectral

acceleration of the structure to static point forces that can be used in linear analysis (similar to the

method employed in the wind load analysis), whilst the multimodal analysis, which accounts the

response of all the modes of vibration contributing significantly to the total response. The latter is best

performed using FEA software as it can prove time-consuming and is complex in nature, thus not suited

to manual calculation. EC8 states that where the first mode of the structure T1 < 4Tc or 2.0s, then thelateral force method is applicable to analysis. As T1≈ 2.6s for the first mode (from modal analysis), for

TC = 0.3, this means that multimodal analysis is required for A4 structure.

Figure 38 - Values of horizontal response spectrum parameters recommended in EC8

Figure 39 - Methods of analysis for new buildings permitted by EC8 (IStructE manual to EC8)



51

SAP2000, has been used to perform the multi-modal analysis, where in the above parameters for q, T B,

TC, TD and s have been input. This in turn will be used to automatically determine the structural loading,

and thus the deformations exhibited in the structure.

5.4.4 Multimodal analysis response spectrum analysis using SAP2000

Multimodal analysis is permitted by Eurocode 8 in where the response of all the modes of vibration

contributing significantly to the total response must be taken into account (T1 < 4Tc). This principle is

deemed to be satisfied if either of the two conditions below is fulfilled in each principal horizontal

direction:

the sum of the effective modal masses for the modes considered reaches at least 90% of the

total mass of the structure.

all the modes whose effective modal mass is higher than 5% of the total mass are taken into

account.

Both parameters were verified in SAP2000, for the numerous modal cases considered. Multimodal

response spectrum analysis encompasses the principle of modal superposition, which states that any setof modal displacements can be expressed as a linear combination of the eigenvalues (modes shapes).

This relationship is denoted by the equation:

The coefficients Yi denote modal displacement, the displacement of the structure/component during that

mode. The modal displacements are a function of time whilst mode shapes are functions of their

position. The equation for modal superposition allows us to transform the equation of motion into a set

of equations in terms of modal displacements (as opposed to degrees of freedom):

However, in the case of a 3D model where it may prove difficult to satisfy this condition, the principle

may be satisfied by taking into account a number of modes at least equal to k ≥ 3√n, where n is thenumber of levels in the structure. This equates to a multimodal analysis of 15 horizontal modes for A4.

The period Tk cannot be greater than 0.2 seconds. This was verified in FEA calculation, and is evident

from the first three modes that we have considered as the T 3 ≈ 0.216, and the period will only getshorter for the higher modes of excitation.

To combine the modal responses, the total seismic action effect E E (force, displacement, and other FEA

variables) may be taken as the square root of the sum of the squares (SRSS) of the action effects EEi due

to individual modes, provided all the modes can be regarded as independent. The SRSS method equates

the peak overall response as the sum of the squares of the peak modal responses, as .

Defining the previously mentioned parameters in SAP2000, and computing modal responses for a

ground acceleration of 0.04g and 0.018g in the global Y-direction of A4, the seismic effects were

determined as modal contributions, and combined using SRSS to give estimates of total response by

SAP2000. Figure 40 demonstrates the input methods for multimodal analysis in SAP2000. The outputs

where then added to the gravity load combinations through means of linear elastic superposition, and

the structural behaviour determined under seismic actions for A were determined.



52

Figure 40 - Response spectrum parameters for EC8 in SAP2000



53

6 Design PerformanceUnder analysis, the figure 41 demonstrates the behaviour of the structure under the different loading

types, with a deflection scale factor of 100 applied to the graphical output from SAP2000. Figure 41

demonstrates the different deflected forms in the different loads applied to the structure, with the grey

wire-shadow representing the original undeformed shape of the structure.

From a qualitative perspective, it is evident that the prevailing lateral displacement is due to windloading, and the expected uniform gravity loading which includes structural self-weight (dead) and

Gravity load Wind Gravity & wind load

Figure 41 - Deformed shapes in FEA model

Seismic 0.04g Seismic 0.18g Gravity & Seismic 0.18g



54

occupancy (live) loads is evident. In comparison, the earthquake induced displacements for 0.04g and

0.81g PGA are very small. From the deflected shapes, it appears that similar form of bending is

occurring in all three cases of lateral loading, and therefore the stress distribution induced in the

structures for all three lateral loads will be similar in distribution. However, they will differ in relative

magnitude to each other, with the largest stresses induced by the greatest lateral loads. A tabular output

for the stresses in the shear core or specific structural joints would be too lengthy to include in a

succinct report due to the. As wind is the prevailing lateral load scenario the discussion will focus on

the prevailing outcomes of this form of lateral loading.

