TECHNICAL REPORT S-74-3

ANALYSIS OF PILE TESTSby

W. C. Sherman, Jr., D. M. Holloway, C. C. Trahan

TA7.W34tS-74-31974

April I974

Sponsored by Office, Chief of Engineers, U. S. Army

Conducted by U. S. Army Engineer Waterways Experiment Station

Soils and Pavements Laboratory

Vicksburg, Mississippi

APPROVED FOR PUBLIC RELEASE; DISTRIBUTION UNLIMITED

LIBRA Rv

MAR 6 75Bureau or Reclamsfi-sfl

Denver, Cd&rs&Q

Destroy this report when no longer needed. Do not return it to the originator.

The findings in this report are not to be construed as an offic Department of the Army position unless so designated

by other authorized documents.

> 4

) BUREAU OF RECLAMATION DEN)

N?( V

92067275

0 .TECHNICAL^PORT J-74-3 ^

ANALYSIS OF PILE TESTS^ by

W. C. Sherman, Jr., D. M. Holloway, C. C. Trahan

Mm,

IDI

ç==?I—tf in. Ii f i

IDI

IDI

April 1974ySponsored by Office, Chief of Engineers, U. S. Army

Conducted by U. S. Army Engineer Waterways Experiment Station

Soils and Pavements Laboratory Vicksburg, Mississippi

A R M Y - M R C V I C K S B U R G . M I S S

APPROVED FOR PUBLIC RELEASE; DISTRIBUTION UNLIMITED

92067275

THE CONTENTS OF THIS REPORT ARE NOT TO BE USED FOR ADVERTISING, PUBLICATION, OR PROMOTIONAL PURPOSES. CITATION OF TRADE NAMES DOES NOT CONSTITUTE AN OFFICIAL EN­DORSEMENT OR APPROVAL OF THE USE OF SUCH

COMMERCIAL PRODUCTS.

iii

FOREWORD

The study described in this report was initiated as a Civil Works Engineering Study, ES 038, entitled "Analysis of Pile Tests," and sub­sequently incorporated into CWIS 31203 "Analyses of Structure and Foun­dation Interaction." It was conducted by the U. S. Army Engineer Waterways Experiment Station (WES), under the sponsorship of the Office, Chief of Engineers. The initial phase of this study involved collecting information on pile load tests conducted by Corps of Engineers offices. Subsequent work involved analysis of the data.

Appreciation is expressed to those Corps of Engineers offices that replied to the request for information.

The compilation of data was accomplished by Messrs. J. L. McCall and C. R. Furlow. The data were analyzed and this report was prepared by Messrs. W. C. Sherman, Jr., D. M. Holloway, and C. C. Trahan under the general supervision of Mr. J. P. Sale, Chief, Soils and Pavements Laboratory, WES.

Directors of WES during the preparation and publication of this report were COL L. A. Brown, CE, BG E. D. Peixotto, CE, and COL G. H. Hilt, CE. Technical Director was Mr. F. R. Brown.

v

CONTENTSPage

FOREWORD...................................................... vNOTATION...................................................... ixCONVERSION FACTORS, BRITISH TO METRIC UNITS OF MEASUREMENT . . . xiS U M M A R Y ...................................................... xiiiPART I: INTRODUCTION........................................ 1

Purpose of Study........................................ 1S c o p e .................................................. 1Review of Pile Load Test D a t a .......................... 2

PART II: PILE LOAD T E S T S .................................... 5Conduct of Load T e s t s .................................. 5Time Effects............................................ 5Effects of Water Table .................................. 8Distribution of Load in P i l e ............................ 8Residual Load in Piles.................................. 11Calculation of Failure Loads ............................ 12

PART III: ANALYTICAL DETERMINATION OF PILE BEARING CAPACITYIN COHESIONLESS SOILS.............................. l6

Basic Concepts............... 16Tip Capacity................................... 17Skin Friction.......................................... 31

PART IV: PILE TESTS IN COHESIONLESS SOILS 36CE Design Procedure .................................... 36Data from CE Pile T e s t s ................................ 36Distribution of Load in P i l e ............................ 39Unit Skin Friction...................................... b5Tip Capacity................................. 51

PART V: PILE TESTS IN COHESIVE S O I L S ........................ 52Basic Concepts.......................................... 52Data from CE Pile T e s t s ......................... 52Soft or Firm Clays...................................... 52Stiff or Hard C l a y s .................................... 5bLong-Term Capacity ...................................... 57

vii

CONTENTS

PagePART VI: CONCLUSIONS AND RECOMMENDATIONS...................... 58

Conclusions.............................................. 58Recommendations.......................................... 58

LITERATURE CITED................................................ 60TABLES 1-8APPENDIX A: COMPILATION OF D A T A ................................ A1TABLES A1-A8APPENDIX B: INSTRUMENTATION FOR PILE LOAD TESTS................ B1

Introduction............... B1Pile Load Measurements .................................... B1Movement of Pile B u t t .................................... B3Load Transfer T e s t ........................................ B5Lateral Load Tests.......................... B13

TABLE B1

viii

NOTATION

AAsAtB

c andc T and ft'

caC

d ,d c ? q 7D

DrDwe

Efsfs

FIk

K-

K

KsL

Nc,N ,N5 q» y

Cross-sectional area of the pile, sq in.Shaft area Tip areaWidth of the footing = 2BMohr-Coulomb strength parametersModified Mohr-Coulomb strength parametersAdhesion component of the shaft resistanceUndrained shear strength intercept R cohesionBearing capacity depth factorsDepth of pile penetrationRelative densityDepth of water tableDistance from the bending axis of the pile to the straingageModulus of elasticity of the pile, psi Unit skin friction at a given depth

Average unit skin friction Pile capacity ratioMoment of inertia of the pile about its bending axis Empirical compressibility factor Passive earth pressure coefficient

Coefficient of lateral earth pressure in compression

Coefficient of lateral earth pressure in tension Passive earth pressure coefficientPrimary bearing capacity factors for general shear failure

ix

N'c ,W ,N'* q 7N*q

pq

qfqoqz

qlQsQtSiSR

s ,s ,s c’ q5 7tzaP7

7 T5

AeAS€07aT

Primary bearing capacity factors for local shear failureEquivalent bearing capacity factor for deep circular foundationsUnit normal stress on the free surface Unit normal stress on the shaft surface Average load between two strain rod anchors, lb Surcharge pressureEffective vertical stress at failure Unit tip resistanceVertical overburden stress at depth zAverage vertical stress Equivalent overburden pressure Skin friction Tip resistanceUltimate pile capacity with the water table at depth DwUltimate pile capacity with the water table at ground surfaceCylindrical foundation base radius Bearing capacity shape factors Soil sensitivity ratio Depth below the soil surface Empirical adhesion coefficientAngle from the horizontal to the slip surface terminus Average effective unit weight of the overburden Average effective unit weight below water surface Pile-soil interface friction angle Vertical deformation between the rod anchors, in. Vertical distance between two rod anchors, in.Measured strain at the gage location Angle defined by the geometry of the failure zone Semiempirical bearing capacity shape factor Measured stress at the gage location Shear stress

x

CONVERSION FACTORS, BRITISH TO METRIC UNITS OF MEASUREMENT

B r it i s h u n its o f measurement used in t h i s re p o rt can be co n verted to

m e tric u n its as f o l lo w s :

______ M u ltip ly ______

in ches

f e e t

tons

pounds p e r square in ch

pounds p e r square fo o t

pounds p e r cu b ic fo o t

to n s p e r square fo o t

_______ ___________2.5I*

0.301*8

8.896W*0.6891*757

1*7.88028

16.0185

95.760567

_________ To O btain____ _____

cen tim e te rs

m eters

kilonew tons

newtons p e r square cen tim eter

newtons p e r square m eter

kilogram s p e r cu b ic m eter

kilonew tons p er square m eter

xi

SUMMARY

The purposes of this study were to evaluate pile load test data obtained from Corps of Engineers offices and to compare these data in light of analytical design methods for predicting pile load capacities. Though many tests.were performed* very few permitted a detailed analysis of the behavior of the pile-soil system. Carefully instrumented pile load tests* such as those performed at Old River Low-Sill Structure and Arkansas River Lock and Dam No. U* provided the only sources of data for which the pile-soil interaction could be examined in sufficient detail. It was found that the conventional static pile capacity formulas do not adequately describe the behavior of piles in cohesionless soils.

Load test results for piles in sands indicate that pile-soil in­teraction and soil compressibility in the vicinity of the tip may make the frictional resistance and tip resistance interdependent. The unit skin friction computed from field measurements tends to increase lin­early with depth only at shallow depths; thereafter it approaches a limiting value below a depth of 10 to 20 pile diameters. For tension piles* this limiting value remains essentially constant* whereas for compression piles* the unit skin friction decreases near the pile tip. Other investigators have reported similar observations. Extrapolation of field data to design pile foundations* based upon conventional meth­ods* may produce significant* unconservative errors.

For piles in soft to medium clays* the load tests indicate that the conventional methods of analysis using undrained shear strength are generally satisfactory and probably conservative for long-term behavior. Limited data for piles in stiff clays suggest many uncertainties in evaluating foundation performance. Time effects and related phenomena make these conditions most difficult to analyze.

The results of this study indicate that further research is nec­essary to provide clearer insight into the pile-soil interaction prob­lems. The behavior of piles in cohesionless soils deviates signifi­cantly from that predicted by conventional theories. In order to design pile foundations properly and interpret the results of pile load tests correctly* a more rational method of analysis is urgently needed.

xiii

ANALYSIS OF PILE TESTS

PAPT I: INTRODUCTION

Purpose of Study

1. The Corps of Engineers (CE) is responsible for the design of many structures that require the use of pile foundations to provide sup­port and to minimize objectionable settlements. Structural and economic considerations are causing a trend toward the use of high-capacity pil­ing, which requires more careful assessment of pile behavior under load. Because of the inherent uncertainties in designing pile foundations, the designs are often verified by pile load tests. In the past 30 to i)-0 years, CE offices have performed numerous pile load tests throughout the United States and in overseas areas. The pertinent data and anal­ysis of these tests are normally filed in the responsible office after the tests have served their intended purpose.

2. The specific purposes of this study were to:a. Compile and make available to CE offices the results of

these pile load tests.b. Review theoretical solutions for determining pile load

capacity.c . Compare the pile load test results with the theoretical

solutions.d. Develop improved methods for conducting pile load tests.e. Develop design guidelines.

The overall purpose was to develop empirical or theoretical solutions that will provide a better understanding of the mechanics of pile fail­ure, and thus possibly lead to a reduction in the need for future pile testing.

Scope

3. This report contains a compilation and analyses of static pile

1

load tests that have been conducted by CE offices • The data are pre­sented in tabular form in Appendix A, Factors affecting the conduct and analysis of pile load tests are reviewed. The analysis of data was lim­ited to tests on single vertical piles loaded in compression and tension. Load tests on instrumented piles in cohesionless soils provided the best source of detailed information; hence, greater emphasis was placed on these tests in the analysis.

Review of Pile Load Test Data

b. 3h the review of the pile load test data compiled in Appendix A, it was apparent that no consistent criteria were employed for deter­mination of the failure loads for single piles loaded in compression or tension. Since load-deformation data were not provided for many of the load tests 5 no uniform method of determining pile failure loads could be applied.

5. Many of the load tests were terminated when the loading reached a certain levels such as twice the design load, or when deflec­tion under the applied load reached prescribed limits. Therefore, ul­timate failure loads could not be determined for these piles. Of bl2 compression load tests performed on single vertical piles, 1k6 tests were carried to failure, 1 test showed an estimated failure load, and failure loads were obtained by extrapolation of the load-deformation curves in 6 tests. Forty-two tension load tests on single vertical piles out of 82 tests reported were carried to failure, while 3 of the b vertical load tests on pile groups were carried to failure.

6. Histograms of failure loads according to pile type for com­pression and tension tests are shown in fig. 1. These histograms dem­onstrate the extremely wide range of failure loads that may be developed for compression piles, with failure loads ranging trcsm less than 50 tons* to b^O tons. On the other hand, failure loads in tension are seldom

* A table of factors for converting British units of measurement to metric units is presented on page xi.

2

12

8

4

Oi

12

8

4

'Oi

16

12

8

4

O

12

TIMBER PILES (34 TESTS)

_J________I________i_350 400 450 500

TIMBER PILES (4 TESTS)

—H - n50 100 150 200

0250

T---------1---------1---------1---------1-------- 1---------1---------r~

STEEL H-PILES (32 TESTS)

50 100 150 200 250 300 350 400 450 500

I I

STEEL AND CONCRETE DISPLACEMENT PILES

(80 TESTS)

i_________L_200 250 300FAILURE LOAD, TONS

COMPRESSION TESTS TENSION TESTS

Fig. 1. Histograms of pile failure loads

NUM

BER

OF P

ILE

LOAD

TES

TS

greater than 150 tons. Generally, the timber piles are associated with the lower failure loads, while the steel H-piles and the steel and con­crete displacement piles are associated with the higher failure loads. The range of failure loads reflects the dominating influence of soil conditions on pile behavior.

7. Piles derive their support from the soil by skin friction along the embedded length of the pile, by end bearing on the pile tip, or by both.

8. A rational method for analyzing measured load capacities of test piles requires not only a detailed knowledge of the soil properties and groundwater conditions, but also adequate instrumentation to deter­mine the resistance distribution. Since these data were unavailable for the majority of the pile tests reported, not even an approximate analy­sis of these tests was possible. Therefore, the analysis was limited to those pile tests that were supported by appropriate data. Particular attention was given to analysis of data from instrumented test piles, since these provided the most enlightening information concerning pile behavior. A more comprehensive summary of the remainder of the pile tests included in fig. 1 is impractical, due to insufficient supporting data.

h

PART II: PILE LOAD TESTS

Conduct of Load Tests

9. The results of pile load tests can be affected considerably by the manner in which the tests are performed and the results are in­terpreted. Some of the factors that may affect the results include er­rors in load and deformation measurements, the time allowed between driving and load testing of the piles, the rate of load application, and the manner of load application. The depth of the groundwater level dur­ing testing and the procedures used in calculating the failure load also have an important bearing on the results. A rational analysis of pile behavior requires knowledge concerning the distribution of load along the length of the pile. To obtain this, various types of strain­measuring techniques may be employed, which in themselves may be sub­ject to significant errors. As will be subsequently discussed, residual loads induced by driving can seriously affect the interpretation of strain measurements along a pile. Proper procedures for conducting pile load tests are described in EM 1110-2-2906. Instrumentation required for proper interpretation of pile load tests is described in Appendix B.

Time Effects

Cohesionless soils10. The capacity of piles driven in sands and gravels generally

will not change with time. However, under some conditions, usually in­volving saturated fine or silty sands, the time effects may be signifi­cant. The dissipation of high negative pore water pressures induced bypile driving can account for significant changes in pile capacity with

2time. Terzaghi and Peck point out that occasionally the bearing ca­pacity of piles in sands decreases conspicuously during the first 2 or 3 days after driving. They state that the high initial bearing capacityis probably due to a tenporary state of stress that develops in the sand

3surrounding the point of the pile during driving. Feld describes the

5

case of l8-in.-diam pipe piles driven to depths of 50 to 60 ft intovarved silty sand overlain by about 10 ft of medium sand. The pileswithstood static loads of 120 to l6o tons. However, load tests made onthe same piles about a month after driving, and after additional pileshad been driven adjacent to them, showed excessive settlement underloads of only 80 tons. Parsons reports on two projects in the NewYork City area in which piles exhibited lower resistance at redriving.He referred to this phenomenon as "relaxation." Additional examples of

5relaxation are presented by Yang, who attributes the relaxation of pile resistance for piles driven into dense fine sands to the adjustment of the soil structure and the equalization of pore water pressure. He suggests that the driving resistance of piles in loose sands increases with time. Tavenas^ presents data for concrete piles driven into sands of medium density for which the bearing capacity showed a tendency to increase by about 70 percent in the first 2 or 3 weeks after driving.

11. Relaxation of pile resistance with time was noted at Jones- ville Lock and Dam during driving of piles at Monolith 1-L. The results of driving tests for two piles are shown in fig. 2. These data indicate

DRIVING RESISTANCE, BLOWS PER FOOT

Fig. 2. Effects of time on driving resistance of piles in sands at Jonesville Lock

6

that piles driven initially to refusal can subsequently be redriven with much lower driving resistances. It appears that the load-carrying ca­pacity of piles in dense sands can be substantially overestimated if the capacity is based on dynamic formulas and observed driving resistances.Such errors are not likely to occur if the capacity is based on static7formulas. Tomlinson suggests that because of time effects, load tests on piles in sands should not be made until at least k days after driving. Cohesive soils

12. Time effects for piles in cohesive soils are extremely im­portant and depend considerably on the nature of the soil, the type of pile, and other factors. Consequently, the period of time between driv­ing and load testing can have a great influence on the results. Gener­ally, the pile capacity will increase with time due to consolidation effects. As discussed in Part V, effective stresses in the soil imme­diately adjacent to the pile increase with time due to the decrease in excess pore pressures induced by pile driving. Field data which demon­strate that significant pore pressures can be generated in clays due togpile driving have been summarized by Horn. There appears to be no gen­eral agreement as to how these pressures can be used quantitatively inan engineering analysis. No data on pore pressures around piles were

2obtained on CE projects. Terzaghi and Peck present an example of an increase in bearing capacity with time that indicated development ofomaximum bearing capacity after about 1 month. Thorburn and MacVicar^ describe a case in which the length of time necessary to dissipate ex­cess pore water pressures around a pile in clayey silts was about 2 months. The time required to achieve complete re consolidation of the soil after driving is difficult to assess. Generally, on CE projects for load tests involving friction piles in clay, a waiting period of about 7 days has been specified. However, most of the clays have been relatively insensitive, and significant changes in strength with time may not have occurred. The rate at which test loads are applied can also have an important influence on the capacity. CE practice has been to increase test loads in increments, allowing each load to remain until

7

movement of the pile butt is essentially complete, before adding the next load increment.

Effects of Water Table

13. Both theory and experience indicate that the position of the water table can have a marked effect on the capacity of test piles, par­ticularly those installed in cohesionless materials. Ideally, the load tests should be conducted with the water table at ground surface since such a condition, if it should develop during the life of the structure, would result in the minimum bearing capacity. In practice, the water table during testing may be several feet or more below foundation grade, and it is then necessary to adjust the measured pile capacities for the influence of the lower water table. Examples of curves that may be used for adjusting pile capacities to the condition of zero groundwater depth are shown in figs. 3 and b. These curves are based on conventional static pile formulas that will be described subsequently. The impor­tance of measuring groundwater depths during driving and testing of test piles was brought out in the analysis of pile load tests for the Arkan­sas River lock and dam system.

lk. Another case in which groundwater levels are important occurs when piles are driven through a thick stratum of cohesive materials into sands. The pile capacity is dependent on the effective overburden pres­sure at the base of the clay stratum, which in turn is a function of the piezometric head in the sands. Where the sands interconnect with a river or other source with a varying head, and thus are subjected to varying piezometric heads with time, it is essential that proper consid­eration be given to this factor in evaluating the results of driving and load testing of piles.

Distribution of Load in File

15. Piles are frequently driven through soft compressible mate­rials into underlying stiff or dense materials of low compressibility.

8

WATER TA BLE AT GROUND SURFACEQu = y DAtNq + 1/2 y D2As K‘ TAN 8

WATER T A B LE AT DEPTH Dw

Qy = [yDyy + y '(D ~ Dyy[] AjNq +[1/2 r D * + y DW(D - Dw) + 1/2 y '(D - Dw)] As Kcs TAN 8

ASSUMED:

18-iN-SQUARE CONCRETE PILE

Nq = 21 = 0.92

s = 30°NOTE: CURVES OF F VS D SHOWN BELOW ARE VALID ONLY

FOR ASSUMED SOIL PROPERTIES AND PILE SIZESYMBOLS USED IN THIS AND SUBSEQUENT FIGURES ARE LISTED AND DEFINED ON PAGE ix

1.0 1.2 1.4 1.6 1.8 2.0 2.2_ PILE CAPACITY WITH WATER TABLE AT INDICATED DEPTH Qu

PILE CAPACITY WITH WATER TABLE AT GROUND SURFACE Q‘u

Fig. 3* Effects of groundwater table on compression testsin cohesionless soils

9

WATER TABLE AT GROUND SURFACEQJ, = 1/2 y>D2As K‘ TANS

WATER TABLE AT DEPTH Dw

Qu = [ l/ 2 y D 2 + rD w(D - Dw) + 1/2 y(D - Dw)2] As TAN

Qu 2D2 4DW(D - Dw) (0 - Dw)2 Q u = D2 + D2 + D2

NOTE: CURVES OF F VS D SHOWN BELOW ARE VALID FOR ANY STRAIGHT-SIDED PILE IN COHESIONLESS SOIL

S

1.0 1.2 1.4 1.6 1.8 2.0 2.2PILE CAPACITY WITH WATER TABLE AT INDICATED DEPTH Qu

PILE CAPACITY WITH WATER TABLE AT GROUND SURFACE

Fig. k. Effects of groundwater table on tension tests in cohesionless soils

10

Load tests of piles wider these conditions reflect the short-term carry­ing capacity of the compressible soils, -which cannot be cowited on for long-term support because of consolidation. Consequently, it is neces­sary to know the distribution of the applied load in the pile in order to determine that portion of the load carried by the incompressible stratum. The distribution of the load in the pile can be determined by means of the instruments that are described in Appendix B.