Isolating the shear core we can display the maximum axial forces induced during gravity, wind, and

combined loading. Figure 43 shows the deflected form of the structure in the YZ plane under

deformation with the axial forces induced on the shear core in 103kN (see legend). It is evident that the

entire shear core is in axial compression from the evenly distributed gravity loads, with the largest axial

forces occurring at the base of the structure. This result is expected, and is due to the accumulated dead

weight and live occupancy loads occurring at each floor. Thus, axial load at each storey is inversely

proportional the height of the building.

For the wind only case, the core predominantly exhibits the behaviour of a flexural cantilever (evidentfrom the curved profile of deformation). The lateral load applied form the Y direction induces tension

on the windward face, resulting in the positive value for axial forces on that side of the structure,

(extension of that side), whilst the leeward side of the core acts under compression which exhibits the

positive axial force (shortening of that side). Furthermore, the neutral axis (region of zero internal

stress) appears to pass directly through the centre of the core. The basis of structural systems resisting

lateral load can be defined as shown in figure 42 for the hypothetical 7 storey structure.

The transfer of lateral loads acting on the building to the foundation is an action comprised of

transferring the lateral shear. The lateral shear is the sum of the lateral forces imposed on the structureabove the storey that is being considered. Figure 42 demonstrates that lateral shear at a storey decreases

with height from ground, and that the overturning moment is the sum product of lateral forces above the

storey being considered, multiplied by the distance to load from the storey being considered. It is the

overturning moment which puts the leeward side fibres in compression and the windward side fibres in

tension, and exhibits the flexural cantilever deformation of the structure. The overall moment of the

lateral load is resisted at each storey by the couple resulting from the axial and compressive forces in

the fibres and/or columns on opposing sides of the structure.

In the combined loads of figure 43, the forces in the shear wall which induce deformation are exhibiting

the desired performance of resistance to lateral load. The shear core, through its high in-plane stiffness

and strength, is able to carry the gravity loads from the floor slabs at the interconnected storey levels,

but also uses the attracted gravity load induced forces to suppress the maximum tensile bending stresses

in the windward side of the lateral load. It can be seen that the Gravity only axial force of on the bottom

TensionSide

NA

Compression

side

Resulting flexural deformationApplied lateral load

Figure 42 Concept of transferring lateral load and flexural deformation of vertical cantilever

F1

F2

F3

F4

F5

F6

F7

Where h5 = distance

from floor 5 to floor i

Level 4

Level 5



55

Gravity load Unfavourable wind

Figure 43 – Maximum axial forces in shear core

Gravity and Wind Load

right hand corner of the shear core aid in reducing the high tensile stress under wind only loading, to a

relatively low tensile stress.

It is qualitatively evident in the core that it is behaving as a flexural cantilever. The cantilever action is

due to the bending of the core wall. It is known that the relative deflection of the storey will be

dependent on the flexural rigidity (EI) of the element. Conversely, the load attracted by the vertical

element will depend on the flexural stiffness k(EI/L). As the vertical elements between each floor are

the same for the shear core, then there is a constant value for I, and therefore the load resisted by each

member will be proportional to I. Due to the constant I all the way along the shear core for A4, the

relatively proportional distribution of bending forces are evident.



56

As previously discussed, there is also a shear component to the deformations of the core, but this is due

to the lateral load resisted by shear action in the cantilever action. The relative deflection of each storey

in this case is primarily dependent on the axial rigidity of the structure (EA). This however becomes,

the predominant mode in cases of squat wall structures, where the length of the wall is relatively large

compared with the height of the wall. As this is not the case with A4‘s core, this deformation mode isnot prevalent.

Although the predominant behaviour shows that the majority of the lateral load is resisted by the shear

core, there is likely a frame action contribution from the external columns and adjoining slabs. This is

anticipated to be the location as shown in figure 44 for the lateral loading, where there is a change in the

overturning moment. At the point of change in moment, this is the point of contraflexure where zero

overturning moment is present. It can be seen for the wind only loading of the core that this anticipated

point in which the axial forces flip is likely the region of change from flexural shear wall deformation to

predominant shear deformation from the frame component. This appears to be occurring at the level of

storeys 19-20.

As shown previously, a rigid frame under lateral loads predominantly exhibits shear deformation i.e.

side-sway. The inter-storey drift should begin to diminish with the shear deformation of a rigid frame

structure, as the lateral shear at a storey decreases with height from ground (due to lateral shear at a

storey being the sum of the lateral forces imposed on the structure above the storey that is beingconsidered). As the prevailing mode of deformation is due to shear, the greatest inter-storey drift will

occur at the base, and diminish as height increases. Therefore, if the region above the anticipated point

of contraflexure is deforming predominantly in shear action, then the inter-storey drift, dr, under lateral

loading should begin to diminish beyond the point of contraflexure. In determining the overall

deformation and plotting to show the inter-storey drift and inter-storey drift ratios, this point of

contraflexure can be determined.

The overall lateral deformations are as shown in table 8. It is evident that there is a similar trend of

lateral deformation for all load combinations. Although favourable and unfavourable design scenarios

have been considered, due to the linear elastic nature of analysis, the lateral deformations for the

unfavourable load cases are simply 1.5 times the magnitude of the favourable deformations. This is due

to partial factor of 1.5 applied to the unfavourable loads. Hence, for the rest of the evaluation, only theunfavourable load combinations will be considered as they exhibit the largest deformations.

Flexural

deformationshape of shear

wall component

Shear deformation

shape of Frame

component

Point of

contraflexure

Typical wall-frame interaction

Typical wall-frame deformation

Anticipatedpoint of

contraflexure

Figure 44 - anticipated region of wall-frame interaction



57

The Graphs in figure 45 and 46 indicate the profile of the structures overall deformation under overall

lateral deformation (lateral storey drift). It is evident that the largest lateral deformation occurs due to

wind loading, and that both seismic load cases are relatively small. Furthermore, it confirms that the

structure is undergoing deformation of a vertical cantilever, up until around storey 20 (as is evident

from the curved profile of the graph to that point). From then on, the relationship of overall drift is

somewhat linear. The top deformation for unfavourable wind is Δtop = 0.068m. The allowable total top

deflection from section is H/500 = 82.64/500 = 0.165m. Therefore the total deflection of the structure

under unfavourable wind load does not approach even half the SLS lateral deflection limit, and thus

satisfies the SLS condition. The maximum top deflections for seismicity are negligible, and for the

worst case seismic loading of 0.18g PGA the Δtop = 0.010m, and for PGA 0.04g Δtop = 0.0035m which

are 15% and 5.2% of the magnitude of the wind-induced lateral deflection respectively.

Table 8

Overall storey deflections, ∆(mm)

Storey height (m)Storey No.

Gravity + WindGravity + Seismic 0.04g Gravity + Seismic 0.18g

unfav. favourable unfav. favourable unfav. favourable

0.00 0 0.00 0.00 0.00 0.00 0.00 0.00

8.00 1 2.82 1.88 0.30 0.20 0.63 0.42

11.11 2 4.38 2.92 0.38 0.25 0.88 0.59

14.22 3 6.17 4.11 0.47 0.31 1.16 0.77

17.33 4 8.17 5.45 0.57 0.38 1.47 0.98

20.44 5 10.35 6.90 0.65 0.43 1.79 1.19

23.55 6 12.69 8.46 0.77 0.51 2.13 1.42

26.66 7 15.17 10.11 0.89 0.59 2.48 1.65

29.77 8 17.78 11.85 1.01 0.67 2.83 1.89

32.88 9 20.48 13.65 1.13 0.75 3.19 2.13

35.99 10 23.27 15.51 1.26 0.84 3.56 2.37

39.10 11 26.13 17.42 1.38 0.92 3.94 2.63

42.21 12 29.05 19.37 1.51 1.01 4.32 2.88

45.32 13 32.02 21.35 1.64 1.09 4.72 3.15

48.43 14 35.01 23.34 1.77 1.18 5.12 3.41

51.54 15 38.03 25.35 1.90 1.27 5.53 3.69

54.65 16 41.06 27.37 2.04 1.36 5.95 3.97

57.76 17 44.09 29.39 2.19 1.46 6.38 4.25

60.87 18 47.12 31.41 2.33 1.55 6.82 4.5563.98 19 50.15 33.43 2.48 1.65 7.27 4.85

67.09 20 53.16 35.44 2.63 1.75 7.73 5.15

70.20 21 56.15 37.43 2.79 1.86 8.19 5.46

73.31 22 59.13 39.42 2.95 1.97 8.67 5.78

76.42 23 62.09 41.39 3.11 2.07 9.15 6.10

79.53 24 65.02 43.35 3.26 2.17 9.62 6.41

82.64 25 67.92 45.28 3.52 2.35 10.21 6.81



58

Figure 45 - Lateral storey drift for Unfavourable loading

Figure 46 - Lateral storey drift for favourable loading

0

5

10

15

20

25

0 10 20 30 40 50 60 70 80

S t o r e y n u m b e r

Lateral storey drift (mm)