1 6 . In addition to the determination of short-term loads carried by compressible strata, instrumentation to determine the distribution of load in piles provides fundamental knowledge concerning the mechanics of load transfer in pile-soil systems, which is necessary for the proper interpretation of pile load test data. A critical review of the experi­mental field investigations and methods of analysis of load transfer from single piles and pile groups in various types of soils has been presented by Vesic."^° Much of this data is derived from pile load tests on CE projects. The observations show that the magnitude and distribu­tion of skin friction of piles change with the pile penetration and time, and depend also on the variability of the soil, the method of in­stalling the piles, and the complete stress history in both piles and soil. The development of reliable instrumentation methods has led to their extensive use on major pile test programs by the CE. The develop­ment of basic data concerning soil-pile interaction from these instru­mented pile tests forms the basis for much of the data presented in this report.

Residual Load in Piles

17. During pile driving operations, each hammer blow induces compressive strains in the pile. The rebound of the pile after each blow is partly restricted by negative skin friction along the upper part of the pile and by positive skin friction along the lower part. Conse­quently, residual compressive stresses are present in the driven pile.As pointed out by Davisson, much of the load transfer test data that is available in the literature is in error because of an oversight in

11

interpretation of the test data. The oversight is that readings are referenced to a zero load condition after the pile is in the ground, as­suming that the pile contains no residual loads.

18. The importance of residual loads was recognized during the12pile load tests for the Old River Low-Sill Structure in 1955* At­

tempts were made to measure residual loads by means of strain rods. However, despite all precautions , it was found that significant errors were introduced as a result of temperature effects and the manipulation of the strain rods before and after driving. Consequently, no accurate determination of the residual stresses after driving could be obtained. Residual loads were considered in the analysis of pile load tests for Arkansas River Lock and Dam No. b. A procedure was introduced whereby the residual loads were presented after completion of the tension test. The procedure applied to one of the test piles at the Old River Low-Sill Structure is illustrated in fig. 5- Additional calculations of residualload for test piles at this project are presented by Hunter anditDavisson. in some instances, residual loads of 25 to bo tons were computed for conventionally driven piles.

1 9 . Direct measurements of the residual loads were made in con­nection with the test piles for the Columbia Lock and Dam project. A typical result is shown in fig. 6, which further demonstrates that re­sidual loads induced by driving cannot be ignored if the true distribu­tion of load in the pile is to be determined.

Calculation of Failure Loads

20. On important CE pile projects, the load tests are usually carried out until sufficiently large displacements of the pile butt oc­cur, so that the -ultimate resistance is realized. The ultimate resis­tance is defined if the pile plunges into the soil. On the basis of

✓15instrumented pile load tests, Vesic concluded that the ultimate resis­tance of a driven pile should be taken as the load corresponding to a butt displacement of 10 percent of the pile diameter, or to a pile tip settlement of 8 percent of the pile tip diameter, whichever is smaller.

12

EM

BE

DM

EN

T,

FTLOAD IN PILE, TONS

COMPRESSION200

MEASURED LOAD DISTRIBUTION IN COM­PRESSION TESTRESIDUAL LOAD AFTER COMPRESSION TEST

MEASURED LOAD DISTR IBU TIO N IN T E N ­SION TESTRESIDUAL LOAD AFTER TENSION TEST

ADJUSTED LOAD DISTRIBUTION IN COM­PRESSION TEST

Fig. 5* Determination of residual load in pile

IO

SOFTTOM ED IU MCLAYS

; F INE• TO• M ED IU M i SA N D S

(LujQ

30

40

50

60

ST IFFCLAYS

LOAD IN P IL E . TONS

Fig. 6. Measured residual loads at Columbia Lock

The Building Research Institute, Tokyo, also has concluded on the basis of instrumented load tests that the ultimate resistance corresponds to a settlement of 10 percent of the pile diameter for a driven pile, and to a settlement of 1 to 1.5 pile diameters for buried piles.Leonards suggests that, unless the pile fails by plunging, the test should not be stopped unless the pile deflection exceeds 20 percent of the tip diameter (or the structural strength of the pile is being ap­proached) to insure that the full point resistance is mobilized.

21. The failure loads for CE pile tests have generally been based on a variety of empirical procedures which insure a safe load from the standpoint of settlement rather than bearing capacity. A summary of the procedures that have been used in some recent CE large pile test pro­grams is presented in table 1. In order to provide data for analyses which could be tied into conventional laboratory tests, the actual fail­ure loads based on tip movements of 10 percent of the pile diameter were

Ik

FAIL

URE

LOAD

BAS

ED O

N A

TIP

MOV

EMEN

T OF

10

PERC

ENT

OF T

HE P

ILE D

IAME

TER

recomputed as shown in fig. 7. The actual failure loads are approxi­

mately 20 percent higher than the average failure load determined by

the procedures shown in table 1.

0 50 100 150 2 00 2 50 300 350REPORTED FAILURE LOAD, TONS

Fig. 7- Comparison of actual and reported failure loads

15

PART III: ANALYTICAL DETERMINATION OF PILE BEARING CAPACITYIN COHESIONLESS SOILS

Basic Concepts

22. The ultimate bearing capacity of a single, axially loaded pile in cohesionless soil is a fundamental problem in foundation engi­neering. Several methods may be used to predict the bearing capacity of a pile foundation. This portion of the report discusses the application of rational mathematical models to the problem of static bearing ca­pacity of deep foundations in sand.

23. Conventional pile bearing capacity theories generally sepa­rate the ultimate pile capacity into two components (fig. 8): the

tip resistance Q and the skin friction (frictional resistance)Q . Each component is usuallysassumed independent of the other component. The tip bearing ca­pacity is described by the prod­uct of the tip area andthe unit tip resistance q The unit tip resistance is eval­uated using an appropriate limit equilibrium method. The shaft capacity is determined by the product of the shaft area Ag and the average unit skin fric-

Fig. 8. Pile bearing capacity problem

tion f A distinction ismade between the average unit

skin friction f and the unit skin friction at a given depth f s sThe average unit skin friction is also prescribed by a particular theoretical and/or empirical method.

2k. Recently3 numerical methods have been applied to pile founda­tion problems in which a mathematical model of the pile-soil system is

l6

devised to smulate the load-settlement behavior of the pile. These techniques require both deformation and failure parameters to model the behavior of the pile-soil interaction.

25. There are many factors which profoundly affect the ultimate bearing capacity of piles in cohesionless soils. Some of these factors include the geometry and method of installation of the pile, the non- homogeneous composition of natural soil deposits, the complicated de­formation behavior of cohesionless soils during shear, and the complex interaction of the pile-soil system in the vicinity of the foundation base. Incorporation of these and other factors into the rational anal­ysis of a deep foundation requires considerable engineering judgment.

26. The subsequent paragraphs discuss some basic theoretical and semiempirical solutions to the bearing capacity problem applied to deep foundations. Emphasis is placed on the assumptions made in each case and on some of the difficulties encountered by investigators who have applied a particular approach.

Tip Capacity

27* Classical bearing capacity theories generally apply limit equilibrium techniques of plasticity theory to solve a given problem. The most common assumptions made for the material behavior in these approaches include:

a. Mohr-Coulomb failure is a valid criterion for soil.b. Strength at any point is independent of strain.c. Elastic deformations are negligible with respect to

plastic deformations.d. Volume change due to shear or compression loading is

negligible.

28. These assumptions describe a rigid-plastic material which conforms to the Mohr-Coulomb strength criterion.

29. Limit analysis of plasticity theory imposes additional as­sumptions. Essentially, such a solution requires the establishment of a representative kinematic failure mechanism with associated boundary

17

and discontinuity conditions. Hie solution to a given problem is theapplied stress state which satisfies the equations of equilibrium forl8the kinematic configuration based on a prescribed failure condition.A detailed description of extremum principles may be found in appropriate texts on plasticity theory. It should be noted that these methods pro­vide an approximation to the exact mathematical solution of an idealized problem of continuum mechanics. The remainder of this section concerns several analytical solutions for base bearing capacity which are most prominent in the literature.Prandtl-Reissner-Terzaghi pattern

30. The original theoretical solution of the bearing capacity19 20 21problem is attributed to the work of Prandtl 9 and Reissner. The

failure pattern which they assume is given in fig. 9(a). Most of the17 22later theories are extensions or modifications of this work. Caquot

23and Buisman first applied this failure pattern to deep foundation problems

31. The solution found by Prandtl and Reissner concerns theplastic flow in a semi-infinite , weightless solid due to a uniformly distributed infinite strip load on the surface. The soil is assumed to be a rigid-plastic material having Mohr-Coulomb strength components c and ft . A failure configuration is prescribed which combines three types of plastic equilibrium shear zones: an active Rankine zone be­neath the base5 two passive Rankine zones, and two zones of radial shear (Prandtl zones) which permit the transition between the active and pas­sive zones. A uniform surcharge q is assumed to act on the surface of

2kthe solid adjacent to the strip load.2k32. Terzaghi describes the failure mechanism of Prandtl and

Reissner as general shear failure. He derived the equation for the strip load-bearing capacity problem by superposition of three different solutions to related problems. These solutions are added to determine the bearing capacity per unit length of a continuous footing Q, as

Qt = 2B(cNc + qN + 7^ ) = (1)

18

REISSNER (1924) CAQUOT (1934) BUISMAN (1935) TERZAGHI (1943)

DE BEER (1945) J/Ck Y (1948) MEYERHOF (1951)

Cc)BEREZANTSEV &

JAROSHENKO (1962)VESIC (1963)

Fig. 9- Pile foundation failure patterns (after reference 17)

where2B = width of the footing

Nc,N^,N^ = hearing capacity factors for shallow continuous footings33. Terzaghi introduced the primed factors N' , N* , and N' ,

0 /

which he recommends for the condition of local shear failure. Local shear failure occurs as a result of loose or highly compressible soil beneath the foundation base. It is characterized by sinkage of the foundation without full mobilization of the failure pattern to the base level. Terzaghi recommends modification of the strength parameters to reflect this effect. He employed c' = 2/3c and tan 0' = 2/3 tan 0 to compensate for the compressible soil behavior, yielding reduced val­ues for the bearing capacity factors. His criterion for selection of failure mode is based upon the stress-strain behavior of the soil ob­served in the laboratory. A material which continuously mobilizes shear strength with increasing strain with no definite peak strength is con­sidered likely to fail in local shear.

3^. Terzaghi applied shape factors sc > s , 8X1(1 s/ 0 respective terms of the basic equation to correct for finite shape of the foundation. He adopted values for these factors based upon empir­ical results of his own and other investigations. For shallow founda­tions, Terzaghi prescribes values for sc 5 Sq 5 8X1(1 Sy aS ‘''*3’ and 0.6 for circular base foundations and 1.3? 1.0, and 0.8 for square

2kbase foundations.35» For the problem of a deep foundation, Terzaghi describes a

failure surface for the term, in which he considers the complexinteraction in the vicinity of a cylindrical foundation shaft. He de­rived the equation for equivalent overburden stress due to the hollow cylinder of soil, with inner radius R and outer radius nR , surround­ing the foundation. The term q^ replacing q in equation 1 includes the forces resisting upward movement of the surface of the general shearpattern due to average unit skin friction f at radius R , and mobi-slized shear stress t at radius nR , which transmit resultants through the soil cylinder to the foundation base in addition to the weight of soil within the cylinder.

20

36. The equation for determining the point bearing capacity of a cylindrical deep foundation according to Terzaghi is

Qt = ttR ( 1 .3 cNc + cqjW + 0 .6 /N ) (2)

where R is the radius of the base of the cylindrical foundation and the equivalent overburden pressure

D 7 + 2f + nr s(n - 1) R

(3)

whereD = depth of the foundation base7 = average effective unit weight of the overburden

He recommends fully mobilized average skin friction be used in equa­tion 3? hut he indicates that the value of t greatly depends on the compressibility of the soil beneath the foundation base. The numbern £ 1 is defined as that value for which the computed point resistance

2 kQ, in equation 2 is a minimum.37. Terzaghi emphasizes the uncertainty involved in using these

parameters to evaluate tip resistance. The values of bearing capacityfactors are derived for the plane deformation problem of a shallow stripfoundation on a rigid-plastic solid. The effect of volume compress-

211.ibility on the point resistance has been consistently disregarded. A common assumption made in using this method of analysis is that the overburden pressure is conservatively prescribed as the pressure exerted by the unit weight of soil times the depth of the foundation base q^ = D7 . Other investigators have determined experimentally that this assumption may be unconservative for foundations placed at depths greater than 15 to 20 base diameters D/2B > 15 to 20 .^5*25 Esther- more ? the influence of the method used to install the deep foundation is not considered analytically in determining the point bearing

2kcapacity.38. A semi-empirical approach to the problem was taken by Brinch

26Hansen. He used the primary bearing capacity factors and of

21

Prandtl and proposed an empirical equation for . The superposi­tion solutions are generalized by the introduction of appropriate dimen- sionless shape, depth, and inclination factors, which all depend on the relative depth D/2B and the friction angle f> . For deep foundations, he combined the shape and depth factors to a term SQdc = s^d^ and ig­nored the effect of the term. This combined term is multiplied bythe unit base resistance of a continuous foundation under the same over­burden. Fig. 10 contains equivalent bearing capacity factors for deepcircular foundations N* which summarize the theories of several

15 qinvestigators.DeBeer-Jaky-Meyerhof -pattern

39. One modification of the Prandtl-Reissner-Terzaghi failure pattern involves a solution for which the failure pattern extends abovethe foundation base and includes the shear resistance of the overburden

2Vsoil. A study of DeBeer evaluated the penetration resistance of an incompressible material using a failure pattern for which the boundary reverts back to the shaft, as shown in fig. 9(b). A paper by Jakyemploys a similar pattern to solve the same problem. The most extensive

29work using this type of failure mechanism has been done by Meyerhof.ho. Meyerhof solved the problem of a continuous strip foundation

based on the same material behavior assumptions of Terzaghi's theory, with similar superposition of related solutions combined to form the general solution. The basic difference involves the assumption of a kinematic failure mechanism which is fully developed above the founda­tion base. To determine values of N and N he assumes a weightlessc qmaterial within the failure zone which is bounded by an "equivalent free surface" extending from the base at an angle P with the horizontal, which intersects the failure surface at ground level. The values of are developed from passive pressure calculations using the log spiralmethod and the same failure pattern. The equation for the total base

29resistance per unit length is written

®t = Eqo = B(cNo+poH(i + f Nr) 00

19

22

Fig. 10. Bearing capacity for deep circular foundations(after reference 17)

23

BEAR

ING

CAPA

CITY

FA

CTO

R

where B is the base width of the strip foundation.Ul. The unit normal stress on the free surface pQ depends on

the unit skin friction at the vertical surface of the foundation, the weight of the wedge of soil above the equivalent free surface, and the shear resistance which is mobilized tangential to the free surface. The values of the general bearing capacity factors depend on the depth, ge­ometry, and roughness of the foundation as well as the friction angle of the soil. These terms are uniquely defined for interrelated depth pa­rameters. The details of the computations are given in Meyerhof's

29paper. A clear discussion of Meyerhof's method is contained in the 18text by Scott, with same further assumptions which simplify the computations.

U2. The curves representing the three bearing capacity factorsindicate that the mobilized shear stress on the equivalent free surfaceis of little consequence. For the problem of deep foundations, thefailure pattern is assumed to revert back to the shaft (fig. 9(b)). Theunit normal stress pQ acting on the equivalent free surface is givenas the average normal stress acting on the pile shaft within the failure

29zone. The value is computed as in an earth pressure calculation,

Po =

KCrDs (5)

0where Kg is the lateral earth pressure coefficient. For buried founda­tions in cohesionless soils, Meyerhof recommends K values of 0.5 forloose sands to 1.0 for dense sands. For driven foundations K should

29be determined by in situ testing.i+3• In his original paper, Meyerhof introduced a complex, semi-

empirical shape factor X , which is multiplied by the bearing capacity of an infinite strip of the same width to determine the point bearing capacity. This shape factor depends upon the base geometry L/B (L =foundation length), the relative depth D/B , the friction angle ft ,

29and the method of installation.l+U. For relatively shallow foundations (D/B < 5) in dense sands,

Meyerhof's solution predicts greater point bearing capacity than

Terzaghi's method. For such foundations, Meyerhof's method more closely approximates the bearing capacities observed in the field. For deep foundations (D/B >5) and foundations in very compressible sands, Meyerhof's solution gives far greater tip capacity than observed in field load tests. Meyerhof attributed this behavior to volumetric com­pressibility of the soil which causes local shear failure beneath the foundation. He recommends, therefore, an empirical compressibility factor k to modify the friction angle in the same fashion as Terzaghi; i.e.,

tan = k tan ft (6)

where k takes the value of O .85 for deep, buried foundations. Herecommends a value of 0.95 for driven foundations since the compress-

29ibility is somewhat offset by compaction of the soil due to driving.1+5. Vesic presents factors after Meyerhof^ for bored and driven

piles that apparently incorporate the semi-empirical factors, whichspecify a deep foundation bearing capacity factor, N* , as a function

17 ^of friction angle only. For deep foundations, the term containing is small compared with the others and is usually neglected. For cohe­sionless soils, c = 0 such that the point bearing capacity may be written as^

where q. is the effective vertical stress at failure at the level of the foundation base and N* is the bearing capacity factor for a deep circular foundation, a function of jZi only. As described previously, curves summarizing appropriate N* values after Meyerhof are shown in fig. 10.

1+6. In a subsequent paper, Meyerhof extended his theory to in­corporate the effects of compaction on point bearing capacity of driven piles by considering the compaction due to installation resulting in prestressing and increased internal friction in the vicinity of the

25

foundation base. Other articles by Meyerhof that consider in detail various aspects of bearing capacity theory are references 32-3 5? which will not be discussed herein.Berezantsev-Jaroshenko-Vesic pattern

^7. Berezantsev and Jaroshenko^ were the first investigators toemploy a modified failure pattern (fig. 9(c)) to study local shear fail-

17ure beneath foundations. Based upon extensive experimental observa­tions, Vesic derived values for the Nq factor using a similar pattern15 37to represent local shear failure. Observations and conclusions ofthe studies by Vesic are discussed in this section.

i+8. Vesic examined the load-deformation behavior of foundationsin sand using laboratory and field tests under absolutely controlled

15conditions, similar to the procedure employed in separate work re-n r o Q

ported by Kerisel. J > ~ > The laboratory model tests included considera­tion of the relative depth and shape of the foundation, the relative density of the sand, and the method of foundation installation. Great care was taken to insure homogeneity of the air-dried, cohesionless ma­terial. Field tests were conducted on large-scale piles installed at a

15site composed of fairly homogeneous, moist sand.^9. Types of failure. In order to establish failure criteria for

the model tests, Vesic described three characteristic modes by which a surface foundation might fail, depending upon the relative density of the sand. For dense sands (D > 70 percent), Terzaghi's pattern of general shear failure represents the mechanism exactly. For medium dense sands (35 percent < < 70 percent), the abrupt failure was notobserved; nor did the load necessarily reach a peak, but a slight sur­face bulge was apparent. This phenomenon has been described as local shear failure. For relatively loose sand (D < 35 percent), the foundation penetrated without surface bulge, and base resistance in­creased continuously with penetration. Vesic defines this behavior as punching shear failure. He defines failure load in the latter two cases as the load at which the settlement rate becomes a maximum.^

50. For foundations placed at various relative depths and

31

2 6

relative densities, Vesic examined the failure pattern for two basic shapes: long rectangular foundations and circular foundations. Heplotted the boundaries between the three different failure modes ob­served in the model tests on buried foundations as a function of rela-

15 57tive depth and relative density of the sand. 5 In the case of deep circular foundations (D_p/B > 5), the failure mechanism is always punch­ing shear failure, regardless of relative density.

51. Failure pattern. Recognizing that punching shear failure oc­curs beneath deep foundations, Vesic proposed a failure mechanism which closely resembles the localized deformation pattern observed (fig. 9(c)).

52. Vesic assumed that the overburden q is great enough to ne­glect the soil weight within the failure pattern. He made the same ma­terial behavior assumptions as Terzaghi (rigid-plastic, Mbhr-Coulcanb solid). Assuming that the effective overburden pressure at failure,q , is the minor principal stress, Vesic used equations after Reissner to compute the unit tip capacity qQ as

. 2 /rr , 20 tan 0 /o\% = qf tan ^ e r (8)

where 0 is an angle defined by the geometry of the failure zone.53. Based upon experimental observations Vesic assumed a value

of 0 = 1.9ft such that

1, = qf tan2 (j + |) e3'80 tan ? (8a)

ttj = tan2 (j + §) e3 '8 tan $ (8b)

where is the bearing capacity factor for continuous deep founda­tions in cohesionless soils.

5b. Using a relationship for and a related shape factorequation, Vesic wrote the general equation for unit point bearing capacity as'*’

*o = cN s c c 4- W q (9)

27

where

q_p = effective vertical stress at failure at the level of the foundation base

It should be noted that this set of equations has only two independent variables, N and s . For the case of circular deep foundations

q qin cohesionless soils, he combines these factors to form the bearingcapacity factor N* , as defined in equation 7. Appropriate values

*1/15for N* after Vesic are included in fig. 10.55* Experimental observations. Field and model load tests by

Vesic led to several interesting conclusions. Using the theoretical formulas for deep foundations, equation 9? if is commonly assumed that q_p is equal to the overburden stress at the foundation base level prior to foundation installation. This would indicate that in a homogeneous sand (constant ft) the N* value would be constant, such that qQ would increase proportionally with depth. Vesic found this to be true only for foundations at relatively shallow depths (d/b k ) . In fact, for greater relative depths (D/B > 1 5 to 20), the value of qQ remained es­sentially constant, independent of depth. In addition, the value of theunit skin friction f also reached a limiting value at the same rela-stive depth. The values of unit tip and skin capacities were found to be functions of relative density (or friction angle) only.