Wind

unfavourable

Seismic 0.04g

unfavourable

Seismic 0.08g

unfavourable

0

5

10

15

20

25

30

0 10 20 30 40 50


Storey drift (mm)

Wind

favourable

Seismic 0.04g

favourable

Seismic 0.18g

Favourable



59

Figure 47 - Inter-storey drift for unfavourable loading

Figure 48 - Inter-storey drift ratio for unfavourable loading

0

5

10

15

20

25

30

0.00 0.50 1.00 1.50 2.00 2.50 3.00 3.50


Inter-storey drift (mm)

Wind

unfavourable

Seismic 0.04g

unfavourable

Seismic 0.18g

unfavourable

0

5

10

15

20

25

30

0 5 10 15 20 25


Inter-storey drift ratio (%)

Wind Y

unfavourable

Seismic 0.04g

unfavourable

Seismic 0.18g

unfavourable



60

Graph 47 indicates the interstorey drift, whilst graph 48 indicates the inter-storey drift ratio. The inter-

storey seismic drift limit dr for the two storey heights of 3.11m (for residential) and 8.11 (ground) are

0.031m and 0.08m respectively. The seismic dr equates to the same h/100 limit for wind induced

interstorey drift. From the graph, it is evident that, the maximum inter-storey drift occurs at the 20th

storey, with a value of 3.30mm for wind loading, which does not exceed the values for interstorey-drift

limits. From this maximum point, inter-storey drift begins to diminish. Therefore, all inter-storey drift

values fall within their respective ranges of acceptable SLS limits. It can therefore be confirmed that

can be confirmed that seismic loading does not govern (as expected), and poses no threat on the SLS

performance of the structure.

Therefore Hence, for a typical structure London based high-rise structure for the 2500 year return

period PGA of 0.04g and the hypothetical 0.18g PGA for exceptional exceedance, the consideration

into seismicity in UK design can be deemed unnecessary. Even if the worst case wind and seismic load

was to be superimposed in an adverse situation of unfavourable wind and seismicity acting together,

the total top deflection would equate to Δwind + Δseismic, 0.18g = 0.068m + 0.010m = 0.078m. This is still

well within the allowable deflection limit, and means that the structure‘s global behaviour is acceptable

for lateral deflections at this stage.

Table 9 - Inter-storey drift

Inter-storey drift, dr (mm)

Storey height

(m)

Storey No. Gravity + Wind Gravity + Seismic 0.04g

unfav.

Gravity + Seismic 0.18g

unfav.unfav.

0.00 0 0.00 0.00 0.00

8.00 1 1.56 0.30 0.63

11.11 2 1.79 0.08 0.25

14.22 3 2.00 0.09 0.28

17.33 4 2.00 0.10 0.31

20.44 5 2.18 0.08 0.32

23.55 6 2.34 0.12 0.34

26.66 7 2.48 0.12 0.35

29.77 8 2.61 0.12 0.35

32.88 9 2.70 0.12 0.36

35.99 10 2.79 0.13 0.37

39.10 11 2.86 0.12 0.38

42.21 12 2.92 0.13 0.38

45.32 13 2.97 0.13 0.40

48.4314 2.99 0.13 0.4051.54 15 3.02 0.13 0.41

54.65 16 3.03 0.14 0.42

57.76 17 3.03 0.15 0.43

60.87 18 3.03 0.14 0.44

63.98 19 3.03 0.15 0.45

67.09 20 3.01 0.15 0.46

70.20 21 2.99 0.16 0.46

73.31 22 2.98 0.16 0.48

76.42 23 2.96 0.16 0.48

79.53 24 2.93 0.15 0.4782.64 25 2.90 0.26 0.59



61

The highlighted region in table 9 is the point at which the values for inter-storey drift begin to diminish.