56. Vesic discussed similar experimental results which led Kerisel to conclude that is a complex function of ft , D/B , andB that decreases with depth. Vesic, however, suggests that at greater depths q^ is no longer equivalent or proportional to the initial over­burden stress 7D . He recalls that the true meaning of q^ in the various theories is the effective overburden pressure at failure at the level of the foundation base

57« The rational explanation for the asymptotic final values of q and f considers an arching phenomenon of the soil surrounding the0 s 2kpile, similar to the yielding pattern described for silo design. The

28

mass of soil beneath the base is compressed downward, and the sand around the pile tends to follow the downward movement causing stretching in the soil mass (extension) and, consequently, vertical stress relief^ (see fig. 1 1 ).

LEGENDq - I N I T I A L OVERBURDEN

S T R E S S * r Z

q z * V E R T IC A L S T R E S S AT DEPTH Z

q 7 - AVERAGE V E R T IC A LSTRESS = q j . l / 2/ J q 2d z

q -L IM IT IN G V E R T IC A L ' S T R E S S

Fig. 11. Vertical stress distribution around a deep foundationin sand (from reference 1 7 )

58. Laboratory observations indicate a loosening of dense sand15 87adjacent to a model pile tip. ’ In an X-ray graph study of displace­

ments in sand surrounding model piles, this stretching phenomenon wasclearly observed 39 In a finite element analysis of the model testsconducted by Vesic, Ellison presents numerical results that predictthe onset of tensile stresses near the pile tip. A subsequent finiteelement analysis of a pile in cohesionless soil likewise exhibited soilUlextension and vertical stress relief in the vicinity of the pile tip.

59« For the case of driven piles in cohesionless soils, Vesic modified the friction angle to accommodate the increased relative den­sity due to pile installation. He computed N* from measured qQ and f values and plotted these versus relative density. He found that the relationship given by Berezantsev et al. (fig. 10) would fit the data quite well for both driven and buried foundations. He used the mean density of that measured before and after driving to determine the

29

relative density. He used the correlation between relative density andfriction angle to adopt the theoretical values as a function of relative

15density. A similar correction of friction angle for driven piles was31adopted by Meyerhof in analyses of driven piles.

60. A final observation made by Vesic in the original study was that the ratio of point and skin resistance q ^ f is apparently a parameter directly related only to the relative density of sand and the method of placement of the pile.^" In a later study, Vesic proposed em­pirical correlations of the limiting point and shaft resistances with

ii-3relative density as

qo £ (4)(10) r (10a)

1.5D^f £ (0.08)(10) r (11a)s

for driven piles, in tons per square foot; as2.ifD3

qQ £ (1.5)(10) r (10b)k1.5D

f £ (0.025)(10) r s (Hb)

for bored piles and piers, in tons per square foot, where is givenas a decimal; and as

q 1.3 tan f6 (12)(9) (10)

s

for both driven and bored piles * where jZi is determined from a drained triaxial test at a 10-psi confining pressure.

6l. These limiting values are reached at a critical depth varying between 10 pile diameters in very loose sand and 20 pile diameters in very dense sand. Vesic also cautions that these equations represent piles in dry or submerged sands and that the effects of negative pore pressure and capillary cohesion in moist sands need to be considered

30

separately. Another condition which may render equations 10 and 11 in­correct may he the case in which a sand layer is first encountered and

lj.3only partially penetrated by the pile.

Skin Friction

62. The Mohr-Coulomb strength criterion is generally assumed to represent the skin friction resistance to the movement of the pile rel­ative to the soil. It is also generally assumed that the failure is developed at the pile-soil interface, such that the maximum unit shear resistance on a vertical pile shaft at depth z may be written as

f (z) = c + p (z ) tan 6IS U S

(13a)

where c is the adhesion component of the shaft resistance, p (z) isQj S

the normal stress on the shaft at depth z from the surface, and 6 is the friction angle between soil and shaft.

63. For cohesionless soils, the adhesion term is zero. The shaft resistance, therefore, is purely frictional. The angle of interface friction depends on the roughness of the pile shaft, and may depend on the initial relative density of the sand. Appropriate values for 6 may be selected from laboratory tests or approximated from the results of other investigations.

Gb* The remaining term in equation 13a, p , represents the hori-szontal effective stress at the pile-soil interface at depth z . It iscommonly linked to the effective vertical stress q by a proportion-zality constant K . It is usually written in the form s

Ps K q s^z (lit)

where K is the coefficient of lateral earth pressure on the shaft at failure.

65. hi this manner, the term K incorporates the effects of ini­tial in situ stresses and stresses introduced by pile installation and

31

by loading to failure. Various relationships proposed to evaluate thegnormal pressure on the shafts of driven piles are given in table 2.It is obvious that there is wide variation in both measured and theo­retical values for the proportionality constant.

66. It is commonly assumed that the effective vertical stress at failure is given by the initial effective overburden pressure which exists in situ prior to pile installation. This assumption coupled with equation 13a yields the relationship for cohesionless soil as

which, when integrated over the shaft surface and embedded pile length D , determines the skin friction capacity Q^g as

where J is the effective unit weight of the soil.67- Experimental measurements of pile load distribution for piles

installed in homogeneous sands indicate that, for deep foundations, theunit skin friction f is proportional to depth only at moderatelysshallow relative depths (D/B < U). Thereafter, the skin friction in­creases at a decreasing rate, reaching a limiting value at some critical relative depth.10,15*16,25,37 i^e critical relative depth depends upon the initial relative density of the sand and the method of pile instal­lation, varying from 10 to 20 pile diameters#-^,15*37 instrumented pile load tests in this report exhibit similar skin friction distributions, which reach limiting values at relative depths in this range.

68. Based on large-scale model tests in sand, Vesic gives em­pirical relationships for the limiting values of unit skin friction that depend only on the method of installation and the soil relative

hodensity. Equations 11a and lib are given in tons per square foot, as °

f (z) = K 7z tan 6 s s (13b)

<yQ . = Kg tan 6 (effective perimeter) (15)

f £ (0.08)(10)s (lla bis)

32

fo r d riv en p i l e s , and as

-1 - • J U

f £ (0 .0 2 5 )(10 ) r ( l i b b i s )s

fo r b u rie d p i le s and p i e r s .

69* V e s ic e x p la in s t h is observed b eh a vio r in terms o f an a rch in g

phenomenon which i s most pronounced w ith in a few p i l e d iam eters o f the

p i l e p o in t ( f i g . 11 ) . " ^ ^ The a rch in g b eh a vio r causes e x ten sio n in

t h is r e g io n , which a llo w s both v e r t i c a l and h o r iz o n ta l s t r e s s r e l i e f ,

and co n seq u en tly reduces sh earin g s tre n g th a t the p i l e - s o i l in t e r f a c e .

F in ite elem ent a n a ly se s o f s i n g le - p i le fo u n d atio n s a ls o p r e d ic t th is

a rch in g phenomenon w ith th e r e la t e d asym p totic v a lu e s o f sk in f r i c t i o n l^Q i+1

and s t r e s s r e l i e f . * An e m p ir ic a l method o f computing the e f f e c t i v e

v e r t i c a l s t r e s s a d ja c e n t to th e p i l e s h a f t , c o n sid e rin g th e a rch in g e f -k2f e e t , i s g iv e n by B ere za n tse v e t a l .

7 0 . V alues o f K computed from p i l e lo a d t e s t d ata u s u a lly g iv e

an average v a lu e under the assum ption th a t th e v e r t i c a l s t r e s s i s e q u iv ­

a le n t to i n i t i a l e f f e c t i v e overburden s t r e s s . This approach autom ati­

c a l l y assumes th a t th e u n it s h a ft r e s is ta n c e i s p ro p o rtio n a l to d epth ,

p r e s c r ib in g an e q u iv a le n t t r ia n g u la r sk in f r i c t i o n d is t r ib u t io n ( f i g .

1 2 ) . The shape o f the sk in f r i c t i o n d is t r ib u t io n curve can a ls o v a ry

w ith r e la t i v e depth o f embedment. For lo n g e r p i l e s o f th e same diam­

e t e r , th e sk in f r i c t i o n in c re a se s l i n e a r l y w ith depth to a g r e a te r r e l ­

a t iv e d epth . The slop e o f th e sk in f r i c t i o n d is t r ib u t io n c u rv e , how-g

e v e r , i s l e s s fo r the lo n g e r p i l e s . I t appears th a t an average Ksv a lu e may a ls o depend upon th e depth o f embedment.

7 1 . The sk in f r i c t i o n d is t r ib u t io n may be e v a lu a te d u sin g average

v a lu e s o f th e sk in f r i c t i o n to a g iv en d ep th , based on u n it sk in f r i c ­

t io n determ ined from f i e l d m easurem ents. The average u n it sk in f r i c t i o n

f fo r th e p i l e i s computed assuming an e q u iv a le n t uniform d is t r ib u - st io n . F ig u re s l i b and H e d e sc r ib e th e v e r t i c a l e f f e c t i v e s t r e s s d i s ­

t r ib u t io n around a deep fo u n d atio n in sand. F igu re l i b re p re s e n ts th e

average v a lu e q o f a h y p o th e t ic a l curve re p re s e n tin g the a c tu a l zv e r t i c a l s t r e s s d is t r ib u t io n qz v e rsu s depth ( f i g . 1 1 c ) .

33

EM

BE

DM

EN

T R

AT

IO,

z/D

EFFECTIVE PRESSURE UNIT SKIN FRICTION LOAD IN PILE

Fig. 12. Distribution of load in pile based on static pile formula

o<>

72. The relative displacement necessary to mobilize full skin friction in sands is on the order of 0.3 in., independent of initial relative density, pile diameter, and method of installât ion. may be important to consider the direction of the relative shaft dis­placement subsequent to installation. For field conditions in which the soil may move downward relative to the pile, the skin friction could even be fully mobilized in the downward direction. This phenomenon is described as negative skin friction. It may cause significant settle­ments of pile foundations.

73* The tip displacement necessary to mobilize full tip resis­tance depends upon the base diameter, the method of installation, and the initial relative density of the sand. For driven piles in sand, the tip displacement at ultimate resistance is about 8 percent of the base diameter. For bored piles or piers, a tip displacement in the range of 20 percent of the base width is needed to mobilize full resis­tance. Thus, the full skin friction is mobilized before the point sus­tains any significant load. For a given method of placement, the re­quired point displacement for -ultimate load increases with increasing, .. 15*25density. 5

7^. For dense sands which exhibit a marked peak and subsequent strain softening in laboratory tests, the total skin friction may be reduced by the time the full point resistance is mobilized. Under these conditions, the ultimate bearing capacity may be significantly less than the sum of the peak values of shaft and point resistance determined using peak strength parameters.

35

PART IV: PILE TESTS IN COHESIGNLESS SOILS

CE Design Procedure

75. Procedures used by the CE in estimating pile capacities in cohesionless soils are based on the conventional static formula using TerzaghiTs formula (equation l) for the tip capacity. The unit skin friction is assumed to vary directly with depth, as shown in fig. 12. The angle of wall friction is usually based on laboratory shear tests. The coefficient of lateral earth pressure is usually determined empiri­cally from previous pile load tests. On the basis of the above assump­tions, the load in pile, as shown in fig. 12, decreases parabolically with depth with the maximum frictional load developed at the pile tip.

76. The static formulas are relatively simple and have been used extensively by the CE in interpreting pile load tests and extrapolating the results to field conditions. The formulas have apparently provided reasonable predictions resulting in satisfactory and economic pile foun dations. However, some of the basic assumptions in the formulas have not been consistent with field measurements in CE load tests, and it ap pears that a critical appraisal of these observations and the simple static formula is necessary if present criteria are to be improved.

77. Data on the magnitude and distribution of skin friction for piles in cohesionless soils can only be obtained from carefully instru­mented pile load tests. Such tests have been performed by the CE for

12the Old River Low-Sill Structure and the Arkansas River Lock and Dam No. 1+.13

Data from CE Pile Tests

78. A tabulation of compression pile tests by the CE in cohesion less soils is shown in table 3* Only those tests are shown for which sufficient data on soil and groundwater conditions are available for proper interpretation. A similar tabulation for tension tests is shown in table In addition to the reported failure loads, the ultimate

36

failure loads based on a settlement or rise of 10 percent of the tip diameter are also shown. Because the groundwater table has an important influence on the failure loads, as previously discussed, failure loads corrected to a groundwater level at ground surface are also shown. The corrections were made in accordance with the equations shown in figs. 3 and k . No corrections were made for possible errors in measured applied loads due to the uncertainty of the correction factors.Compression versus tension capacity

79« A relatively large number of piles have been tested by the CE in both compression and tension. These tests provide a useful indi­cation of the relative capacities in compression and tension. Based on data in tables 3 and. the relation between the capacities for piles in silts and sandy soils is shown in fig. 13* Also shown is a similar re­lation, which will be discussed subsequently, for piles in clay. For piles in cohesionless soils, the tension capacity varies from about 30

to 50 percent of the compression capacity.Average skin friction

80. All of the load in tension tests is carried by skin friction. The average unit skin friction f was computed for the tension tests shown in table k assuming a linear distribution of skin friction, as indicated in fig. 12. The calculations were based on the ultimate ca­pacity corrected to a zero groundwater depth. The average unit skin friction varied from 0.20 to 0.6l ton/sq ft. The average unit skin friction could not be reliably computed from the compression tests be­cause of uncertainties regarding the distribution of tip load and skin friction. As will be discussed later, average skin friction values can be computed only for those compression piles which are instrumented to determine the load distribution with depth.

81. A further breakdown of values of f for various types ofspiles is shown in table 5* Hie ranges of f for displacement and non-sdisplacement piles, tested singly and after adjacent piles were driven,are shown. Driving adjacent piles (group effect) is shown to increasef for both the displacement and the nondisplacement piles. Jetting of sdisplacement piles produces a significant reduction in f .

37

FAIL

UR

E LO

AD I

N C

OM

PRES

SIO

N

TON

S

50 100 150 200 250FA ILU R E LOAD IN T EN S IO N Q* ,T O N S

Fig. 13. Compression versus tension capacity from CE pile tests

400

350

300

250

200

150

100

50

0 300

FAIL

URE

LO

AD I

N

CO

MPR

ESSI

ON

Q

f. ,

TON

S

82. Laboratory determinations of the angle of skin friction 5 have been made in connection with important CE pile test projects. The tests are usually made with a direct shear box in which the sliding friction between the sand at natural density and pile material is mea­sured. The results of these tests are summarized in fig. lU. These values were used, as shown in table b, in computing the coefficient of lateral earth pressure K* . A summary of K* values computed from tension tests is shown in table 5*

Distribution of Load in Pile

83. The basic data defining the distribution of load in pile forthe load tests at the Old River Low-gill Structure and the ArkansasRiver Lock and Dam No. b were replotted and corrected for residual loadsinduced by pile driving using the procedure described in reference 13 •The corrected curves of pile load distribution for test piles at theLow-Sill Structure are shown in figs. 15 and 16 for compression andtension tests, respectively. The curves represent conditions at or nearfailure. The corrected curves of pile load distribution for test pilesat Arkansas River Lock and Dam No. U are shown in figs. 17 and 18 forcompression and tension tests, respectively. The latter are essentially

lbsimilar to those computed by Hunter and Davisson with the exception that the slope of the load distribution curve at its intersection with ground surface was considered to be zero. This assumes that zero skin friction at ground surface for a pile in cohesionless soils is valid.The discrepancy between load indicated by the uppermost strain rod or strain gage readings and the hydraulic jack gage was assumed to be due to errors in the jack gage readings.

8H. The distribution of measured load in pile is subject to con­siderable interpretation as the points seldom define a smooth curve.It is not certain whether erratic points represent instrumentation er­rors or actual variations in skin friction due to changes in density of materials. The piles at the Low-Sill Structure were driven through stratified silty sands into denser clean sands. The stratification

39

AN

GLE

O

F S

KIN

F

RIC

TIO

N

6, D

EG

50 50

0 10 20 30 40 50

ANGLE OF INTERNAL FRICTION <)>, DEG

40

30

20

10

00 10 20 30 40 50

ANGLE OF INTERNAL FRICTION <J>, DEG

Fig. Ik Angles of skin friction

AN

GLE

O

F S

KIN

F

RIC

TIO

N ,

6,

DEG

EM

BED

ME

NT,

FE

ET

E

MB

EDM

EN

T,

FEE

T

LOAD TONS0 50 100 150 200 250 300 350

LOAD, TONS LOAD, TONS

NOTE! LOAD DISTRIBUTION CURVES BASED ON STRAIN ROD DATA

Fig. 15 Distribution of compression load for the Old River Low-Sill Structure

4l

EMBE

DM

ENT,

FEE

T

EMBE

DM

ENT,

FEE

T

LOAD,TONS

0

10

20

50

60

70

LOAD, TONS0 50 100 150 200 250

LOAD, TONS LOAD,TONS

n o t e : l o a d d i s t r ib u t io n c u r v e sBASED ON STRAIN ROD DATA,

Fig. 16 . Distribution of tension load for the Old River Low-Sill Structure

h2

EMB

EDM

ENT,

FE

ET

EMB

ED

ME

NT,

FE

ET

LOAD, TONS LOAD, TONS IOO 150 200 250

10

20

I -

5 30 2 û u (X)3Ui

JACK LOA1<D = 162 T<

1DNS

i ••

<

• j

1•

<r1^ J E T T E L ? TO 40 F T/52.7 FT— 1ST

16-

1

P IL EIN -OD

_____ 1

NO. 1 6PIPE

LEGENDO STRAIN ROD DATA • STRAIN GAGE DATA

Fig. 17 Distribution of Lock

compression load for Arkansas Riverand Dam No. k

3

EM

BE

DM

EN

T,

FE

ET

E

MB

ED

ME

NT

, F

EE

T

LOAD, TONS0 50 100 150 2 0 0 2 5 0

0 ----- -- 1--------------- 1---------- 1----------•

• 1JACK LO/

1------------------kD * 78.6 T<

1-----------------DNS

~~1*/ •

f

r IE T TED TO 40 F T

------ 5 2 .7 FT TE:s t P ILEI6 -IN .-O D

__________ 1N O . 16PIPE

LEGENDo STRAIN ROD DATA • STRAIN GAGE DATA

Fig. 18. Distribution of tension load for Arkansas RiverLock and Dam No. k

undoubtedly has some effect on the distribution of load in the piles. Bending of the piles during driving and load testing may also affect the distribution of load. In general, the load distribution curves were drawn through all observational points, although in some instances smooth curves (the dashed lines) are also shown to represent the more likely overall distribution.

85. Comparison of the corrected load distribution curves in figs. 15-18 with the assumed parabolic distribution given by the conventional static formula in fig. 12 indicates substantial differences. The com­pression tests indicate a curved distribution of load in pile inthe upper part of the pile with a tendency for less load to be carried by friction near the pile tip. On the other hand the distribution of tension load in the pile indicates a tendency for a linear reduction in load with depth below the curved upper portion.

Unit Skin Friction

86. Based ion the observed distribution of load in pile corrected for the residual loads induced by pile driving, curves of unit skin friction f versus depth were drawn. For test piles at the Old River Low-Sill Structure and Arkansas River Lock and Dam No. k, plots of unit skin friction versus depth are shown for both tension and compression tests in figs. 19-22. The construction of these curves is very sensi­tive to errors in either the strain gage or the strain rod readings.The assumptions employed in determining the residual loads also have a significant effect on the shape of the unit skin friction versus depth curves. Consequently, it is not certain whether the distributions of unit skin friction reflect these possible errors or actual variations in density and shear strength along the pile. Nevertheless, it appears that the skin friction tends to reach a maximum at about a depth of25 ft, after which it tends to remain constant or else decrease. Simi­lar phenomena were observed by Vesic.^ It is important to note that, for both compression and tension tests, the unit skin friction does not increase linearly with depth, as indicated by the static pile load

EMB

EDM

ENT,

FEE

T E

MB

ED

MEN

T FE

ET

1.2 1.0 1.2UNIT SKIN FRICTION, TONS/SQ FT

Fig. 19« Skin friction versus depth from compression tests at theOld River Low-Sill Structure

U6

EM

BE

DM

EN

T,

FE

ET

E

MB

ED

ME

NT

, FE

ET

U N IT SK IN F R IC T IO N , T O N S /S Q FT U N IT S K IN FR IC T IO N , T O N S /S Q FT 0 0 .2 0 .4 0 .6 0 .8 1.0

TA(

— A VG Fs - i T O N S / S i

0 . / 2 ? F T

TES

___ i>T P IL E 1

I4 B P 7 3

|MO. 3

U N IT S K IN F R IC T IO N ,T O N S /S Q FT U N IT SKIN FR IC T IO N , T O N S /S Q FT

Fig. 20. Skin friction versus depth from tension tests at the Old River Low-Sill Structure

EMBE

DM

ENT,

FE

ET

EMBED

MEN

T, F

EET

UNIT SKIN FRICTION, TONS/SQ FT UNIT SKIN FRICTION, TONS/SQ FT

UNIT SKIN FRICTION, TONS/SQ FT 0 0.2 0.4 0.6 0.8 1.0

UNIT SKIN FRICTION, TONS/SQ FT

Fig. 21. Skin friction versus depth from compression tests at Arkansas River Lock and Dam No. k

EM

BE

DM

EN

T, F

EE

T

EM

BE

DM

EN

T,

FE

ET

UNIT SKIN FRICTION, TONS/SQ FT

UNIT SKIN FRICTION, TONS/SQ FT 0 0.2 0 .4 0.6 0.8 1.0 1.2

!

L!AVG fs * 0 . 25 TO NS/5 Q F T

!

______ L_

I--------- 1------------1

TE1ST PILE I4BP7.

1____________ i

NO. 7 3

UNIT SKIN FRICTION, TONS/SQ FT

Fig. 22. Skin friction versus depth from tension tests at Arkansas River Lock and Dam No.

formula. Analysis of the pile load tests in cohesionless soils using the static pile formula can, under certain circumstances, lead to sig­nificant errors. The analysis of the pile load test at Jonesville Lock indicated that the assumption of a constant skin friction below a certain depth provided a more consistent interpretation of the test results.

87. The reduction in skin friction near the tip of the pile forthe compression tests (figs. 19 and 21) was observed in model tests by

89Robinsky and Morrison, who used radiographs to determine the limits of visible displacements. Vertical compaction and radial expansion took place immediately below the pile tip, resulting in a reduction of lat­eral pressure and/or a lessening of the relative vertical strain be­tween pile and soil.

88. Plots of unit skin friction versus depth were derived from pile load distribution versus depth data which were corrected for re­sidual load estimates. Smoothed curves were drawn to approximate the load distribution data, and skin friction values (slopes) were calcu­lated from these curves. As such, the data provide a useful guide in estimating values of skin friction and earth pressures which can be used for preliminary design. Values of maximum unit skin friction and the coefficient of lateral earth pressure corresponding to the maximum unit skin friction are tabulated for compression piles in table 6 and for tension piles in table 7. The maximum unit skin friction for piles tested in compression varied from 0.2k to 1.13 tons/sq ft and was de­veloped at pile depth to diameter ratios of 8 to 27. The maximum values of the coefficient of lateral earth pressure ranged from O.ij-O to 2.1+0. The maximum unit skin friction for piles tested in tension varied from 0.25 to 0.7*+ ton/sq ft and was developed at pile depth to diameter ra­tios of about 6 to 23. The maxim-urn values of the coefficient of lateral earth pressure varied from O .58 to 2.60. The relatively wide range for values of maximum unit skin friction and coefficients of lateral earth pressure reflects difficulties in interpolating field strain and defor­mation measurements along the pile. Nevertheless, it is possible that

50

the values shown in tables 6 and 7 can provide an empirical basis for improving currently employed static pile formulas.

Tip Capacity

89. 3h the Terzaghi expression for the tip capacity of piles in cohesionless soils, equation 2, the term involving N^ is relatively small and can be disregarded, and c is equal to zero. The tip ca­pacity can be defined as

S> = V D\ (16)

Based on the results of compression tests on instrumented piles shown in figs. 15 and 17* the load distribution curves permit an estimate of the tip load for an applied load near failure on the pile butt. If it is assumed that, at the larger applied M d s , the frictional resistance is completely mobilized and any changes in applied load are reflected only in the tip load, the tip load at failure, based on a tip movement of 10 percent of the pile diameter, can be readily computed. These calcu­lations are summarized in table 8. No corrections were made for pos­sible errors in the applied load. Except for test pile No. 5 at the Old River Low-Sill Structure, which was imbedded in silty sands, values of N^ for the test piles varied from 31 to 77? with corresponding angles of internal friction from 33 to ^3 deg. These values appear reasonable for sands in a medium dense to dense condition. However, the range of values for the Arkansas River test piles is somewhat greater than that

1kcomputed by Hunter and Davisson. Test pile No. 5 indicated an N value of 2k and an angle of friction of 31 deg for the silty sands.

51

PART V: PILE TESTS IN COHESIVE SOILS

Basic Concepts

90. The bearing capacity of a pile placed in cohesive soil is in­fluenced by the in situ stress conditions and stress history, the method of pile installation, the type of pile, and especially time. Pile driv­ing generates high positive pore pressures and total stresses in the vicinity of the pile. The resulting high gradient causes flow away from the pile such that the remolded clay reconsolidates most rapidly near the pile surface. Clearly, the effective shear strength varies with time and depends directly on the stress history and stress-deformation properties of the remolded soil. The bearing capacity of a pile in clay should increase with elapsed time after driving such that a conservative estimate of capacity should be obtained for a test pile loaded shortly after installation. During a pile loading test, the rate of load appli­cation is generally much faster than the rate of pore pressure dissipa­tion. Therefore, the pile load test will probably generate excess pore pressures which may be significantly affected by the loading rate.

Data from CE Pile Tests

91. Review of pile load tests conducted by CE offices indicated arelatively small number of test piles loaded to failure in cohesivesoils. Comprehensive pile load tests in clay were performed in eonnec-

b5tion with the Mbrganza Floodway Control Structure, the Wolf River1+6 i+7Floodwall, and Columbia Lock and Dam. These tests provided a basis

for evaluating available analytic methods used for estimating the load­carrying capacity of piles in cohesive soils.

Soft or Firm Clays

92. The bearing capacity of a friction pile driven into soft or firm clay (undrained shear strength less than 0.5 ton/sq ft) is

52

determined by assuming that the adhesion component of skin friction is equivalent to the undrained shear strength. For most piles in such soils, the tip resistance is relatively small and can be neglected.Thus, the total capacity is usually written as the product of the aver­age soil adhesion c and the imbedded shaft area A , ora s

= cg/ s = ca ' D • (effective perimeter) (17)

93• The validity of this analytical approach has been substan-i+8tiated by Peck, who compared load test data with the results of field

and laboratory studies. As shown in fig. 23, the computed adhesion from compression tests by the CE agrees closely with the undrained shear strength. The CE pile load tests in soft or firm clays are described below.

O O.l 0.2 0.3 0 .4 0.5AVERAGE U N D R A IN E D S H EA R S T R E N G T H ,T O N S /S Q F T

Fig. 23» Computed adhesion versus undrained shear strength

Wolf River Floodwall9k. Pile load tests in compression and tension were conducted

prior to construction of the Section IB floodwall along Wolf River neark6Memphis, Tennessee. Precast concrete and steel pipe piles were driven

to various depths and subsequently tested to failure. The average unit

53

skin friction f was computed for each pile in both tension and com­pression. For the compression load tests, f varied from 0.22 tos0.32 ton/sq ft, with an average value of 0.28 ton/sq ft, which agreed reasonably well with the average undrained shear strength of 0.25 ton/ sq ft, obtained from laboratory tests. There was also no noticeable dif­ference between computed average skin friction values for the concrete piles and the steel pipe piles loaded in compression.

95. The average unit skin friction computed for the test piles loaded in tension was computed as 0.23 ton/sq ft for the concrete piles and 0.18 ton/sq ft for the steel pipe piles. These values are, respec­tively, 18 and 36 percent less than the average values obtained from the compression test results. These reductions were attributed to elapsed time effects. The compression tests were conducted about 20 days after driving, whereas the tension tests were performed only about 7 days after completion of the compression tests; therefore, the difference in excess pore pressure dissipation was the cause cited for the difference in values. The discrepancy between values computed for steel pipe and concrete pile skin friction in tension was not explained.Morganza Floodway Control Structure

96. A comprehensive pile testing program was undertaken at theh5Morganza Floodway Control Structure site near Morganza, Louisiana.

Seven test piles, including steel pipe, precast concrete, and monotube piles, were driven into soft clay. The average skin friction values computed for each pile tested in compression varied between 0.26 and 0.37 ton/sq ft, with an overall average of 0.32 ton/sq ft. This value is very close to the average undrained shear strength of 0.33 ton/sq ft obtained from laboratory tests. The tension load tests were conducted on piles which were driven through the clay into an underlying sand de­posit; therefore, the skin friction of the clay could not be determined directly for these tests.

Stiff or Hard Clays

97- Piles in a stiffer cohesive material (undrained shear

strength greater than 0.5 ton/sq ft) usually develop considerable tip resistance. Therefore, the capacity predicted for a pile load test in­cludes a tip resistance component as well as the shaft frictioncomponent Q, , and it is written ass

S i - caAs + ¿» A (1 8 )

where c is the undrained shear strength intercept, cohesion. Thevalue of the bearing capacity factor N is commonly assumed to be 9*

/ ^ 15Vesic has reported values of Nc between 5 and 2k. The value of ad­hesion is usually prescribed as a fraction of the undrained shear strength and is given by the equation

c = ac (19)cl

where a is the empirical adhesion coefficient.98. A series of pile load tests was conducted on instrumented H

piles and pipe piles prior to construction of Columbia Lock and Dam,1*7near Columbia, Louisiana. The piles penetrated through alluvial soils

into a stiff Tertiary clay. The instrumentation data permitted deter­mination of the unit skin friction. It was found that the maximum ad­hesion for all the piles averaged about 1.6 tons/sq ft in compression loading, which is nearly equal to the undrained shear strength of these materials. This computed value of adhesion far exceeds values normally expected for piles in stiff to hard clay soils. The minimum perimeter of the H piles was used in confuting the adhesion values. It was found that adhesion was not constant with depth in the stiff clay, especially for the H piles. The shaft resistance tended to increase with depth and to be fully mobilized near the tip of the pile. For pile load tests in tension, the computed maximum skin friction was about 80 percent of that determined from compression tests.

99. A summary (fig. 2k) of pile load test results on steel pipepiles in stiff, overconsolidated clays, including the Columbia Lock and

k9Dam tests, was prepared by Sullivan. These results not only show that

55

1.50

1.25

a2 1.00 UJulL

o 0 .75o20 (0Ui 0 .5 01 Q <

0 .2 5

0o 1000 2 0 0 0 3 0 0 0 4 0 0 0 5 0 0 0 6 0 0 0

C O H E S IV E S H EA R S T R E N G T H , L B / S Q F T

Fig. 2k. Adhesion of steel pipe piles in stiff overconsolidatedclays (from reference k9)

the adhesion is significantly less than the nndrained shear strength for stiffer materials, but also show that the value of the coefficient a depends as well on the sensitivity of the clay (sensitivity is theratio of undrained strength of the undisturbed clay to that of the same clay remolded, at the same void ratio). For sensitive clays (S _ >2) the remolded strength is considerably less than that of the undisturbed material, and the measured adhesion reflects the reduced strength. For insensitive clays (S^ < 2) the coefficient a describes the maintained strength in the remolded state such that the ratio of adhesion to co­hesion is appreciably greater than that for sensitive clays with the same undisturbed shear strength.

100. The effects of elapsed time between driving and load testing and of the rate at which load is applied may be very important when con­sidering the tip capacity in stiffer materials. Heavily overconsoli­dated materials often develop negative excess pore pressures during shear such that the effective pressures and apparent strength of the

O C L E V E V ARGON

LEGEND 1L A N D ,0 • L E M O O R E , C A L . NEj I L L . ▼ C O L U M B IA , L A .

A TO R O N □ ARROW

IO , UN 1./ L A K E , B .C .* u. r\. D L L . 1 u r ■ S T A N M O R E ,

IN .

EN G .

- IN S E N « C L A Y S

i lT IV ES t < 2

08 ■ ■ ▼ i

iki

1cf> V

M O D ER ATI C L A Y S S t

ELY S E N S IT IV E > 2

1

\

A F T E R SUL ____________________________1

?.L IV A N1__________________________

56

material may be much greater during a loading test than under slow load­ing conditions, which allow these pore pressures to dissipate. Under these circumstances the load tests may produce unconservative estimates of pile capacity.

Long-Term Capacity

101. Mach data has been published which indicates that the skin friction for piles in clay, based on tests conducted a significant pe­riod after driving, may be very different from the undrained shear strength of the clay. The shear strength of cohesive materials greatly depends on time in relation to the development and dissipation of excess pore pressures. Pile load tests in normally consolidated clays gener­ally develop positive excess pore water stresses during shear such that the effective strength is decreased and the resulting load test capacity is a conservative measure of long-term capacity. On the other hand, for piles in heavily overconsolidated materials, the development of excess negative pore pressures may produce a short-term strength that will de­crease as the pore pressures dissipate such that a pile load test may overestimate the long-term capacity.

102 . Bishop"^ and Vesic^ have pointed out that it is unreason­able to relate the long-term load capacity of piles in clay to the un­drained shear strength. Designs based upon effective (or drained) strength parameters and effective stresses would be the more logical approach. Unfortunately, sufficient information is unavailable for ap­plication of this method of analysis, and intermediate loading and re­lated drainage conditions (between quick and infinitely slow) make the actual behavior under a given set of conditions a completely unique problem. Until such time that there are analytical methods which can account for the complex material behavior of cohesive materials, design based upon undrained shear strength and empirical factors is the most reliable alternative. Nonetheless, this method requires great care in its general application.

57

PART VI: CONCLUSIONS AND RECOMMENDATIONS

Conclusions

103. A vast amount of pile load test data has been generated by CE offices. Despite the many tests, very few were found to be adequate for critical evaluation. Carefully instrumented pile load tests such as those performed at the Old River Low-Sill Structure and at Arkansas River Lock and Dam No. 4 were found to provide the only rational means for analyzing the behavior of pile-soil systems. It was found that con­ventional static pile formulas based on classical bearing capacity theories are inadequate to explain observed behavior of piles in cohesionless soils.

IOH. Load test results indicate that the interaction behavior and soil compressibility in the vicinity of the pile tip may make the fric­tional resistance and tip resistance interdependent. The unit skin friction for piles in cohesionless soils varies somewhat erratically with depth but in general is found to increase proportionally with depth only along the upper portion of a pile to a depth corresponding roughly to 10 to 20 pile diameters. Below this point, the unit skin friction tends to remain constant with depth for piles tested in tension, and to decrease near the pile tip for piles tested in compression. The above observations are in general agreement with observations by other investigators.

105 • Load tests by the CE on piles in clay indicate that conven­tional static formulas for load capacity based on the undrained shear strength of the clay are satisfactory for piles in soft to medium clay. Limited test data on piles in stiff clay indicate uncertainties in pre­dicting the behavior of piles in such materials.

Recommendations

106. It is recommended that further research be conducted to pro­vide a better understanding of the behavior of pile-soil systems and

58

to develop improved static pile load formulas for pile design. Test data by other researchers should he carefully reviewed by the CE in con­junction with the available test data to provide a basis for development of an improved static formula for piles in cohesionless soils.

107. Recent applications of the finite element method to deep foundation problems have proven valuable for analyses of bored piles in cohesionless soils. It is recommended that this method be employed to trace the stress-deformation behavior throughout the entire pile-soil system for driven piles in cohesionless soils.

108. In order to obtain information on the nature and extent of disturbances induced by driving of piles in cohesionless soils, care­fully controlled experiments on model pile systems are necessary. The use of the stocked ring device at the U. S. Army Engineer Waterways Ex­periment Station would permit such tests on a larger scale than employed by previous investigators. Freezing techniques and the X-ray pellet technique might be profitably employed to determine density variations. It is recommended that a detailed program of investigations along these lines be developed.

59

LITERATURE CITED

1. Headquarters, Department of the Army, "Design of Pile Structures and Foundations," Engineer Manual EM 1110-2-2906, Jul 1958» Washington, D. C.

2. Terzaghi, K. and Peck, R. B., Soil Mechanics in Engineering Practice, Wiley, New York, 19 -8.

3 . Feld, J., "Discussion on Session 6," Proceedings, Fourth Inter­national Conference on Soil Mechanics and Foundation Engineering, London, Vol III, 1958, pp 180-181.

i+. Parsons, J. D., "Piling Difficulties in the New York Area," Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, Vol 92» No. SMI, Jan 1966, pp k3-8U.

5- Yang, N. C., "Relaxation of Piles in Sand and Inorganic Silt," Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, Vol 96, No. SM2, Mar 1970, PP 395-^-09»

6. Tavenas, F. A., "Load Test Results on Friction Piles in Sands," Canadian Geotechnical Journal, Vol 8, No. 1, Feb 1971, pp 7-22.

7. Tomlinson, M. J., Foundation Design and Construction, 1st ed.,Wiley, New York, 1963.

8. Horn, H. M., "influence of Pile Driving and Pile Characteristics on Pile Foundation Performance,” Notes for Lectures, American Society of Civil Engineers Seminar on Problems in the Evaluation of Pile Foundations^ New York, 1966.

9. Thorburn, S. and MacVicar, R. S. L., "Pile Load Tests to Failure in the Clyde Alluvium," Proceedings, Conference on Behavior of Piles, Institution of Civil Engineers, London, 1971, pp 1-7.

10. Vesie, A. S., "Load Transfer in Pile-Soil Systems, Proceedings, Conference on Design and Installation of Pile Foundations and Cellular Structures, Lehigh University, Bethlehem, Pa., Apr 1970,PP 7-73.

11. Davisson, M. T., "Static Measurement of Pile Behavior," Proceedings, Conference on Design and Install ation of Pile Foundations and Cel­lular Structures, Lehigh University, Bethlehem, Pa., Apr 1970,pp 159-164.

12. U. S. Army Engineer Waterways Experiment Station, CE, "Mississippi River and Tributaries, Old River Control Low-Sill Structure; Pile Loading Tests," Design Memorandum 1-B, Supplement No. 3, Jan 1958, Vicksburg, Miss.

1 3 . Fruco and Associates, "Arkansas River and Tributaries, Arkansas and Oklahoma; Pile Testing Program, Lock and Dam k >" Final Report, Apr 196 1, St. Louis, Mb.

60

lU. Hunter, A. H. and Davisson, M. T., "Measurements of Pile Load Transfer," Performance of Deep Foundations, Special Technical Publication W+, American Society for Testing and Materials, Philadelphia, Pa., 1968, pp 106-117•

15. Vesic, A. S., "A Study of Bearing Capacity of Deep Foundations," Final Report, Project B-I89, Mar 1967, Georgia Institute of Tech­nology, Atlanta, Ga.

16. Leonards, G. A., "Summary and Review of Part II of the Symposium on Pile Foundations," Pile Foundations, Highway Research Record Ho. 333? National Academy of Sciences-National Research Council, Washington, D. C., 1970, pp 55-59*

17. Vesic, A. S., "Ultimate Loads and Settlements of Deep Foundations in Sand," Proceedings, Symposium on Bearing Capacity and Settlement of Foundations, Duke University, Durham, N. C., Apr 19^5, pp 53-68.

18. Scott, R. F., Principles of Soil Mechanics, Addison-Wesley, Reading, 1963.

19 . Prandtl, L., "Uber die Härte Plastischer Körper," Nachr. Kgl. Ges. Wiss., Göttingen, Math. Phys. KLass, 1920, p

20. __________ , "Uber die Eindringungsfestigkeit Plastischer Baustoffeund die Festigkeit von Schneiden," Zeitschrift fur Angewandte Mathematik und Mechanik 1, Ho. 1, 1921.

21. Reissner, H., "Zum Erddruchproblem," Proceedings, First Inter­national Conference of Applied Mechanics, Delft, 192^.

22. Caquot, A., Equilibre des Massifs ä frottement Interne, Gauthier- Villars, Paris, I93U.

23. Buisman, A. S. K., "De Weerstand van Paalpunten in Zand," De Ingenieur, Vol 50, 19355 pp 25-28.

2k. Terzaghi, K., Theoretical Soil Mechanics, Wiley, New York, 19 -3*25. Kerisel, J., "Fondation Profondes en Milieux Sableux: Variation

de la Force Portante Limite en Fonction de la Densité, de la Profondeur, du Diamètre, et de la Vitesse d'Enfoncement," Proceed­ings , Fifth International Conference on Soil Mechanics and Founda­tion Engineering, Paris, Vol 2, 1961, pp 73-83*

26. Brinch Hansen, J., "A General Formula for Bearing Capacity," Bul­letin Wo. 11, 1961, The Danish Geotechnical Institute, Copenhagen, Denmark.

27* DeBeer, E. E., "Etude des Fondations sur Pilotis et des Fondations Directes," Annales des Travaux Publics de Belgique, Vol k6, 19^5}p 229.

28. Jäky, J., "On the Bearing Capacity of Piles," Proceedings, Second International Conference on Soil Mechanics and Foundation Engineer­ing, Rotterdam, Vol 1, 19^8, pp 100-103.

6l

29* Meyerhof, G. G., "The Ultimate Bearing Capacity of Foundations," Geotechnique, Vol 2, No. 1+, Dec 1951? PP 301-332.

30. __________ , "Recherches sur la Force Portante des Pieux," Annalesde 1'Institut Technique du Bâtiment et des Travaux Publics, Supple- ment, Vol 6, Nos. 63-6*+, 1953, PP 371-37^*

31* __________ , "Compaction of Sands and Bearing Capacity of Piles,"Journal of Soil Mechanics and Foundations Division, American So­ciety of Civil Engineers, Vol 8$, No. SM6, Dec 1959, PP 1-29«

32. __________ , "influence of Roughness of Base and Ground-Water Condi­tions on the Ultimate Bearing Capacity of Foundations," Geo- technique, Vol 5, No. 3, Sep 1955, pp 227-21+2.

33« ______ , "Penetration Tests and Bearing Capacity of CohesionlessSoils," Jburnal of Soil Mechanics and Foundations Division, Ameri­can Society of Civil Engineers, Vol 82, No. SMI, Paper 866, Jan 1956, pp 1-19.

3Î+. ______ , "The Ultimate Bearing Capacity of Wedge-Shaped Founda-tions," Proceedings, Fifth International Conference on Soil Mechan­ics and Foundation Engineering, Paris, Vol 2, 1961, p 105.

35» ________ , "Some Recent Research on the Bearing Capacity of Foun-dations, " Canadian Geotechnical Journal, Vol I, No. 1, Sep 1963, pp 16-26.

36. Berezantsev, V. G. and Jaroshenko, V. A., "Osobennosti Deform- irovanija Peschanych Osnovanii Pod Fundamentami Glubokogo Zalozenija," Osnovaniya I Fundamenty, Vol 1+, No. 1, 1962.

37» Vesic, A. S., "Bearing Capacity of Deep Foundations in Sand," Stresses in Soils and Layered Systems, Highway Research Record No. 39, National Academy of Sciences-National Research Council, Washington, D. C., 1963, pp 112-153*

38. Kerisel, J., "Deep Foundations - Basic Experimental Facts," Pro­ceedings , North American Conference on Deep Foundations, Mexico City, Vol 1, Dec I96I+, pp 5 - b b .