At storeys 19-20, the maximum value is approached for inter-storey drift, and is also the point where no

the overturning moment in the shear-core occurs. This is therefore the indicative zone of contraflexure,

where the change from predominant flexural deformation the shear wall to the shear core changes to

predominant deformation of the rigid frame component. This can be qualitatively observed by the

increasing dr values up until the 19-20th

storey in graph 47, and then prompted by the receding values

for on the curve that is typical of a shear deformation.

Another observable phenomenon (particularly in seismic design) is the large values of inter-storey drift

ratio for the lowest storey. This is due to the soft-storey phenomena, which in the case of A4 is induced

by larger height of that floor (8m). Therefore the concentration of mass at that point is at a different

distance from the other floor masses, and therefore the stiffness of that section is inherently different.

The ―soft-storey‖ is however a concern only in high seismicity seismic events, as the inherent

difference in stiffness may initiate a storey-collapse mechanism. Stiffness soft story irregularity isdefined to exist where there is a story in which the lateral stiffness is less than 70% of that in the story

above or less than 80% of the average stiffness of the three stories above (Taranath, 2010). However,

this effect can be excluded for the future design of A4 as it is negligible.

6.1 Comfortability criteria performance

The dynamic excitation of a structure can induce particular human responses which may result in the

successful end-use of the structure being hindered, and possibly render the structure uneconomical.

Earthquake induced motions, even when they are more violent than those induced by wind, evoke acompletely different human response. This is firstly due to the inherent infrequency of earthquakes

relative to high winds/storms, and secondly, because the duration of motion caused by an earthquake is

generally short. The psychological effect experienced by people during earthquakes are inclined to be

less critical of the building motion, due to the gratefulness of the occupants to have survived such a

natural disaster. Earthquake-induced motions are, therefore, a safety rather than a human discomfort

issue. However, the negligible earthquake effects on A4 will likely not induce this feeling in people.

Although no specific code guidance is given for comfortability criteria calculations, an empirically

based check will be performed to determine the cross-wind and along-wind response. These are based

on a simplified calculation for A4 tower, assuming that it is rectangular in cross-wind and along-wind

profile. For full reference and theory of the calculation, please refer to ‗Dynamic Response to wind

loading‘ in ‗Tall Building Structures: Analysis and Design‘ (Smith, 1991), Where the building has an aspect ratio W/H = 0.424, for a city centre the roughness factor of the surface

is r = 0.425 (Smith, 1991), and a background turbulence factor of B = 0.75

A reduced frequency is then applied where

A size reduction factor, S, is then applied where S = 0.04. The inverse wavelength is = 0.0173



62

From this, a gust energy ratio F, is applied, where F = 0.15

Finally, a resonant turbulence factor

From this, the formula to denote average fluctuation rate,

A peak factor gp= 3.75, and therefore Gust factor = 2.71

The peak dynamic forces and displacements are obtained by multiplying all static values due to the

mean velocity of 22m/s by 2.85.

Therefore, along wind acceleration, is denoted by the formula

The value of 0.955ms

-2, satisfies the criteria of low acceleration, where in section 3.7, it can be classed

as ‗sensible, people can perceive‘. However, this is a small acceleration, and therefore should notinduce discomforting accelerations. It is recommended that a more accurate method such as wind-

tunnel testing be carried out at the final design stage to ensure an accurate representation of the wind

induced accelerations (Smith 1991)

.

6.2 Design implications

The results indicate that performance objectives are met by A4 tower in the case of the applied

loadings, and that there is no need to explicitly consider seismic action due to the minimal impact that it

will have on the structures serviceability performance. Therefore, standard detailed design procedures

to EC2 part 1-1 will suffice for the detailed design stage of the project life cycle.

However the factors affecting structural performance covered only one of the many key aspects of high-

rise building design. Further design will need to give consideration into the following factors. The

structure will in essence provide a frame to support substantial gravity loading through structural dead

weight and occupancy loads. It is evident that a high-rise RC structure will have a significant self-

weight, and over time this will have cumulative effects on the gravity load resisting members (columns

and shear walls).

6.2.1 P-Δ effects

As has been demonstrated in this report, An important characteristic of very tall buildings when

subjected to lateral loads is the expected magnitudes of lateral drift. This is expected due to the fact thatfor a uniform load on a uniform building with a constant stiffness up the height of the building the

lateral deflection is proportional to the fourth power of the height H4. This introduces the

importance of including P-Δ effects into the detailed lateral load analysis of tall buildings.