39* Rob insky, E. I. and Morrison, C. F., "Sand Displacement and Com­paction Around Model Friction Piles," Canadian Geotechnical Journal, Vol I, No. 2, Mar I96I+, pp 81-93*

1+0. Ellison, R. D., "An Analytical Study of the Mechanics of SinglePile Foundations," Dissertation, 1969, Carnegie-Mellon University, Pittsburgh, Pa.

1+1. Desai, C. S. and Holloway, D. M., "Load-Deformation Analysis of Deep Pile Foundations," Proceedings, Symposium on Applications of the Finite Element Method in Geotechnical Engineering, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Miss.,1-U May 1972, pp 629-65I+.

1+2. Berezantsev, V. G., Khristoforov, V. S., and Golubkov, V. N., "Load Bearing Capacity and Deformation of Piled Foundations," Proceedings,

62

Fifth International Conference on Soil Mechanics and Foundation Engineering, Paris, Vol 2, 1961, pp 11-15«

• Vesic, A. S., "Tests on Instrumented Piles, Ogeechee River Site," Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, Vol 96, No. SM2, Paper 7170, Mar 1970* pp 5 6 1 -5 8 ^ .

Furlow, C. R., "Pile Tests, Janesville Lock and Dam, Ouachita and Black Rivers, Arkansas and Louisiana," Technical Report S-68-10,Dec 1968, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Miss.

i+5• U. S. Army Engineer Waterways Experiment Station, CE, "Pile Load­ing Tests, Combined Morganza Floodway Control Structure," Technical Memorandum Wo. 3-308, Jan 1950, Vicksburg, Miss.

k6. Kaufman, R. I. and Sherman, W. C., "Review of Soils Design, Pile Loading Tests, Construction, and Performance Observations, Section 1-B Floodwall, Memphis, Tennessee," Technical Report Wo. 3-*+53j Apr 1957, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Miss.

1+7» Worth, W. L. et al., "Pile Tests, Columbia Lock and Dam, Ouachita and Black Rivers, Arkansas and Louisiana," Technical Report Wo. 3-7^1, Sep 1966, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Miss.

18. Peck, R. B., "A Study of the Comparative Behavior of Friction Piles," Highway Research Board Special Report Wo. 36, 1958, Ra­tional Academy of Sciences-Rational Research Council, Washington,D. C.

U9. Sullivan, R. A., Discussion of Paper, "Some Loading Tests on Long Pipe Piles," Geotechnique, Vol 20, Wo. 2, 1970.

50. Bishop, A. W., "Session A: Discussion; Chairman's Introduction ofPapers 1-1," Proceedings, Conference on Behavior of Piles, Insti­tution of Civil Engineers, London, 1971, pp 31-57»

51. Vesic, A. S., "Main Session 2: Discussion," Proceedings, SeventhInternational Conference on Soil Mechanics and Foundation Engi­neering , Mexico City, Vol 3? 19^9•

52. Cole, K. W., "Section A: Discussion," Proceedings, Conference onBehavior of Piles, Institution of Civil Engineers, London, 1971, pp bl-k-2.

53» Snow, R., "Telltales," Foundation Facts, Vol I, Wo. 2, Fall 1965, pp 12-13»

5k. Geymayer, H. G., "Strain Meters and Stress Meters for Embedment in Models of Mass Concrete, Summary of Information Available as of March 1967," Technical Report Wo. 6-811, Report 1, Jan 1968, U. S. Army Engineer Waterways Experiment Station, CE, Vicksburg, Miss.

55» Hirsch, T. J. et al., "instruments, Performance and Method of

63

56.

Installation," Proceedings, Conference on Design and Installation of Pile Foundations and Cellular Structures, Lehigh University, Bethlehem, Pa., Apr 1970, pp 173-177.Hanna, T. H., "The Bending of Long H-Section Piles," Canadian Geo­technical Journal, Vol 5? Wo. 3? Aug 1968, pp 150-172.

57. Headquarters, Department of the Army, "instrumentation of Earth and Rock-Fill Dams," Engineer Manual EM 1110-2-1908, Aug 1971? Washington, D. C.

6k

Table 1Procedures Used for Determination of Pile

Failure Loads at Four CE Projects

Procedure

Proj ectOld River Arkansas River Columbia Jonesville Low-Sill Lock and Dam Lock and Lock and Structure No. 4 Dam Dam

The load which pro­duced a plastic or net settlement of 0.25 in. X X X X

The load indicated by the intersection of tangent lines drawn through the initial, flatter portion of the gross settlement curve into the steeper portion of the same curve X X X

The load beyond which there was an in­crease in gross settlement dis­proportionate to the increase in load X X X

The load at which the slope of the plas­tic or net settle­ment curve was four times the slope of the elastic defor­mation curve X

The load beyond which there was an in­crease in the plas­tic or net settle­ment dispropor­tionate to the increase in load X X

The load on the gross settlement curve at which the slope equaled 0.01 in./ ton X

Table 2Summary of Information on Horizontal Effective Stresses on Driven Files

Oin Sand (after Horn)0

Reference RelationshipBasis of

Relationship

Brinch Hansen a) p = cos^ 0q rs z a) TheoryLundgren b) ps = °-8^z b) Pile test

Henry Ps = Kpqz = tan2 (l*5 + |) lz Theory

Ireland Ps = Kq = (1.75 to 3) q^z z Pulling tests

Meyerhof PsPs

= 0.5q ; loose sand z 3= l.Oq ; dense sand z 3

Analysis of field data

Sézchy Ps = K-J| % i K = 1 to 2 Theory

Mansur and Kaufman Ps = Kq ; K = 0.3 ; compression z

K = 0.6 ; tensionAnalysis of

field data

Note: q = effective vertical stress in the soil adjacent to the pileshaft at depth z .

p = effective horizontal stress acting on the pile shaft at S depth z .

K = passive earth pressure coefficient,Jr

Table 3

Summary of CE Compression Tests on Piles in Cohesionless Soils

Failure Load

PileNo. Pile Type

EffectiveDiameter

ft

EmbedmentDepthft

DrivingGW Depth Hammer

ft Type

Reported Pailure Load

Fw e s, at 10iTlptons Movement, tons

at 1 0 $ Tip Movement Correction for GWL = 0 Remarks

Old River Low-Sill Structure

1 lU BP 7 3 1 . 1 7 81.0 8.0 OR 292 366 3 1 1

2 20-in. pipe 1 . 7 5 65.O 7 - 3 OR 296 1+00 3 ^ 1

3 ll+ BP 7 3 1 . 1 7 71.0 8.0 OR 1 5 1 212 170 Bottom plates1+ l6-in. pipe 1.1+2 66.0 7 - 3 OR 3 6 1 1+00* 3 3 8

5 l6-in. pipe 1.1+2 U5.0 8.0 OR 1 1 7 1 1 + 5 1126 18-in. pipe 1 . 5 8 65.O 7 . 3 OR 3 2 9 3 7 0 3 1 3

7 18-in. pipe 1.50 65.O 9 .1+ OR 3 1 7 360 2 9 3

Arkansas River Lock and Dam No . 1

C-8 18-in. concrete 1.50 50.0 21.0 200C 1+1+5 __ __G-8 20-in. concrete 1 . 6 7 1+5 . 0 6.0 200C 1+00 — —G- 2 18-in. concrete 1.50 1+8.0 7 . 5 2000 3 5 2 1 + 9 0 3 5 0

Arkansas River Lock and Dam No . 2

B- 5 18-in. concrete 1 . 5 0 1+3 . 0 2 . 5 0-l6 280 3 2 5 301G- 2 lU-in. concrete 1.16 1+2.8 9 . 0 ll+OC 296 — —

5 lip-in. concrete 1.16 1+I.7 7 . 0 ll+OC 129 180 ll+ 3 Jetted to 37 ftJ- 2 Timber 1.08 3 5 - 0 6.0 65C 112 Prejetted 2 7 ft

Arkansas River Lock and Dam No ■ 3

E-ll I k BP 7 3 1 . 1 7 1+2.8 7 - 5 ll+OC 1 3 0 180 ll+O In groupE-ll lU BP 7 3 1 . 1 7 61.8 9 . 9 ll+OC 1 8 5 236 186 Second test on Pile E- UTP- 1 I k BP 7 3 1 . 1 7 52.8 1 + . 9 ll+OC 1 5 1 208 180TP- 2 I k BP 7 3 1 . 1 7 62.8 1 + . 9 ll+OC 169 21+1+ 2 1 5

TP- 3 I k BP 7 3 1 . 1 7 7 3 . 0 1 + . 9 ll+OC 226 272 2 l+ 3

k-8 lU BP 7 3 1 . 1 7 3 5 - 0 1.2 S-8 122 180 170k - 9 ll+ BP 7 3 1 . 1 7 3 5 - 5 7 - 7 S-8 1 3 7 180 1 3 l+F- 9 lit- BP 7 3 1 . 1 7 65.0 5 - 2 ll+OC 220 — —G- 7 Timber 1.08 3 8 . 3 6 . 5 C- 5 9 5 120 9 9 In groupG-8 lU BP 7 3 1 . 1 7 1+3 - 0 5 . 2 Foster 8 5 130 108 In groupJ- 3 lU BP 7 3 1 . 1 7 1+6.7 8 . 9 Foster 1 0 5 1 8 5 ll+O In group

Arkansas River Lock and Dam No . 1+

l 12-in. pipe 1 . 2 5 5 3 . 1 2 . 5 ll+OC ll+O 172 1 5 9

2 l6-in. pipe 1.68 52.8 2 . 5 ll+OC 1 9 5 2 3 7 2192 16-in. pipe 1.68 52.8 2 . 5 ll+OC 210 2 l+ 5 227 Second test on Pile 2

3 20-in. pipe 1 . 8 5 5 3 - 0 2 . 5 ll+OC 2 1 5 262 21+31+ l6-in. concrete 1 - 3 3 1+0.2 2 . 5 ll+OC 170 — —5 l6-in. concrete 1 . 3 3 51.0 2 . 5 ll+OC 21+0 2 8 5 271+6 I k BP 7 3 1 . 1 7 1+0.0 2 . 5 8 0 C ll+O 180 1637 I k BP 7 3 1 . 1 7 52.1 2 . 5 8 0 C 190 2 3 5 2178 Timber 1.08 38.6 2 . 5 65C 80 — —9 I k BP 7 3 1 . 1 7 5 3 - 2 2 . 5 Bo dine 210 2 1 + 9 23010 16-in. pipe 1.68 5 3 . 1 2 . 5 Bodine 180 2 3 5 21711 l6-in. concrete 1 - 3 3 38.8 2 . 5 Bodine 150 — —16 l6-in. pipe 1 . 5 0 5 2 . 7 2 . 5 ll+OC ll+O 1 7 5 162 Jetted to 1+ 0 ft

C- 3 l6-in. concrete 1 - 3 3 3 9 - 6 2 . 5 0-16 27I+ 310 281L- 1 0 l6-in. concrete 1 - 3 3 3 7 - 5 5 - 5 0-16 217 250 206 Jetted to 33 ftH- 3 Timber 1.00 5 0 . 3 6.1+ 65C 100 120 100 Jetted to 1+6 ftB- 2 Timber 1.00 1+7 . 3 3 - 6 65C lll+ 120 112 Jetted to 37.3 ft.B-1+ 18-in. concrete 1.50 1+5 .1+ 1+.8 0-16 5 0 6 580 501+C- 2 18-in. concrete 1.50 1+1 + . 9 0.0 0-16 389 500 500

Jonesville Lock and Dam

1 l8-in. concrete 1 . 5 0 38.0 7.0 0-16 3 5 6 1+10 3252 18-in. concrete 1.50 1+5 . 0 7.0 0-16 3 0 3 3 l+ 0 2763 18-in. concrete 1.50 5 1 + . 0 7.0 0-16 3 l+ 7 3 9 0 3 2 5

2 A 18-in. concrete 1 . 5 0 1+5 '.0 7.0 0-16 196 250 203 Jetted to 3 9 ft

Note: GW denotes groundwater; GWL denotes groundwater level; BP denotes bearing pile.* Estimated.

Table 1+

Summary of CE Tension Tests on Piles in Cohesionless Soils

Failure Load Based on 10% Tip Movement

PileNo. Pile Type

Effec­tive

Diam­eterft

Embed­mentDepth

ft

GWDepthft

DrivingHammerType

ReportedFailure

Loadtons

Corrected Failure Q Load -UQ for GWL = 0tons tons

AverageUnitSkin

Friction tons/ sq ft K* tan <5 s 6 , deg

tKs Remarks

Old River Low-Sill Structure

2 20-in. pipe 1.75 65.0 9.4 OR 135 200 176 0.49 0 .5 4 1 25 I.I6

3 lb BP 73 1.17 71.0 9.4 OR 50 74 68 0.20 0.203 28.5 O .3 7 Bottom plates4 l6-in. pipe 1.42 66.0 9-4 OR l6l 200 178 0 .6l 0.660 25 1.415 l6-in. pipe 1.42 45-0 9.4 OR 55 84 59 0.30 0.470 25 1 . 0 1

6 l8-in. pipe 1.58 65.0 9-4 OR 137 185 162 0.50 0 .5 5 4 25 1.19

Arkansas River Lock and Dam No. 1

C -8 l8 -in. concrete 1.50 50.0 2 1 .0 200C 210 240 145 0.48 0.620 30 I .08

Arkansas River Lock and Dam No. 2

B-5 l8 -in. concrete 1.50 43.0 2.5 O-I6 100 150 135 0 .52 0 .78 5 30 I .36G -2 14-in. concrete 1 . 1 6 42.8 9.0 140C 105 1 18 85 0.43 0.645 30 1 . 1 2

5 lU-in. concrete 1 . 1 6 41.7 7.0 140C 53 70 42 0.22 0.338 30 0.59 Jetted to - 37 ft

J -2 Timber 1 .0 8 35.0 6.0 65C 41 62 48 0.40 0.735 30 1.27 Prejetted 27 ft

Arkansas River Lock and Dam No. 3

G -8 lb BP 73 1.17 43.0 5.0 Foster 25 58 48 0.24 0 .35 2 28.5 0 .65 In groupE-ll 14 BP 73 1.17 42.8 7.5 140C 34 67 51 0.25 0 .38 3 28.5 0.70 In groupJ-3 lb BP 73 1.17 46.7 9.2 Foster 31 80 59 0.27 0 .372 28.5 0.69 In groupTP-1 14 BP 73 1.17 52 .8 4.9 140C 40 55 49 0.20 0.229 28.5 0.45TP-2 14 BP 73 1.17 62.8 5.1 I4OC 51 80 69 0.23 0.240 28.5 0.44G -7 Timber 1 .0 8 38.3 7.6 0-5 31 50 37 0.29 0.485 30 0.84 In group

Arkansas River Lock and Dam No. 4

1 1 2 -in. pipe 1.25 53.1 2.5 140 c 70 92 85 0 .4 1 0 .4 9 1 25 1.052 l6-in. pipe 1 .6 8 52 .8 2.5 140C 91 1 1 6 107 0 .38 0.462 25 0.993 20-in. pipe 1.85 53.0 2.5 140 c 90 120 110 0 .36 0.433 25 O.934 l6 -in. concrete 1.33 40.2 2.5 140 c 71 95 86 0 .40 0.642 30 1 . 1 1

7 14 BP 73 1.17 52.1 2.5 80c 41 70 64 0 .26 0 .3 2 2 28.5 0.598 Timber 1 .0 8 38 .6 2.5 65c 25 37 33 0 .2 5 0 .4 19 30 0.90

10 l6 -in. pipe 1 .6 8 53.1 2.5 Bodine 87 1 10 100 0 .3 5 0 .4 3 1 25 O.9316 l6-in. pipe 1.50 52.7 2.5 14OC 63 78 71 0 .2 7 0.335 25 O.72 Jetted to

40 ftC-3 l6 -in. concrete 1.33 39-6 2 . 1 0 -I6 113 139 127 0.60 O.975 30 1 .6 9

L -10 l6-in. concrete 1.33 37.5 5.1 O-I6 93 106 85 0 .4 1 0.682 30 1 . 1 8 Jetted to33 ft

H-3 Timber 1 .0 0 50 .3 6.3 65c 25 40 31 0.20 0.248 30 0.43 Jetted to 46 ft

B -2 Timber 1 .0 0 47.3 4.7 65c 55 60 51 0.34 0.465 30 0.34 Jetted to 37-3 ft

Arkansas River Lock and Dam No. 6

K -8 14 BP 73 1.17 39.3 9.7 0 -10 70 120 83 0.47 0.733 28.5 1.35

Jonesville Lock and Dam

1 l8 -in. concrete 1.50 38.0 7.0 O-I6 88 130 99 0.44 0 .726 30 1 .2 62 l8 -in. concrete 1 .5 0 45-0 7.0 O-I6 115 150 1 1 2 0.42 0.620 30 I .08

3 l8 -in. concrete 1.50 54.0 7.0 0 -16 1 1 2 130 1 1 5 0 .36 0 .38 6 30 0 .6 72A 1 8 -in. concrete 1 .5 0 45.0 7.0 O -16 69 80 56 0 .2 1 0 .3 3 1 30 0 .58 Jetted to

39 ft4 l8 -in. concrete 1.50 45-0 10 .0 0 -16 97 130 89 0.33 0.494 30 0.86

Note: GW denotes groundwater; GWL denotes groundwater^vel; BP denotes bearing pile.

Table 5

Summary of Average fs and K*" s Values from TensionTests in Cohesionless Soils

Average Unit Skin Friction fs tons/sq ft

Coefficient of ^ Lateral Pressure Ks

Pile TypeNo. of Tests

Range of Values Average

Range of Values Average

Single H-piles 4 0.20 to 0.26 0.22 0.37 to 0.59 0.46

H-piles in group 4 0.24 to 0.29 0.26 0.65 to 0.84 0.72

Single displacement piles (concrete, steel pipe, and timber) 13 0.27 to 0.61 0.40 O .67 to i.4l 1.03

Displacement piles in group 4 0.43 to 0.60 0.51 I .08 to 1.6 9 1.31

Displacement piles jetted 6 0.20 to 0.4l 0.28 0.34 to 1 .18 0.64

Table 6Maximum Unit Skin Friction for Compression P iles in Cohesionless Soils

P ro jec tTestP ile P ile Type

Old River Low- 2 20-in.-0D pipeS i l l S tructure 3 1UBP73

1+ l6-in.-QD pipe5 l6 -in .-0D pipe6 l8 -in .-0D pipe

Arkansas River 2 l6 -in .-0D pipeLock and DamNo. k 3 20-in.-0D pipe

7 ltaP7316 l6 -in .-0D pipe

Maximum Unit Skin

F r ic tio n f s

tons/sq f t

Depthz

f t

P ileDiameter

f t

P ileDepth/

DiameterRatio

0.9b 1+7 1.75 26.9

0.52 23 1.17 19.71.0 8 34 1.1+2 23.90.93 25 1.1+2 17.61.13 28 1.58 17.7

0 . 1+9 16 1.68 9-50.21+ 15 1.85 8.10.65 20 1.17 17.10.1+2 36 I .50 21+.0

to n s/sq f t fA 6 , deg

Maxi­mumKCs

1.51+ 0.61+9 25 1 .3 90.892 0.583 28.5 1 .0 71.18 0.915 25 1.960.9^8 0.976 25 2.09

1.01 1.12 25 2.1+0

0.992 0.1+91+ 25 1.06

0.930 0.258 25 0.551.21+ 0 . 521+ 28.5 0.96

2.23 0.186 25 0 . 1+0

Table 7

Maximum U n it S kin F r ic t i o n f o r T ension P i l e s in C o h e s io n le s s S o i ls

P r o je c tT estP i le P i le Type

Old R iver Low- 2 20- i n . - 0D p ip eS i l l S tru ctu re

3 1UBP73k l 6- i n . - 0D p ip e

5 l 6- i n . - 0D p ip e

6 l 8- i n . - 0D p ip e

Arkansas R iver 2 l 6- i n . - 0D p ip eLock and DamNo. k 3 20- i n . - 0D p ip e

7 l t a P 7316 l 6- i n . - 0D p ip e

MaximumU n it S kin F r ic t i o n

fs

to n s / s q f t

P i l eDepth

zf t

P i l eD iam eter

f t

P i l eD epth/

D iam eterR a t io

o .6i 33 1.75 1 8 .9

0.32 1 7 1.17 11+.5

0.71+ 32 1 . 1+2 2 2 .5

0.1+2 11+ 1 . 1+2 9 .9

0.59 23 1 .5 8 11+.6

0.1+7 10 1.68 5 .6

0.35 21+ 1 .8 5 1 3 .0

0.25 23 1.17 1 9 .7

0 .1(1 20 1 .5 0 1 3 .3

tons/sq f t f / qS/ z 6 5 deg

Maxi­mum

K*s

1 .2 2 0.500 25 1 .0 7

0.767 0 . 1+17 28.5 0.77

1.19 0.621 25 1-33O.683 0 .615 25 1.32

0.935 0 .631 25 1.35

O.388 1 .2 1 25 2.60

0.822 0.1+26 25 0 .9 1

0.791 0 .3 16 28.5 0.58

O.698 0.587 25 1 .2 6

Table 8Computed Friction Aggies for Compression Tests in Cohesignless Soils

Estimated

ProjectPileWo. Pile ïÿpe

Overburden Pressure tons/sq ft

Failure Load in tons

Observed Loads, tons Butt Tip*

Tip Load at Failure

tons w_a AOld River Low- 2 20-in.-0D pipe 1 .8 0 1+00 332 98 16 6 1+2 35Sill Structure 3 lteP73 1.90 212 2k5 13 8 10 5 1+0 35

1+ l6-in.-0D pipe 1 .8 3 1*00 3ko 136 19 6 77 k35** l6-in.-0D pipe 1.1+6 145 lk5 1+8 1+8 21+ 316 l8-in.-0D pipe 1.79 370 3^5 12 6 1 5 1 U8 37

Arkansas River 2 l6-in.-0D pipe 1.6k 237 250 118 10 3 1+0 35Lock and Dam No. b 3 20-in.-0D pipe 1.6k 262 257 111 1 1 6 3 1 33

7 1UBP73 1.6l 235 255 97 77 35 3^l6t l6-in.-0D pipe 1 .6 5 175 16 2 10 5 1 1 8 1+8 37

* From load distribution curves adjusted for residual loads induced by driving.** Pile tip in silty sands, t Jetted to bO ft.