The P-Δ effect is the creation of additional moments and lateral deflections caused by the eccentricity of gravity loads through lateral deformation of the structure. In terms of ultimate limit state design, lateral

deflections must be limited to prevent this second order P-Δ effect, due to gravity loadings in high-rise

buildings being of such a magnitude as to initiate collapse. To satisfy serviceability limit state design,

deflections must be limited to levels such that proper functioning of non-structural members (and other

criteria listed in section 3.4). Furthermore, P-Δ effects may lead to undesirable dynamic characteristics,

causing further discomfort to building occupants.



63

As the lateral deformations for A4 fall within the range of acceptable deflections, it is not expected that

significant P-Δ effects should occur under gravity loading design of members. However this can only bediscerned through more detailed and final design and analysis. As is often the case in many ongoing

structural design projects, the client and architectural needs may change with time, thus imposing

changes that must be made to the structural layout. Therefore, should any changes be made, the same

procedures as above, plus those listed here should be made, and analysed thoroughly.

6.2.2 Axial shortening

With increasing height of tall building structures, axial shortening of columns and wall members can

become very significant, compared to that of low-rise buildings. The factors affecting this are:

Elements which are subjected to higher stress levels due to accumulated loads from many floors

with the added pressure of minimising cross sectional areas with the use of higher strength

concretes.

Accumulation of strains over longer lengths corresponding to the height of the building.

It is important to recognise that shrinkage and creep is time dependant, and that creep is also stress

dependant. Hypothetically, if all vertical elements were subjected to constant stress and comprised of identical concrete mixtures and geometrical shapes the building would shorten uniformly. Although

some limitation would have to be placed on the total allowable shortening it should not cause too great

a concern. Unfortunately this idealised situation is never the case (although the large part of A4 is

uniform in section sizes and concrete type). It is the differential shortening between adjoining vertical

elements that tends to cause problems. The magnitude of this movement can be enough to cause distress

to non-structural elements and induce moments and shear forces into connecting horizontal elements.

Axial shortening is made up from the following constituent parts:

Elastic Shortening - Short term strain from applied stress

Creep – Long term gradually increasing strain from applied stress

Shrinkage – long term strain caused by moisture evaporation from the surface

The calculation of expected axial shortening will require significant engineering judgement. The

influence of construction time and sequence is an important factor, and will play a significant role in the

construction of A4

Another assessment that is vital to design of high-rise structures is for movement due to temperature

effects. This will result in thermal loads being applied to the structure, and can result in further

differential shortening or movement of the structure. However, in the case of A4 it should pose no issue

to ignore this, as the structure will fall within the insulated body of the cladding and therefore negligible

differential movement can be expected.

7 Conclusions

Under typical UK design procedures utilising relevant Eurocodes, the performance based design of a 26

storey high-rise RC building located in central London is still wind-controlled. Static wind loading

effects have been compared with seismic multimodal response spectrum analysis for a 2500 year return

period PGA of 0.04g and a hypothetical worst case scenario PGA of 0.18g, and shown to prevail in

magnitude of lateral effects. Neither of the lateral load deformations exceeds the defined serviceability

limit states for inter-storey drift. Furthermore, seismic inter-storey drifts have been found to be a very

small proportion of the magnitude of inter-storey drift due to wind loading to Eurocode 1. The adopted

design has shown to be adequate in satisfying serviceability limit states defined for lateral deflection

limits, so as to prevent damage and/or stressing of non-structural elements. Furthermore, no ductility

specific detailing requirements would be necessary for the final design stages of the proposed project,

as the low seismic activity and EC8 specified ‗ductility class low‘ requirement states that generaldetailing to EC2 part 1-1 is applicable to ensure satisfactory lateral load performance.



64

However, since the method employed to determine seismic loading was using a simplified response

spectrum method in one predominant direction, a more accurate representation to confirm the findings

is suggested. This is as it cannot be said for certain whether the seismic load will not inhibit larger

deformations or even take control in some instances. To ensure confidence in findings, it is suggested

that a non-linear time-history analysis be undertaken to provide any information regarding the non-

linear behaviour that response spectrum analysis cannot provide. Furthermore, for a more detailed and

accurate lateral performance investigation, lateral loads in more than one direction should be

considered. It must be noted that earthquakes are highly unpredictable, and it is very difficult to

ascertain the size of future seismic events. Therefore, there is still desirability for seismic analysis of

structures within the UK.