APPENDIX A: COMPILATION OF DATA

1. A multiple letter dated 2 June 19^7 was sent to all Division and District offices in CONUS and to the Pacific Ocean Division and Districts requesting titles of published reports and copies of un­published reports concerning pile tests. In response to this request, pile load test data for 578 load tests were received from 22 CE offices. The load test data varied in form from rather sketchy field notes to comprehensive reports in bound volumes. Information on some load tests was furnished complete with detailed supporting data, such as laboratory tests on foundation soils, while other reports included no data as to soil type. The data received varied considerably with respect to the type of piles involved, pile alignment (vertical or batter), method of installing, direction and Intensity of test load, and rate of load application.

2. The pile test data were grouped according to District or Di­vision. A legend describing the symbology for source, soil conditions, and pile type is shown in table Al. The test data are tabulated in tables A2-A8. Tables A2 and A3 show results of compression and tension tests, respectively, performed on single vertical piles. Results of lateral load tests on single vertical piles are tabulated in table Ai+. Tables A5 and A6 show results of vertical and axial load tests, respec­tively, on single battered piles. Results of vertical load tests on pile groups are tabulated in table A7. The data compiled for instru­mented piles are shown in table A8.

Al

Table AlLegend for Tables A2 Through A8

District or Division Generalized Soil Condition Type of Pile MiscellaneousSymbol Name Symbol Meaning Symbol Meaning Symbol Meaning

LMM Memphis District Cl Clay or clayey Wd Wood tr TraceLMN New Orleans District Sd Sand or sandy Pipe Steel pipe w/ WithLMK Vicksburg District Si Silt or silty BP Bearing pile OD Outer diameterLMS St. Louis District G Gravel or gravelly C Reinforced concrete ID Inner diameterMRK Kansas City District 0 Organic PC Prestressed concrete CE Closed end

MRO Omaha District F Fine Mono Monotube pileNAN New York District M Medium CIP Cast^in-placeNAO Norfolk District C Coarse CCC Centrifugally cast concreteNAP Philadelphia So Soft RST Raymond step taper

DistrictNOS St. Paul District St Stiff

NPA Alaska District MSt Medium stiffNPS Seattle District B1 BlueORP Pittsburgh District Gr GrayPOF Far East District Wh WhitePOH Honolulu District

P00 Okinawa DistrictSAM Mobile DistrictSAS Savannah DistrictSPL Los Angeles

DistrictSPN San Francisco

District

SWL Little Rock District

RED New England Division

Table A2Compression Load Tests on Single Vertical Piles

Dis­trict

Driving DataMaxTest

Divi- of GWL* Test Pile Em- Type of Energy Last Load Loadsion Pro,! ect Tests Generalized Soil Conditions ft No. Type of' Pile Driven bedded Hammer ft-lb ft tons tons RemarksLMM Wolf River Floodwall, Sec­ 195*t 0-20 ft fill, 20-80 ft Cl Si and 6-13 l-N(N) Pipe, 12-3A in. OD (CE) 55.0 Vulcan 5OC 15,100 19 60 1+2 Wall thickness = 3/8 in.

tion 1-B, Memphis, Tenn. Si Cl w/strata of fat Cl andSd Si

l-N(S) Pipe, 12-3/1+ in. OD (CE) 70.0 18 80 71 Wall thickness = 3/8 in.1-A Pipe, 12-3 /*+ in. OD (CE) 60.0 19 70 52 Wall thickness = 3/8 in.57 C-l6 in. sq 1+5.0 30 100 8U62(S) C-l6 in. sq 1+0.0 25 80 6062 (N) C-l6 in. sq 55.0 51 100 907*4(S) C-l6 in. sq 1+0.0 1+1+ 70 50

Wolf River and Nonconnah Creek Floodwall Memphis, Tenn.

1939

GM & 0 RR Bridge, M-l+36.58 1967S. Fork, Obion River,Obion City, Tenn.

LMN Morganza Floodway Control Structure

19*49

Morganza Floodway,New Orleans, Texas and Mexico Railway Co.

19*+0

.0-10 ft Sd Si, tr Cl, G, 0 IQ-60 ft Si, some Cl, tr Sd

0-15 ft Si Cl, 15-*+0 ft F Sd UO-70 ft Sd G, 70- ft Cl Sd

0-27 ft Cl, 27-35 ft Sd Si & Si Cl, 35-5*7 ft Cl, 5*1-77 ft Cl Si, Sd Si,;Si Sd, Si Cl, 77-? F Sd

10-13 1/3 +83.5

1/36 + 1+1.5 2/25 + 80 3/36 + bo 6/lb + 30 6/36 + 82.1+3

O-30 ft Cl, 30-1+0 ft Sd Si, I+O-65 ft Cl, 65-75 ft Si Cl

0-27 ft Cl, 27-35 ft Sd Si, & Si Cl, 35-5*1 Cl, 5*1-77 ft Cl Si,Sd Si, Si Sd, Si Cl, 77-? F Sd |

0-20 ft Cl, 20-35 ft Cl Si & Sd Si 35-60 ft Cl, 60-70 ft Cl Si, 70- 75 ft Cl, 75-78 ft Sd Si, 78-100 ft Sd

O-3O ft Cl, 30-1+0 ft Sd Si, I+O-65ft Cl, 65-75 ft Si Cl, 75-100ft Sd

0-5 ft Cl, 5-10 ft Cl Si, 10-15ft si ci, 15-30 ft Cl, 30-1+0 ft Si Cl, 1+0-55 ft Cl

tà, butt-ll+-l/l+ in. Tip-9 in.

in. Tip-8-1/2 in.

(Continued)

52.O Vulcan 1 15,000

60 1+5.01+0 35.01+0 23.51+0 28.01+0 33.0

10I6O302327

2525252525

1 ll+ BP 117 73.3 Vulcan 06 19,500 51 150 —

C-l-a Pipe 2l+ in. OD (CE) - 58.6 Vulcan-OR 30,225 15 135 120 Wall thickness =

C-l-b Pipe 2l+ in. OD (CE) 71.8 Vulcan-OR 30,225 265 390 355 Wall thickness =C-2-a Mono 8 in. Tip -- 67.5 Vulcan 1 15,000 13 80 60 Tapered sectionC-2-b Mono 8 in. Tip — 8I .8 167 170 150 Tapered sectionC-3-a Mono 12 in. OD — 67.6 9 60 bo Constant sectionC-3-b Mono 12 in. OD — 75-9 322 150 ll+O Constant sectionC-l+-a Pipe 18 in. OD (CE) — 67.1 21 90 80 Wall thickness =C-l+-b Pipe 18 in. OD (CE) - - 75.O 26O 270 250 Wall thickness =C-5-a Pipe 2l+ in. OD (CE) — 67.5 Vulcan-OR 30,225 lb li+o 90 Wall thickness =C-5-b Pipe 2l+ in. OD (CE) — 79-9 230 31+0 2i+0 Wall thickness =C-6-a Pipe 30 in. OD (CE) — 67.5 23 180 160 Wall thickness =C-6-b Pipe 30 in. OD (CE) 75-3 35*4- 1+00 1+00 Wall thickness =C-7-a C 22 in. sq - 6b.6 1+0 li+o 80

C-7-b C 22 in. sq - 86.7 299 320 2l+0

T-l Pipe 2l+ in. OD (CE)

"

6b.6 I60 85 Wall thickness =

3/8 :

3/8 in 3/8 in 3/8 in 3/8 in 3/8 in

T-2 Mono 8 in. Tip 8O .2 Vulcan 1 15, 000 __ 305 85 Tapered section

T-3 Mono 12 in. OD 75-9 Vulcan 1 60 50 Constant sectionT-1+ Pipe 18 in. OD (CE) - - 7I4.9 Vulcan 1 160 85 Wall thickness = 3/8 in.T-5 Pipe 2l+ in. OD (CE) -- 79-9 Vulcan-OR 30,225 193 85 Wall thickness = 3/8 in.T-6 Pipe 2l+ in. OD (CE) -- 93.3 Vulcan-OR 30,225 326 85 Wall thickness = 3/8 in.

T-l Wd, butt-12-1/2 __ 52.O Vulcan 1 15,'000 16 66 __

* Ground water level (l of 2l+ sheets)

D is­t r i c t

o rD iv i­s i o n ___________ P ro je c t_________

LMN Morganza FloodwayI New O rlean s, Texas and

Mexico R ailw ay Co.

Dateo f

T e s ts G en era lized S o i l C onditions

DepthGWL T est P ile f t No.

19^0 0 -30 f t C l , 30-1+0 f t S i C l , 1+0-60f t C l

0-35 f t C l , 35-^5 f t S i Sd , 1+5-75 f t C l , 75-78 f t veg m atter, 78-92 f t Sd

0-10 f t C l S i , 10-35 f t C l , 35-^0 f t S i , 55-65 f t Sd & C l

0 - 10 f t c i s i , 10 - 15 ' f t S i C l, 15 -6 5 f t C l , 65-100 f t wet Gr Sd & C l

0-35 f t C l , 35-^5 f t Si C l, 1+5-60 f t C l , 6O-8O f t Gr Sd & C l

0-1+0 f t C l 1+0-50 f t S i C l , 50-65f t F Sd S i , 65-90 f t wet Gr Sd & C l

0 - 1+0 f t C l 40-55 f t S i , 55-85 f t B1 C l & S i

T - 3 1

T-32

T-33

T-3I+

0 -30 f t C l , 30- 1+0 f t S i C l, 1+0-55 f t C l, 55-80 f t B1 C l & S i 8O-9O f t Gr Sd1 & C l

0-35 f t C l, 35-55 f t S i , 55-70 f t So B1 C l & Sd

T-36

O-3O f t C l ,30- 1+0 f t S i C l, I+O-60 f t C l

0-30 f t c i , 30-1+0 ft s i c i, 1+0-50f t C l, 5O-8O f t B1 Gr C l, 8O-IOO f t Gr Sd

0-30 f t c i , 30-1+0 f t s i c i , 1+0-80f t C l , 8O-IOO f t Sd

M -l

M-2

O-8O ft Cl, 80- 1 1 0 f t Cl & Sd M-3

0-50 f t s i c i, 50-100 ft Cl S - l

o-i+o f t s i c i 5 1+0-90 f t Cl S-2

LMN C a lc a s ie u R iv e r S a ltw a te r 1966 B a r r ie r

f t Sd 1

f t Sd

T ab le A2 (Continued)

__________ D riv in g Data__________Length o f Blows Max F a i l -

P i l e , f t p er T est ure

Type o f P i l e D rivenEm­

beddedType o f Hammer

Energy f t - l b

L astf t

Loadton s

Loadton s Remarks

Wd, b u t t - 17-3Ain . T ip -11- 1 / 2 in .

- 59.5 Vulcan 1 1 5 .,000 13 96 <96

Wd, b u t t - 17- 1 / 2 in . T ip -8 in ..

" 8 5 .0 26 10 0 - -

Wd, b u t t - l 8 - 3 Ain . T ip -8 - 1A in .

81+ .0 29 80

Wd, b u t t - 19 in . T ip -8 - 1/ 2 in .

" 8 7 .0 3 1 80 "

Wd, b u t t - l8 in . T ip-8 in .

- 79.0 55 80

Wd, b u t t -18 in . T ip-8 in .

— 8I+.0 ll+ 80 —

Wd, b u t t - I 6 - 3 A in . T ip -8 - 1/2 in .

- - 8 3 .0 8 n o < ; i io

Wd, b u tt - 19 in .T i p - 8 - l A in .

— 8 6 .0 37 80 - -

Wd, b u t t - l6 in . T ip-8 in .

6I+.0 l+l 80. - -

C-Oct (hollow ) b u tt-30 in .

- - 5 1 .0 Vulcan 0 2 A 375 58 8 7 . < 7 8

C-Oct (hollow ) b u tt-30 in .

- 8 1 .5 Vulcan 0 2l+, 375 1 93 12 7 . < 1 2 7

Mono 8 in . T ip - - 9^.5 Vulcan 1 1 5 ,«300 230 12 5 . - - Tapered se c t io n

Mono 8 in . T ip 98.7 1 1 1 202 - - Tapered se c t io n

Mono 8 in . T ip 10I+.5 6 1 15 0 - T a p e r e d 'se c t io n

2 1 BP 96 — 9 2 .0 119 15 0 12 0

21 BP 96 8 8 . 1 138 11+5 -

Wd, b u tt- ll+ in . T ip -10 in .

5 2 .2 Vulcan 06 19,500 1+8 1+0 - -

Wd, b u t t - i t in . T ip -10.5 in .

1+0.0 Vulcan 06 19,500 ll+ 1+0 - -

(Continued) (2 o f 2l+ sh e e ts)

Table A2 (Continued)

Dis­trictorDivi-sion

LMN

__________ Project

Calcasieu River Salt­water Barrier

Freshwater Bayou Lock

Port Allen Indian Village Waterway

Port Allen Indian Village Waterway

Port Allen Indian Village Waterway

Algiers Lock and Canal Plaquemines Parish Pump Station

Algiers Lock and Canal Plaquemines Parish Pump Station

Algiers Lock and Canal Orleans Parish Pump Station

Algiers Lock and Canal Orleans Parish Pump Station

V. A. Hospital (Group l)

V. A. Hospital (Group 2 )

DateofTests Generalized Soil Conditions

1966 0-? ft Sd

1965

1956 0-80 ft Cl

1956 0-80 ft Cl

1956 0-80 ft Cl

1952 0-66 ft Cl, 66-8O ft Si Sd

0-66 ft Cl, 66-8O ft Si Sd

O-56 ft Cl, 56-72 ft Si Sd 72-80 ft Cl

0-56 ft Cl, 56-72 ft Si Sd 72-80 ft Cl

19^7 0-10 ft fill, 10-bb ft Cl, ^1-67ft Sd w/Si & Cl, 67-91 ft St Cl 91-97 ft Si & Sd

Driving Data

DepthLength of Pile, ft

Blowsper

MaxTest

Fail­ure

GWLft

Test Pile No. Type of Pile Driven

Em­bedded

Type of Hammer

Energy ft-lb

Lastft

Loadtons

Loadtons

- 3 Wd, butt-1^ in. Tip-9.5 in. - 28.3 Vulcan 06 1 9,500 6 bo

— 1 Wd, butt-l4 in. Tip-8 in.

29.0 Vulcan 5OC 1 5,100 3 20 20

-- 2 Wd, butt-l4 in. Tip-8 in.

66.0 39.0 Vulcan 50C 15,100 7 Uo

” 3 Wd, butt-l6 in. Tip-9 in.

70.5 45.0 Vulcan 5OC 15,100 19 bo.

1 Wd, butt-l4 in. Tip-8 in.

72.5 69.3 Vulcan 1 1 5,000 18 10 70

2 Wd, butt-l4 in. Tip-8 in.

80.1 65.2 Vulcan 1 1 5,000 28 85 75

— 3 Wd, butt-l4 in. Tip-8 in.

92.2 79-2 Vulcan 1 1 5,000 100 115 115

1 Wd, butt-13-3Ain. Tip- 6.b in.

7^.2 57-6 Drop hammer 3 0,000 3 28 25

" 2 Wd, butt-13.b Tip-6.U in.

73.7 Gb.b h bo. 31

1 Wd, butt-13-3/8 in. Tip- 6 .7 in.

73.0 61.3 5 3b 30

2 Wd, butt-l4 .6 in.

7 ^.0 69.5 6 bo 36

— A Wd, butt-15 in. 91.7 91.5 Vulcan 1 1 5,'000 50 125 __Tip-9 .5 in.

Remarks

Jetted between b2 and 67 ft

D Mono 12 in. 0D 106.2 9 ^.0

E Raymond Com­posite RST (top) butt- l6: in. Pipe bottom 10-3A in. 0D

89.7 87.O

F Pipe 12-3A in- OD (CE)

99.0 9^.2

G Mono 8-in. Tip 101.1 9I.O

A Pipe 12-3A in- OD (CE)

97.3 93.5

28 — Jetted between 53 and 8R ftlower 30 ft tapered

15 — Jetted between ^0 and 68 ft

62 " Jetted between b2 and 76 ft, wall thickness = l A in.

133 ■■ Jetted between Ul and 66 ft, 79-88 ft, tapered section

Jetted to 62 ftwall thickness = l A in-

30

(Continued) (3 of 2k sheets)

Dis­trictor Date DepthDivi- of GWL Test Pilesion _________ Project___________ Tests Generalized Soil Conditions ft ho.LMN V. A. Hospital (Group 2) 1947

V. A. Hospital (Group 2)

V. A. Hospital (Group 2)

V. A. Hospital (Group 3)

0-8 ft fill, 8-22 ft Cl, 22-39 — Bft Cl, 39-67 ft Sd, 67-80 ft Cl 8O-83 ft Si & Sd, 83-91 ft Cl 91-97 ft Si & Sd

C

D

A

D

E

F

G

Old River Control Low-Sill Structure

1955 o-4o f t sd s i , ¿ fo - te 'ftc i, 42-52ft Sd Si, 52-80 ft F & M Sd

O-il.5 ft Sd Si, 4.5-15 ft Sd Si 15-17 ft Sd, 17-39 ft Si w/some Sd, 39-^2 ft Cl w/si seams 42-52 ft Si Sd, 52-8O ft F&M Sd tr of G

46 1

46 2

34

56

7

LMK Steele Bayou Drainage Str 1966 — — 4Columbia Lock & Dam 1965 0-55 ft St Cl, 55-75 ft dense Sd, 5-10 1

75-100 ft Cl w/Si & Sd lenses2

3

4

Planned penetration of 75 ft was never reached.** Extrapolation of tip movement curve. Failure load taken at 0.25 in movement of pile tip.

A2 (Continued)

Driving DataLength of Blows Max Fail­Pile,, ft per Test ure

Em­ Type of Energy Last Load LoadType of Pile Driven bedded Hammer ft-lb ft tons tonsRaymond Com­ 99.3 93.1 Vulcan 1 15,000 53 150posite RST(top) butt- 16 in. Pipe bottom 10-3/4in. OD

Mono 8 in. Tip 95.4 87.5 54 -

Wd, butt-15 in. 91.6 8I.O 34 150 l4oTip-9.5 in.

Pipe 12.75 in. 97.4 8O.O 88 150 __diam (CE)

Mono 8 in. Tip 97.0 80.5 100 -Raymond Com­ 97.5 80.I 55 __ __posite pipe and steptaper

Mono 12 in. OD IOO.5 83.5 38 150 l4o.Wd, butt-15 in. 90.8 80.5 16 __ __Tip-10 in.

14 BP 73 81.9 80.5 Vulcan (DR 30 ,î>25 28 350 292.

Pipe 21 in. 66.6 65.1 4o 375 296diam (CE)

14 BP 73 71.0 70.6 20 240. 151Pipe 17 in. 67.8 66.3 95 400diam- (CE)

Pipe 17 in. 46.5 ^5.1 3 145 117.diam (CE)

Pipe 19 in.(CE) 66.4 65.1 60 380 329diam

Pile 18 in. 66.5 65.0 73 380 317diamconcrete filled .

l4o10 BP 57 — 55.7 Vulcan 1 15,000 33 --

14 BP 73 64.9 63.0 Vulcan 140C 36,000 30 280 245**

l4 BP 73 52.1 51.0 Vulcan 140C & 36,000 >70-75*300 475**Raymond 0000 48,750

l4 BP 73 83.0 81.6 Raymond 0000 48,750 >479 300 320**

Pipe 18 in. OD 63.5 62.6 Vulcan !i4oc 36,000 50 300 4oo**fOE)

(Continued)

Remarks

Jetted to 62 ft

Jetted to 62 ft* tapered section

Jetted to 65 ft

Wall thickness = l/4 in.

Tapered section

Lower 30 ft tstpered

Wall thickness"^ 3/8 in.

Wall thickness = 3/3 in.

Wall thickness = 3/8 in.

Wall thickness = 3/8 in. Wall thickness = 3/3 in.

Piles were instrumented with SR-4 strain gages to get load distribution

Piles were instrumented with SR-4 strain gages to get load distribution

Piles were instrumented with SR-4 strain gages to get load distribution

Piles were instrumented with SR-4 strain gages to get load distribution

Wall,thickness = 7/l6 in.