The analysis of the structure has shown that the predominant resistance to lateral load has been

provided through flexural bending resistance of the shear core. The grouped behaviour of wind and

gravity loading upon members reduces the net tensile forces imposed on the structure during lateral

bending. The analysis has also successfully confirmed the point of shear-wall frame interaction for the

proposed design, the corresponding changes in the form of lateral deformation, and demonstrated the

ability of the external frame components to potentially resist lateral load. Therefore, it is suggested inthe instance of detailed design for individual structural members that this shear wall-frame interaction

be considered. The typical procedure of only designing the shear-core to resist lateral load prevents the

advantages on detailed design imposed by considering the wall-frame interaction, where:

The estimated storey drift may be less than if the walls were considered to be the only

horizontal load resisting members.

The estimated bending moments in the walls/cores will be less if than if only wall action is

considered alone.

The columns of the frames can be designed as fully braced, limiting requirements to design for

considerable secondary effects (i.e. PΔ effect) The estimated shear in the frame will likely be uniform throughout the height, and as a

consequence of this can be designed and constructed repetitively and economically.

Preliminary calculation of wind induced acceleration of the structure has shown that comfortability

criteria for human perception has not been exceeded. However, to truly ascertain this, it is suggested

that in place of the conservative code based wind load calculations (for which A4 is outside of the

traditional scope) a specialist wind-tunnel test is undertaken. The use of a wind model for tall buildings

typically offers significant advantages over a conservative use of the code, as it:

Provides an accurate distribution of wind loads, especially for structures in a built-up

environment, as results can directly determine the impact of wind load on surrounding

structures as well.

Provides predictions of wind-induced building motions (accelerations and torsional

velocities) likely to be experienced by occupants of the top floors, and compares the

test results to available serviceability criteria.

Provides an assessment of expected pedestrian wind comfort along with any conceptual

recommendations for improvement to key pedestrian areas (e.g., main entrances,

congregational areas, etc.).

Because wind-tunnel studies consider the effect of nearby buildings and directional

variations in the local wind climate, the overall design wind loads are generally (but not

always lower than code wind loads resulting in lower cost in construction due to lack of

over reinforcing.



65

8 References

Main literature

British standards institution (2010) ‗BS EN 1991-1-1:2002 Eurocode 1: Actions on structures – Part 1: General actions — Densities, self-weight, imposed loads for buildings‘. British

standards institution, London.

British standards institution (2010) ‗BS EN 1991-1-4:2002 Eurocode 1: Actions on structures – Part 1: General actions — Wind actions‘. British standards institution, London.

British standards institution (2002) ‗UK National Annex to BS EN 1991-1-1:2002‘. British

standards institution, London.

British standards institution (2008) ‗BS EN 1998-1-1:2004 Eurocode 2: design of concrete

structures – Part 1: general rules for building‘. British standards institution, London. British standards institution (2010) ‗BS EN 1998-1-1:2004 Eurocode 8: design of Structures for

Earthquake resistance – Part 1: general rules, seismic actions, and rules for buildings‘, Britishstandards institution, London.

Institution of Structural Engineers (2010) ‗ Manual for the seismic design of steel and concretebuildings to Eurocode 8‘. Institution of Structural Engineers, London

Institution of Structural Engineers (2010) ‗ Manual for the design of building structures toEurocode 1 and Basis of structural design‘. Institution of Structural Engineers, London.

Institution of Structural Engineers (Manual for the design of concrete building structures to

Eurocode 2‘. Institution of Structural Engineers, London.

Other references

Bird, J.F. and Bommer, J.J. (2004) ‗Earthquake losses due to ground failure‘, EngineeringGeology, 75:147 – 179.

Bommer and Stafford (2009) ‗Seismic Hazard and earthquake actions‘ in Elghazouli, A.Y(2009) ‗ Seismic design of Buildings to Eurocode 8‘, Spon press (1st

edition), New York, pp.