(4 of 24 sheets)

Dis-trictorDivi­sion Project

Dateof

Tests Generalized Soil ConditionsDepthGWLft

LMK Columbia Lock & Dam 1965 0-55 ft St Cl, 55-75 ft dense Sd, 75-100 ft Cl w/Si & Sd lenses

5-10LMK Columbia Lock & Dam 1965 0-55 ft St Cl, 55-75 ft dense Sd,

75-100 ft Cl w/Si & Sd lensesLMS Lock & Dam 26, Alton, 111. -- 0-20 C Sd, 20-*l0 F Sd, *10-85 ft

C-MF Sd

Test Pile

56

Pier 3

Pier 7

Lock 26

— Pier 17

Pier 18 Pier 30

— Pier 33

0-k2 ft M-F Sd, *+2-52 ft C Sd & pea G, 52-78 ft M-F Sd

Lock & Dam 25, Cap Au Gris Mo. Tainter gate

Lock & Dam 25, Cap Au Gris Mo. Tainter gate

Roller Gate

Lock 25 Monolith 126

Lock 25 Monolith 126

0-21 ft M Sd, 21-36 ft C' Sd Pier 3w/pea G

0-8 ft MF Sd w/pea G, 8-33 ft — Pier 7F-M Sd, 33-37 ft M Sd w/G

0-1*1 ft F-M Sd, 1^-34 ft F-C Sd — Pier 10w/G, 3 *1-5*1 ft M-C Sd w/G

0-55 ft F-C Sd, tr G

Monolith 213

Table A2 (Continued)

Driving DataLength of Blows Max Fail­Pile,, ft per Test ure

Em­ Type of Energy Last Load LoadType of Pile Driven bedded Hammer ft-■lb ft tons tonsPipe 18 in. OD

(OE)C 18 in. Oct.

83.2 81.8 Vulcan 1*10C 36,000 >3000 300 18959.2 57.2 Raymond 0000 U8 ,750 >906 300 175

Wd, butt-10-3/i „ 26.0 Steam hammer 15,000 80in. Tip-8 in.

Wd, butt-12-1/2 28.0 80in. Tip-8-l/2in.

Wd, butt-13-1/1 — 36.0 — 70 —

in. Tip-10 in.Wd, butt-12-3/+ — *10.0 — 80

in. Tip-9 in. Wd, butt-13 in. 36.0 60Tip-9 in.

Wd, butt-13-l/2 _ 36.0 80in. Tip-9 in.

Wd, butt-12-l/l __ 25.0 _ 50 35in. Tip-8-l/2in.

Wd, butt-12-1/8 -- 30.0 -- 50 25in. Tip-8-3/8in.

Wd, butt-13-3/8 — 35.0 — 50 30in. Tip-8-5/8in.

Wd, butt-ll-l/2 — U o . o — 80 50in. Tip-6-l/2in.

Wd, butt-12-1/8 — 50.0 — 50 30Tip-6-5/8 in.

C 16 in. diam „ 35-0 100 87C 16 in. diam -- 30.0 — 100 90C (tapered) -- 30.0 -- 100 90butt-18 in.Tip-10-3A in.C (tapered) — 30.0 — 100 86butt-l8 in.Tip-8-l/*l in.

C (tapered) „ 35.0 100 75butt-18 in. Tip-8-l/l in.

Wd, butt-13 in. __ 31.0 _ 75 _Tip-8 in.

Wd, butt-12-l/l 32.0 _ 75 _in. Tip-8 in.

Wd, butt-12-1/1 — 31.0 __ 75 _in. Tip-8 in.

Wd, butt-17-1A __ 29.0 , __ 100 60in. Tip-13 in.

Wd, butt-l*l-l/*l — 3*1.0 __ 60 *10in. Tip-10-l/*l

Wd, butt-17 in. - 32.0 -- 100 80Tip-11 in.

RemarksJetted to depth of 63

ness = 7/l6 in.No instrumentation

wall thick-

continued) (5 of 2*1 sheets)

Table A2 (Continued)

D is­t r i c tor

D ivi­sion P ro je c t

IMS Lock 25 M onolith 227

Lock 25 M onolith 227

Lock 25 M onolith 305

Lock & Dam 18 B u rlin g to n , Iowa

L yn x v ille , Wis. Lock & Dam 9? R o lle r Gate

L yn x v ille , Wis. Lock & Dam 9 , R o lle r Gate

T a in te r ga te

Abutment M onolith C-2

MRK North Kansas C ity Unit

C e n tra l I n d u s t r ia l D i s t r i c t M onolith 2

M onolith 25

R.W. a t Mo. Pac. RR Bridge

Armourdale P ro je c t M onolith 15 S ta 279+00

Dateof

T ests

19 6

19I+8

191+8

191+8

19 9

D riving Data

G eneralized S o il C onditions

DepthGWLf t

T est P i le No. Type o f P ile

Length o f P i le , f t

Em-Driven bedded

Type of Hammer

Energyf t - l b

Blowsp er

Lastf t

MaxTestLoadto n s

F a i l ­ure

Loadto n s

O-I8 f t F-C Sd, some G, I 8- 5O f t _ Wd, b u t t -12-3 /8 _ 32.0 Steam hammer 15, 000 60 50F-M Sd, 50-? f t F-C Sd, some G in . T ip -8 -3 A& sandstone in .

O-I8 f t F-C Sd, some G, I 8- 5O f t — — Wd, b u t t - I 7- I /2 - - 31.0 - - 100 60F-M Sd, 50-? f t F-C Sd, some G in . T ip -1 3 -3 A& sandstone in .

0-6 f t M Sd, 6-1+7 f t M-C Sd, some _ __ Wd, b u t t -11-3A __ 36.0 __ 120 60G, 1+7-? f t F-C Sd, G, Sand- in . T ip -8 -3 Asto n e , lim estone in .

O-II+ f t G & Sd, l k - 2 k f t C G & Sd, _ _ Wd, b u tt-1 2 in . _ 27.6 __ 75 __2I+-3O f t M Sd, 30-? 01 Tip-8 in .

— — Wd, b u t t -12 in . — 22.3 — 75 —Tip-8 in .

— — Wd, b u t t -12-1/2 — 17 A — 7b —in . T ip-8 in .

— — Wd, b u t t -12 in . — 12.2 — 75 —Tip-1 0 -1/2 in .

0-53 f t M Sd, some G __ P ie r 2 Wd, b u t t - I 3- I /2 — 35.0 — 60 —in . T ip-10 in .

0-53 f t M Sd, someì G — P ie r 5 Wd, b u t t -12 in . — 35.0 - - 60, —Tip-9 in .

0-53 f t M Sd, some: G — P ie r 10 Wd, b u tt- ll+ - l/2 — 36.0 — 60 —in . T ip -8- 1/2in .

0 -A f t M Sd, II+-I7 f t Sd Si __ __ Wd, b u t t- 1 3 - l /2 __ 29.0 — 60 —some C l, 17-59 f t M Sd in . T ip-9 in .

0- A f t Si Sd, A -18 f t F Sd __ 1+-8C __ 33.0 Vulcan 1 91 __ __I 8- 2I+ f t C Sd, 2U-28 f t SiSd, 28-1+2 f t C Sd

— 5-1A — 2I+.0 70 — —_ 5-5 A -, — 2I+.0 80 — —~ 6-1+A C l8 in . sq 2I+.0 58 —

0-13 f t Si Sd, 13-■21 f t M Sd _ 11+-2A _ 26.O 71 __ __w/Sd S i , 2I+-32 f t C G, 21-21+f t F Sd, 32-1+6 f t C Sd

1I __ 11+-1+A -- 28.O 1+1 75 --_ 17-7 — 33.0 6 l — —1f _ 11+-8A — 3A 0 1+8 95 —

-- I - Wd, b u tt- ll+ - l/3 1+0. 1+ 39 A McK-T 9B2 8,750 ll+O — --in . T ip -9- 5/16in .

-- 2 Wd, b u tt- ll+ -3 /8 1+0.8 38.7 McK-T 9B2 8,750 72 -- --in . T ip -9- 7/8in .

— 3 C 18 in . X 18- l A 1+0.0 2I+.0 Vulcan 1 15, 000 312 — —in . Tip-5 in .

-- 1A C l8 in . sq 1+0.0 3^.5 Vulcan- OR 30,225 l+o -- --Tip-5 in .

— 1 35.0 3I+.0 21+ " "

__ 2 1+0.0 3I+.0 39 87 __53 3 1+0.0 3 h .o 30 87 60

Remarks

(6 o f 2h sh e e ts )(Continued)

Table A2 (Continued)

Dis­trictorDivi­sionMRK

MRO

NED

Driving DataLength of Blows Max Fail­Date Depth Pile , ft per Test ure

of GWL Test Pile Em­ Type of Energy Last Load ' LoadProject Tests Generalized Soil Conditions ft No. Type of Pile Driven bedded Hammer ft-:lb ft tons tons

Armourdale Unit 19b9 _ _ __ 4 C 18 in. sq 40.0 34.0 Vulcan 1OR 30,225 34 82Sta 296+90 Tip-5 in.

Central Industrial District I9b6 — — 22 C 18 in. sq 34.0 34.0' Vulcan !1 15,1000 32 70 __

Monolith 131Central Industrial District — — 4 24.8 20.0 52 52 __

Monolith 134Central Industrial District — — 9 25.O I9.O 31 45. __

Monolith 134Central Industrial District -- — 3 bo .o 39.0 29 __

Monolith 165Central Industrial District — — 3 and 4 _ 38.0 I60 _ _ __

Monolith 160Central Industrial District __ -- 8 C l6 in. sq 25.O 20.0 20 _ _ _ _

Monolith 169 Tip-3 in.Central Industrial District — — 9 C 16 in. sq 25.O 20.0 27 45 __

Monolith 169 taperedCentral Industrial District — — 4a C 18 in. sq 27.O 28.O 145 75 __

Monolith l4Central Industrial District - - - - 8a C l8 in. sq 34.0 34.0 57 95 6O.OMonolith l4

N. P. Railway Relocation 1966 0-16^ ft Cl & Si Sd, 16-28 ft Cl __ __ Wd, butt-l4 in. __ 35.0 Vulcan 4000 lb 22, '000 8 70 55.OSta 1548+20 2Ö-40 rt Cl Sd Tip-9 in. drop hammerN. P. Railway Relocation at 1965 0-20 ft Sd Cl & Cl Sd, 20-25 ft _ _ _ _ Wd, butt-19 in. 6O.3 5O.O 44,000 27 50

Bridge 11, Bent 5 Cl Si, 25-43 ft Si Sd Sc Cl, Tip-9 in. max43-60 ft Cl Si

N. P. Railway Relocation at 1965 0-5 ft lean Cl, 5-10 ft Cl Sd I8.O __ Wd, butt-l4 in. _ _ 4o.o 24,'OOO 6 100 _ _

sta I3OI+85 w/G, 10-55 ft Sd Si Tip-9 in. maxN. P. Railway Relocation at 1966 0-6 ft Eat Cl, 6-15 ft Cl Sd, _ _ __ Wd, butt^l4 in. _ 48.0 24,'000 11 100 _ _

Sta 1256+60 15-60 ft S.i Sd Cl Tip-9 in. maxE. P. Railway Relocation 1965 0-5 ft Sd Cl, 5-20 ft Cl Sd 22.5 Wd, butt-18 in. 51.0 47.0 40,000 16 50. _

Bridge 11,Bent 7 20-2b ft Cl Si, 2b-b0 ft Tip-ll-l/2 maxSi Sd & Cl, 40-60 ft Cl Si in.

N. P. Railway Relocation 1965 0-7 ft Sd Cl, 7-60 ft shale _ _ __ Wd, butt-l4 in. 42.3 35.0 40,000 18 50Bridge 8, Bent 3 Tip-10-3/4 in. maxRASA Electronics Research 1966 0-4 ft Si Sd, 4-37 ft Cl, 37-62 AA-102 Pipe, concrete- 9O.O 70.6 Vulcan 06 19,500 47 70 4o.o

Center, Cambridge, Mass. ft Cl Sd w/G, 62 -66 ft Sd Cl G 66-72 ft Sd Cl w/weathered shale and G

0-4 ft 0 Si, 4-12 ft Si Sd w/G, 12-36 ft Cl, 56-45 ft Si Sd w/G, 45-47 ft Sd Cl w/G, 47-57 ft Si Cl Sd w/G, 57-81 ft Si Sd w/G, cobbles & boulders, top of rock

0-5 ft peat & 0 Si, 5-9 ft Si Sd w/G, 9-11 ft Cl Sd w/G, 11-46 ft Cl w/G, b6- 62 ft Sd Cl, 62-66 ft Si Sd, 66-79 ft Cl w /mr of Sd & G, 79-85 ft Si Sd w/G, 85-88 ft Si Sd w/weathered rock

filled, 10-3 fb in. OD

g g-io42

8O.O 62.6 ^7 80

84 i4o

3/8. in.

Predrilled 46 ft

Predrilled 66 ft

(Continued)

(7 of 2b sheets)

Dis­trictor DateDivi- ofsion Project Tests

NED NASA Electronics Research 1966Center, Cambridge, Mass.

1966

1966

1967

1967

1967

Fox Point Hurricane Barrier 1963 Providence, R. I.

v Transit Shed Wharf, S. 1942Boston, Mass.

NAN U. S. Quarantine Station 1962Staten Island, N. Y.

NAN U. S. Quarentine Station 1962Staten Island, N. Y.

NAO Langley AFB Chapel 1964Langley AFB Medical 1964Facility

Training Command Head- i960quarters, Fort Eustis,Va.

I960

i960

1 r 1961

Improved Nike Hercules 1962System. TRR Tower Deep Creek, Va.

DepthGWL

Generalized Soil Conditions ft

0-5 ft peat & 0 Si, 5-9 ft Si Sd w/G, 9-11 ft Cl Sd w/G, 11-46 ft Cl w/G, 46-62 ft Sd Cl, 62-66 ft Si Sd, 66-79 ft Cl w/Tr of Sd & G, 79-85 ft Si Sd w/G, 85-88 ft Si Sd w/weathered rock

0-43 ft Cl, 43-45 ft Sd Cl, 45-59 ft Si Sd w/G, 59-64 ft Cl Sd G w/weathered rock, 64-70 ft weathered rock w/Si & Sd seams70- 80, ft argillite, So,highly fractured

0-10 ft fill, 10-15 ft Si Sd w/G 15-69 ft Cl, 69-71 ft Sd Si G71- 74 ft Si Cl Sd w/G, 74-77 ft Cl w/Si lense, 77-81 ft Sd Si G, 81-85 Ft Sd Si Cl G, 85-97 ft argillite

O-3O ft loose M Sd, 3O-5O ft loose Si Sd, 5O-80 ft loose Si, SO-? ft Si Sd

So B1 Cl

Test Pile No.

4

5

6

7

8

9

1 Bent 17

Bent 22

37b

15

12Soldier 1 Soldier 2 Soldier 4 Bent B-l

Bent C-ll

Bent F-19

Bent D-l6 Bent 4 Site N-52C

Table A2 (Continued)

Type of Pile

Pipe, conc- filled,10-3/4 in. OD

Pipe concrete- filled,1 2 -3 A i n - 0D

Pipe concrete- filled,10-3A i n - 0D

Pipe concrete- filled, l4 in. OD

Pipe concrete- filled, 14 in. OD

Pipe concrete filled, 14 in. OD

14 BP 89

Wd, butt-l6 in. Tip-9 in.

WdRst

Wd, butt-l4 in. Tip-7-1/2 in.

Wd, butt-15-1/2 in. Tip-7 in.

Wd, butt-l4 in. Tip-7-lA in*

Wd

Wd

(Continued)

Driving DataLength of Pile, ft

Em-Driven bedded

Type of Hammer

Energyft-lb

BlowsperLastft

110.0 87.7 Vulcan OR 30,225 90

110.0 82.8 120

100.0 80.4 Vulcan 06 19,500 81

- 104.0 McK-T 58 26,000 320

101.0 McK-T 58 26,000 288

- 93.0 McK-T 58 26,000 312

78.0 Bodine Sonicpile driver

66.4 25.6 Vulcan 1 15,000 12

~ 77.0 10

- 94.0 21

_ 24.0 656.1 52.4 Raymond 65c 19,500 11

56.1 51.9 856.1 51.8 ll64.1 62.O 1764.1 63.1 1664.1 62.7 2555-5 54.0 Vulcan 1 15,000 36

55-3 54.0 24

55.1 54.0 25

- 55.0 8

— 51.0 77

Max Fail­Test ureLoad Loadtons tons Remarks

180 __ Predrilled 52 ft

180 — Could not maintain 180 ton load,predrilled 44 ft

140 — Predrilled 53 ft

180

180

180

200 170 Vibrations 67-137 cycles per sec

48 30 Lagged(est)60 Gross settlement greater than 1 in.

90

80

8080

4o.

4o

4o.

4o

30

(8 of 24 sheets)

T able A2 (C ontinued)

D is­t r i c t

o r DateD iv i- o fs io n ___________ P ro je c t____________ T e s ts G en era lized S o il C o n d itio n sMO 200 man airm an

Langley AFB,d o rm ito ryVa.

1967 0-5 f t C l & S i w /Tr o f Sd, 5-8 f tSd w /s h e l l frag m en ts , 8-1+9 f t F Sd w /s h e l l fragm ents and Si

Langley BOMARC F a c i l i t i e s 1959Langley AFB, Va.

Langley BOMARC F a c i l i t i e s 1959Langley AFB, Va.

H e lic o p te r Shop & C lassroom s 1962 F o rt E u s t i s , Va.

1000 Seat T h ea te r 1967F o rt E u s t i s , Va.

I 1967Bare Shop, F o r t S to ry , 196k

Va.R e h a b i l i ta t io n o f P ie r 1

F a llo u t P ro te c t io n , Opera- 1963 t i o n s , Power & GTR Bldg.Cape C h a rles AFS, Va.

F a llo u t P ro te c tio n Manassas AFS, Va.

HAP Penn RR B ridge N. Tower

tPenn RR B ridge S. Tower

0-20 f t F-C Sd w/G, 20-1+0 f t C l S i Sd, I+O-65 f t F Sd w/mica & Si

0*-20 f t Sd & Sd w/G, 20-1+5 f t C l S i w/Sd, 1+5- ? f t F-M Sd (m ica)

0-20 f t Sd and Sd w/G,20-1+5 f t C l S i w /Sd, 1+5-? f t F-M Sd

D riv in g DataLength o f Blows Max F a i l ­

Depth P i le ^ ___ p er T est u reGWL T est P i le Em­ Type o f Energy L ast Load Loadf t Wo. Type o f P ile D riven bedded Hammer f t - : lb f t to n s to n s Remarks

1+.0 1 P-3 Wd, b u t t -13 in . 3 5 .0 30.0 Vulcan 1 15, 000 10 30 _T ip-9 in .

k . O 2 Wd, b u t t -12 in . 35 .0 3 0 .0 13 30 —T ip-9 in .

1+.0 3 Wd — — — 30 —i+.o 1+ Wd, b u t t -12 in . — — — 30 —

T ip-9 in ." 1+3 Wd " — Vulcan 5OC 15, 100 6 30 —- 21 Wd - Vulcan 5OC 15, 100 5 30__ Bent 0,-6 Wd, b u tt-1 4 in . 1+0.0 28.0 Vulcan 1 15,'000 56 30 __

T ip -8 in .- - Bent D-6 Wd, b u t t - 1 3 - l /2 39 .9 30.0 - - - - 60 30 —in . T ip -8 -1 /2in .

— Bent L-l+a Wd, b u tt- l l+ in . I+O.3 31+.0 - - - - 132 — - -T ip-8 in .

— Bent 0-3 Wd, b u t t - 1 3 - l /2 1+0.2 3I+.O - - 109 — - -in . T ip -7 -1 /2

” 71 Bent E-2 Wd U 2 . 0 V ulcan :L 15, 000 103 1+0 P i le j e t t e d 26 f t__ 111 Bent G - l l __ 1+2.0 I1 12 19 _" 1 Wd " " \\ - 90_ 73 Bent 8 PC l8 in . sq __ __ McK-T S10 32, 500 __ 125 _- - 72 Bent 16 - - — - - - -— 173 Bent 27 — — — —— Bent 32 — — — —— 201 Bent 1+0 — — — —- - Bent 1+9 — — — - -— 182 Bent 53 — — — —— 2 CIP 12 in . OD - - 1+3.0 - - ““ 60 “ “

- 17 CIP 16 :in . OD 1+0.0 - - - - - 90

__ 250 ll+ BP 117 __ 3 7 .0 Vulcan 80C 2l+,l+50 288 200 __l l+2 38.0 I 63 200

- - 225 5 3 .0 3OO 200— 237 39 .0 373 200

129 39 .0 198 20039 1+2.0 290 2001+5 39.O 320 20052 37 .0 31+3 200

35 59 .0 I 99 200

— 131 76.0 I 95 200

l l+2 70.0 203 200- - 211 7U.0 1+37 200

7 71.0 I 85 200.8 68.0 185 200

(C ontinued) (9 o f 2U sh e e ts )

Table A2 (Continued)

Dis­trictorDivi­sion Project

Dateof

Tests

NAI Penn RR Bridge, N. Abutment 1962

Penn RR Bridge, S. Abutment 1962

Penn RR Bridge, 1963N. Girder Pier

Penn RR Bridge, 1 9 6 3

S. Girder Pier

Penn RR Bridge, 1963S. Girder Pier

Reedy Pt. Bridge 1966

Generalized Soil Conditions.