Building Research Establishment (1994) ‗Wind around tall buildings‘ in ‗BRE Digest 390‘.Watford: Building Research Establishment

British Geological Society (2009) ‗NOTES ON INDIVIDUAL EARTHQUAKES‘. BritishGeological Society. Available at

http://www.quakes.bgs.ac.uk/earthquakes/historical/historical_listing.htm#PageTop [Accessed

at 20th

March 2011]

Cobb, Fiona (2009) ‗ Exposure classification and recommendation for resisting corrosion of reinforcement‘ in ‗Structural Engineers Pocketbook‖, Elseiver press (3rd

edition), Oxford,

pp.185-186 Computers and Structures Inc. (2009) ‗Diaphragm Constraint‘ in ‗CSI Analysis Reference

Manual For SAP2000, ETABS, and SAFE‘. Berkeley, California. pp. 53-54

Eurocodesexpert.co.uk (2007) ‗Eurocodes‘. Available at:

http://www.eurocodes.co.uk/Content.aspx?ContentId=4 [accessed at 4th

March 2011)

Smith, B. S. (1991). ‗Core Structures‘, in ‗Tall Building Structures: Analysis and Design‘.Canada; John Wiley and Sons, pp. 308-309

Computers and Structures Inc. (2009) ‗Modal Analysis‘ in ‗CSI Analysis Reference ManualFor SAP2000, ETABS, and SAFE‘. Berkeley, California. pp. 348-350

Institution of Structural Engineers (2006) ‗Sizes of reinforcement and columns‘ in ‗Manual for the design of concrete building structures to Eurocode 2‖, Institution of Structural Engineers,London, pp.14-17

Cobb, Fiona (2009) ‗ Preliminary sizing of concrete elements‘ in ‗Structural EngineersPocketbook‖, Elseiver press (3

rdedition), Oxford, pp.187

http://www.quakes.bgs.ac.uk/earthquakes/historical/historical_listing.htm#PageTop


http://www.eurocodes.co.uk/Content.aspx?ContentId=4






Hira, A (2003) ―Lecture 3 – Loads and Design criteria for tall buildings‖ in ―421-415 Design of

High-rise Structures‖, University of Melbourne, Australia. Lin, Liwei (2005) ‗Introduction to Finite Element Modelling‘ University of California at

Berkeley, College of Engineering, Mechanical Engineering Department. Available at:

http://www.me.berkeley.edu/~lwlin/me128/FEMNotes.pdf [accessed at 13th

March 2011].

Musson, R (2003) ‗Seismicity and earthquake hazard in the UK‘, British Geological Survey.Available at: http://www.earthquakes.bgs.ac.uk/hazard/Hazard_UK.htm [Accessed at 20th

March 2011]

Musson, R. and Sargeant, S. (2007) ‗Eurocode 8 seismic hazard zoning maps for the UK‘.British Geological Survey Seismology and Geomagnetism Programme Technical Report

CR/07/125, 2007. Available at: http://www.seced.org.uk/news/UK_seismic_hazard_report-

issue3.pdf [Accessed at 20th

March 2011]

Smith, B. S. (1991). ‗Dynamic Response to wind loading‘, in ‗Tall Building Structur es:

Analysis and Design‘. Canada; John Wiley and Sons, pp. 422-431

Smith, B. S. (1991). ‗Rigid Frame-Structures‘, in ‗Tall Building Structures: Analysis andDesign‘. Canada; John Wiley and Sons, pp. 131-168

Taranath, B.S. (2010) ‗Variation of Wind Velocity with Height (Velocity Profile)‘ in

‗Reinforced Concrete Design of Tall Buildings‘. Florida: CRC Press, pp. 257-260 Taranath, B.S. (2010) ‗ Seismic Design‘ in ‗ Reinforced Concrete Design of Tall Buildings‘.

Florida: CRC Press, pp. 350-355

Taranath, B.S. (2010) ‗Lateral Load-Resisting Systems‘ in ‗Reinforced Concrete Design of Tall Buildings‘. Florida: CRC Press, pp. 32-40

Williams, M.S. (2009) ‗Structural Analysis‘ in Elghazouli, A.Y (2009) ‗Seismic design of Buildings to Eurocode 8‘, Spon press (1st

edition), New York, pp.49-52

Widas, Peter (1997) ‗Introduction to Finite Element Analysis‘, Virginia Tech University,Virginia Tech Material and Science Engineering. Available at:

http://www.sv.vt.edu/classes/MSE2094_NoteBook/97ClassProj/num/widas/history.html

[Accessed at 14th March 2011]

http://www.me.berkeley.edu/~lwlin/me128/FEMNotes.pdf


http://www.earthquakes.bgs.ac.uk/hazard/Hazard_UK.htm


http://www.seced.org.uk/news/UK_seismic_hazard_report-issue3.pdf










CE3099 Individual Project Bekar Bedir 0707295 MEng

Documents

Transcript of CE3099 Individual Project Bekar Bedir 0707295 MEng