0-25 ft F-C Sd, 25-1+5 ft Cl Si Sd i+5 — ? ft micaceous Si Sd

0-15 ft Br F Sd, Si Cl, 15-50 ft Si Sd

0-19 ft Br M Sd, 19-29 ft Gr Br F Sd, some Si & Cl, 29-6U ft Gr M Sd w/Tr Si & mica, 6U-115 ft F Sd w/ci Si & Tr mica, 115-121 ft Cl & Si w/layers of Wh F Sd, 121-130 ft Cl & Si & Tr of F Sd, 130-152 ft Sd w/layers of Si, 152-168 ft Cl & Si

0-5 ft ci Si, w/c to F Sd, 5-1+5 ft C to F Sd w/Tr Si ¿+5-99 ft Gr M-F Sd w/Cl Si & Tr of mica

0-3 ft Si w/Tr of Sd, 3-77 ft Si Sd w/Tr of G

Driving Data

DepthGWLft

Test Pile Wo. Type of Pile

Length of Pile, ft

Em-Driven bedded

Type of Hammer

Energyft-lb

BlowsperLastft

MaxTestLoadtons

Fail­ureLoadtons

- 5 Pipe ll+ in. OD 69.O 6 5.O Vulcan i30C 2k, •I+50 63 100 -

- 15 Pipe ll+ in. OD 59-5 51+.5 28 100

2 Pipe ll+ in. OD 5 3 .0 5 1 .0 76 100 „ll+ BP 117

1 Pipe ll+ in. OD 8O.O 71+.0 62 100lb BP 117

-- 2 8 2 .0 8 I .O 55 "

38 Pier N-2 Mono butt ¥+.8 35-0 Vulcan 1 15,000 51 80Tip-8 in.

0-U ft Sd Cl & F Sd Si, ¿+-30 ft -- 25 Pier W-6F to M Sd w/Tr of Si, 30-53 ft Si Sd

¿+5 .0 3 1 .0 57 80

0-5 ft Sd Si, 5-10 ft Sd, 10-11+ ft Sd Si, 11+-31 ft F to C Sd w/G, 31-71 ft Si Sd, 71-80 ft layers of Si Cl & F Gr Sd

0-5 ft Sd Si, 5-10 ft Sd, 10-11+ ft Sd Si, 11+-31 ft F to C Sd w/G, 31-71 ft Si Sd, 71-80 ft layers of Si Cl & F Gr Sd

0-5 ft Sd Si, 5-10 ft Sd, 10-14 ft Sd Si, 11+-31 ft F to C Sd w/G, 31-71 ft Si Sd, 71-80 ft layers of Si Cl & F Gr Sd

0-1+ ft Sd & G, 1+-8 ft Si w/layers of Sd, 8-1+0 ft F Sd w/G, 1+0-87 "t F Sd w/Si, Cl. & Shell 87-91 ft Si & F Sd w/pieces of sandstone, 91-101 ft- Cl w/layers of Sd

16 Pier W-ll+

16 Pier N-22

1+ Pier N - 3 0

1+1 Pier S-2

37-0

31+.0

39.0

31.0

52 80

56 80

1+1+ 90

62 80

0-3 ft Si Cl, 3-9 ft M-C Sd & G — 12 Pier S-109-52 ft F-C Sd w/some G

0-1+ ft F Si, 1+-26 ft F-C Sd & G — 5 Pier S-2626^29 St Cl 29-53 ft F-C Sd

33.0

26.O

55 80

33 80

Remarks

Tapered section

(Continued) (10 of 2l+ sheets)

Table A2 (Continued)

Dis­trictor

Divi-sion

WAP

NCS

WCS

ÏÏPA

Driving DataLength of Blows Max Fail­

Date Depth Pile,, ft per Test ureof GWL Test Pile Em­ Type of Energy Last Load Load

Project Tests Generalized Soil Conditions ft Wo. Type of Pile Driven bedded Hammer ft-lb ft tons tons Remarks

Reedy Pt. Bridge 1966 0-4 ft F Si, 4-26 ft F-C Sd & G, 20 Pier S-18 Mono, Tip-& in. 45.0 32.O Vulcan 1 15,000 52 80 Tapered section26-29 ft St Cl, 29-53 ft F-C Sd I

1 ““ 5 Pier S-34 Mono, Tip-8 in. 48.0 28.0 15,000 60 80 — Tapered section

Summit Br 1957 0-3 ft water, 3-6 ft F Sd, 6-17 ft 0.0 2 Pier 4 14 BP 102 I79.O 104.61

McK-T S8 26,000 67 180 __

Grand Av & Stockyard W. Pumping Stations, South St. Paul, Minn.

Grand Av & Stockyard W. Pumping Stations, South St. Paul, Minn.

Facilities, Alaska

FacilitiesAlaska

31-38 ft Cl, 38-4l ft F Sd, ill- 55 ft Cl & Sd, 55-61 ft F Sd, 61-lOil ft Sd Cl, 104-122 ft F Sd, 122-134 ft Sd Cl

0-11 ft Cl, 11-48 ft F Sd, 48-52ft Sd, 52-68 ft Cl, 68-100 ftF Sd Cl

0-6 ft Cl & G, 6-28 ft Sd Cl,28-34 ft F Sd, 34-42 ft Cl Si, 42-45 ft Cl & Sd shale, 45-71 ft F-C Sd, 71-74 ft Cl w/some Sd, 74-88 ft ci, 88-104 ft F Sd

0-4 ft Sd w/some Cl, 4-21 ft Cl, 21-30 ft F Sd & Tr of Cl, 30-49 ft Sd Cl, 49-54 ft Sd & G, 54-63 ft Cl, 63-69 ft F Sd, 69-99 ft Sd Cl

0-4 ft Sd w/some Cl, 4-21 ft Cl, 21-30 ft F Sd w/Cr of Cl 30-49 ft Sd Cl, 49-54 ft Sd 8= G, 54-63 ft Cl, 63-69 ft F Sd, 69-99 ft Sd Cl

6.0 11 Pier 5 12 BP 74 79.0

19.O 7 Pier 7

19.O 5A Pier 7

56.O

42.75

24 120

6O.O 48.6 McK-T DE 30

1967

1967

Stockyard N Wd, butt-II-1/2 17.O I7.Oin. Tip-9 in.

Wd, butt-16 in. 23.O 23.OTip-11 in.

55

335

44

44

f Public 1964 Sd Si 0.0 2 Wd, butt-13 in. 57.0 45.0 Vulcan drop 35,000 90 „ __Cordova, Tip-9 in. max

— 1 Wd, butt-l4 in. 57.0 40.0 \ 201 _ _Tip-11 in. f

f Alaska RR 1954 -- — A l4 BP 73 60.3 25.0 Warrington-* 15,000 15 40 40Seward, Vulcan 1

-- -- B l4 BP 73 62.3 35.0 30 66, 66-- -- C 62.O 25.0 58 -- __-- — D 85.O 55.0 33 80 80-- — E 79-3 55.0 89 100 — Lagged with 2-12 in. x 12 in.

timber from 17-32 ft-- -- F 55.0 31.6 68 100 — Lagged with 2-12 in. x 12 in.

for 15 ft-- -- G 44.0 25.0 20 60. -- Lagged with 2-12 in x 12 in.

for 15 ft

(Continued) (ll of 24 sheets)

Dis­trictor Date DepthDivi­ of GWLsion Project Tests Generalized Soil Conditions ft

NPA Alaska RR Dock Seward 1961+1965

19- 2917- 2218- 2620- 6 16-28 14-28 19-20

18-21

NPS Aircraft Control & Warning 1955 Bldg. Portland

Aircraft Control & Warning 1955 Bldg. Portland

Aldercreek, Underpass GNRY 1966 Relocation Libby Project,Montana

Libby Project GNRR Relocation 1966 Wolf Creek Spur

0-4 ft Si, 4-12 ft Cl, 12-14.5 3.0 1ft Si Sd, 14.5-21.5 ft F Sd,21.5- 24.5 ft Sd & Si, 24.5-30 ft Sd Si, 30-36.5 ft Sd & Si,36.5- 50 ft Sd

0-4 ft Si, 4-12 ft Cl, 12-14.5 3.0 2ft Si Sd, 14.5-21.5 ft F Sd,21.5- 24.5 ft Sd & Si, 24.5-30 ft Sd Si, 30-36.5 ft Sd & Si,36.5- 50 ft Sd

0-13 ft Sd G w/Si, numerous cobbles and boulders, 13-19-5 ft Sd G,19.5- 23 ft Cl G, 23-25 ft F Si Sd,25-28 ft Sd G, 28-33 ft Si Sd, 33- 43 ft G Sd, 43-54 ft Sd G, 54-62.5 ft G Si Sd, 62.5-69 ft F Sd, 69- 72. 5 ft G Si Sd, 72.5-75 ft Sd G,75-76.5 ft F Sd, 76.5-79.5 ft GSi Sd, 79.5-88.5 ft Sd G

0-3 ft Sd Si, 3-8.5 ft Sd G w/ cobbles & boulders, 8.5-45 ft Si Cl, 45-54 ft CL Si, 54-67 ft Si Cl, 67-69.5 ft Si Sd,69.5- 76 ft Si Cl, 76-88 ft Sd Cl Si, 88-97 ft Sd Si w/Tr Cl 97-100 ft lean Cl

ORP Shenango 1962’ 0-128 ft Si & Sd, 128- ? ft glacial — 6-1 Pier 2till

0-128 ft Si & Sd, 128- ? ft glacial — 6-2till

7-1 abut 1

0-7 ft M Sd & G, 7-11 ft Sd w/G,11-13 ft Cl, Si, w/G, 13-20 ft Si Cl, 20-30 ft Sd G, 30-33 ft Si Cl, 33-40 ft Cl w/Sd part­ings & lenses, 40-42 ft M Sd w/G,42-49 ft M Sd w/some F G, 49-52 ft Si Sd w/ F G

7-2 abut 1

Table A2 (Continued)

Driving Data

Type of Pile

Length of Pile, ft

Em-Driven bedded

Type of Hammer

Energyft-lb

BlowsperLastft

MaxTestLoadtons

Fail­ure

Loadtons Remarks

14 BP 117 38.7 Linkbelt D520 30,000 90 130 94 20 ft lagging14 BP 117 — 61.7 I I 88 130 — 20 ft lagging

— - - 46.4 i i 124 130 ll4— — 33.0 T T 70 130 70— — 42.5 — — 158 124 —

— — 40.0 — — 210 130 '108— — 35.0 Vulcan 06 & 19,500 15 130 — 35 ft lagging

Linkbelt D520 and30,000

- - - - 25.O Vulcan 06 & 19,500 72 82 - - 20 ft laggingLinkbelt D520 and

30,000

Wd, butt-I6-I/2 75.5 58.0 Raymond 65C 19,500 4o 4o _ _

in. Tip-8 in.

Wd, butt-17 in. 75.5 66.0 Raymond 65c 19,500 38 4o . .

Tip-9 in.

12 BP 53 42.0 42.0 Delmay D12 22,500 110 100

Wd 70.0 6O.O 1 40 40

Pipe, concrete filled, l4 in, OD

22.0 Vulcan 80C

20.5

46.0

154 120 90 Wall thickness = 1/4 in.

148 I6O. 98

56 100 6570

Test da,ta questionable

50 l40- 110

(Continued) (12 of 24 sheets)

Table A2 (Continued)

Driving Datatrict Length of Blows Max Fail­or Date Depth Pile, ft per Test ureDivi- of GWL Test Pile Em­ Type of Energy Last Load Loadsion Project Tests Generalized Soil Conditions ft No. Type of Pile Driven bedded Hammer ft-lb ft tons tons

POR 660 EM Barracks, Camp I960 1 CCC 3OO mm diam 3 2 .8 2 8 .0 2 metric ton 26,400 51 64Chitose III, Hokkaido,Japan

__ _ 2 3 2 .0 37 71— — 3 3 2 .0 17 69 —— — 4 3 1 .0 24 6 1 —

Vehicle Storage Shed, Camp I960 — — l CCC 250 mm diam 2 6 .2 2 6 .0 1 . 5 metric ton 19 ,8 6 0 2 1 48 —Chitose III, Hokkaido, Japan

Vehicle Storage Shed, Camp I960 — — 2 2 6 .2 2 3 .0 37 48 _Chitose III, Hokkaido, Japan

Headquarters Bldg., Camp I96I — — l CCC 200 mm diam 3 2 .8 2 9 .0 25 32 _Chitose III, Hokkaido, Japan

Headquarters Bldg., Camp I96I — — 2 CCC 200 mm diam 3 2 .8 2 7 .0 16 33 —Chitose III, Hokkaido, Japan

BOQ & Officers Open Mess I96I -- — l CCC 250 mm diam 26.2 24.0 26 43 —Camp Chitose III, Hokkaido, Japan

BOQ & Officers Open Mess I96I — — 2 cc c 250 mm diam 26.2 2 5 .0 23 46 —Camp Chitose III, Hokkaido, Japan

AFSS Operations Bldg. 1964 0-5 ft 0 Si, 5-22 ft Cl, 22-29 ft 6.0 1 CCC 300 mm diam __ 3 1 .0 2300 Kg drop 33,200 20 80 __Misawa, Honshu, Japan Sd, 29-41 ft Cl, 41-66 ft Sd hammer

AFSS Operations Bldg. 1964 1 6.0 2 ccc 300 mm diam — 48.0 2300 Kg drop 49,800 58 80 —Misawa, Honshu, Japan T

0 -1 .5 ft 0 Si, 1 .5 -2 2 ft Si, 22-24hammer

AFSS Antenna Array, Misawa, 6.0 C-2 BP — 2 7.0 Diesel IDH12 22,;300 II8 100 —Honshu, Japan

1

ft Blue Clay, 24-? ft, dense Sd GJ 6.0 C-26 C „ 36.0 29 100| T 6.0 AB90 BP — 60.0 128 100 —

Autodin Facility, North 1967 0-2 ft Fat 0 Cl, 2-22 ft highly 7 .5 3 CCC 3OO mm diam 32.5 28.0 36 64 100*Camp Drake, Japan plastic Si, 22-24 ft Cl, 24-?

ft dense Sd G

400-man EM Barracks, North I960 0 -2 ft 0 topsoil, 2 -2 5 ft highly __ 2 CCC 35O mm diam 43.0 37.0 M-22B Diesel 39,300 100 112 175*Camp Drake, Japan plastic Si, 25-35 ft So Gr Si,

35-42 ft dense Sd G w/M denseSi Sd strata

Receiver Bldg., Iwakani, 1965 -- — — C 3OO mm diam 2 6 .2 2 5 .0 2200 lb drop 11,000 24 26 --MCAS hammer

AIC Equipment Bldg. Totsaka 1965 0-13 ft Si, 13-30 ft Si Sd, 30-51 -- — C 400 mm diam 32.8- -- 4400 lb drop 43,300 27 24 —Naval Radio Sta. ft Si, 51-54 ft Si G hammer

Autovon Microwave Facilities 1966 0-3 ft loose Si Sd w/G, 3-12 ft 7.9 C 300 min diam I9 .9 15 .0 4400 lb port­ 28 ,600 10 60 __Fuchu, Japan Sd Si, 12-20 very dense Sd G able pile

hammer (cartype)

POH Air National Guard Hangar I960 0-2 ft Si Sd w/coral fragments 2 .5 1 RST, tip-9.5 - 1 8 - Vulcan 1 15,000 48 60 -Hickam AFB, Hawaii 2-3 ft So Sd Si CL, 3-4 ft

dense coral crust, 4-15 ft loose Si Sd w/coral fragments

Remarks

Test results questionable

(Continued)

Extropolation of load deformation curve. (13 of 24 sheets)

Table A2 (Continued)

Dis­trictDivi­sion

DepthGWL Test Pile

Length of Pile, ft

Driving Data

Air National Guard Hangar Hickam AFB, Hawaii

2nd Entrance Channel, Honolulu Harbor, Hawaii

Tests Generalized Soil Conditions

i960 0-2 ft Si Sd w/coral fragments2-3 ft So Sd Si Cl, 3-4 ft dense coral crust, 4-15 ft loose Si Sd w/coral fragments

0-8 ft loose to M dense Si Sd w/ coral fragments, 8-11 ft Si Cl Sd w/coral fragments, 11-22 ft Si Sd G w/coral fragments,22-37 ft coral ledge w/Sd pockets

i960 0-30 ft water, 30-32 ft So Si Sd,-32-38 ft very dense coral, Si & Sd, 38-75 ft Si Sd w/coral fragments

2.5 JK-1 RST, tip-9.53.0 in.

3.1 G-9 RST, tip-9.5

Blows Max Fail­per Test ure

Energy Last Load Loadft-lb ft tons tons Remarks

15,000 108 60 -

15,000 48 60

0.0 4 PC, 16 in. Oct. -- 62.O McK-T S8 26,000 72 150Tip-l6 in.

Berthing Pier, Tengan 1964 +19-O ft water, 0-39 ft Si Sd, +I9.O* 63 Bent 10 CO > C 30 in. diam 97.8 55.O Diesel IDH-22 — 68 110 —39-41 ft Cl, 41-47 ft Si Sd,47-48 ft Cl, 48-51 ft Si Sd,51-52 ft ci, 52-60 ft60- ? ft limestone

Si 0 ,

Berthing Pier, Tengan 1964 +I8-O ft water, 0-40 ft Si Sd, +I8.O* 58 Bent 28-D C 30 in. diam 97.8 56.O Diesel IDH-22 -- 64 1104o-43 ft ci, 43-44 ft Si Sd,44-55 ft ci, 55-62 ft Si Sd

Hamby Airfield

STARCOM Facilities, Taiwan

1963 0-2 ft ci, 2-7 ft poorly gradedSd, 7-15 ft fat Cl, 15-35 ft poorly graded Sd, 35-51 ft Si, 51-60 ft fat Cl, 60- ? ft limestone

0-4 ft Cl G, 4-25 ft fat Cl,25-30 ft Si Sd, 30-35 ft Cl Sd, 35-43 ft Si, 43- ? ft decom­posed limestone & limestone

1962 0-5 ft St Sd Si, 5-9 ft Si Sd,9-29 ft Si Sd, 29-34 ft ande­site w/Gr Si Sd, 34-46 ft Sd Si w/andesite fragments, 46-76 ft plastic Si w/weathered andesite

0-5 ft So Si w/G, 5-22 ft Si w/ weathered andesite, 22-73 ft dense C Sd w/weathered andesite

0-5 ft So plastic Si, 5-30 ft Si Sd w/lenses of Sd Si, some weathered andesite, 30-48 ft St Sd Si w/weathered andesite & G, 48-65 ft dense Sd G w/weathered andesite

6.0 20 Footing C 14 in. sq 57.0 6O.O Delmay D-12 22,500 ll4 80A-2

4.0 70 Footing C 14 in. sq 57-0 53-0 Delmay D-12 22,500 80 80E-8

62.0 B-l C 14 in. sq

51.0 Between C 12 in. sqB-9 & B-10

55.O Between A-12 C 12 in. sq & A-13

37-2 Kikuchi (Single 21,300 l44 70acting steam)

24.5 133 4o

84 4o

(Continued)

* Pile driven in sea bottom. (l4 of 24 sheets)

Table A2 (Continued)

Dis­trictorDivi­sionPOO

SAM

SAS

Project

STARCOM Facilities.Taiwan

Dateof

Tests

1962

Generalized Soil Conditions

)-5 ft Cl Si w/Sd, 5-9 ft Cl Si w/G & weathered andesite, 9-12 ft Si w/weathered andesite G, 12-16 ft Cl Si, 16-22 ft C Sd Si w/weathered andesite, 22-46 ft Si G w/weathered andesite frag­ments, 46-4-7 ft andesite, 47-52 ft Sd Si w/F G

DepthGWL Test Pile ft No.

Between B-l & B-2

Length of Pile, ft

Type of Pile Driven bedded

Driving Data

C 12 in. sq

Type of Hammer

Energy ft-lb

Blows Max Fail- per Test ure

Last Load Loadft tons tons

Kikuchi (Single 21,300 84acting steam)

40.0

NASA Mississippi Test 1962- O-38 ft Cl, 38-59 ft Sd w/G 10 S-l 14BP73 __ 70.0 Fairchild 20 20,400 30 300 _Facility SII Test Stand 1963 59-66 ft G

12 S-2 - 64.0 25 270 -

Ammo loading terminal 1957 0-24 ft Cl Sdx 24-45 ft Cl Sd T-3A C 20 in. oct. 43.0 36.0 McK-T S8 26,000 13 45 40 Spudding & jetting used for settingKing's Bay Wharf 3 & w/shells, 45-48 ft Sd Si Cl Tip-8 in. & cleaning 17 ft of 30 in. diamApproaches „ , ___ 1 shell

fragments, 52-55 ft limestone,So, porous loosely cemented shells, 55-56 ft Sd w/limestone fragments, 56-64 ft Cl w/0

0-24 ft Cl Sd, 24-45 ft SI Sd w/shells, 45-48 ft Sd Si Cl,48-52 ft Sd w/limestone & shell fragments, 52-55 ft limestone,So, porous loosely cemented shells, 55-56 ft Sd w/limestone fragments, 56-64 ft Cl w/0

0-13 ft Cl Sd w/shells, 13-15 ft Sd Cl w/shell fragments, 15-24 ft Cl Sd w/shells, 24-52 ft Sd w/ shells, 52-54 ft limestone, So, porous loosely cemented shells, 54-55 ft Sd w/shells, 55-58 ft limestone, So, porous loosely ce­mented shells

0-8 ft Sd, 8-9 ft Cl w/0, 9-25 ft limestone, So, porous loosely ce­mented shells, 25-26 ft Cl w/o, 26-30 ft Cl Sd, 30-36 ft lime­stone, hard, dense, 36-42 ft limestone, hard, porous, sili­ceous

0-9 ft limestone, So, porous,loosely cemented shells, 9- 10 ft Cl w/0, 10-14 ft Cl Sd, 14-19 ft limestone, So, argillaceous,19-26 ft limestone, hard, porous, siliceous

0-l4 ft Cl, 14-15 ft limestone, 15-20 ft Cl

63.0 55.0

43.0 37.0

51.0 4l.0

51.0 4 3 .O

64.0 39.0

41.0 11.0

58.0

47.0

435

47 80

44 100

150 100

84 100

35 100

Spudding & jetting used for setting & cleaning 36 ft of 30 in. diam casing

Spudding & jetting used for setting & cleaning 17 ft of 30 in. diam casing

Spudding & jetting used for cleaning & setting 17 ft of 30 in. diam casing

Spudding & jetting used for cleaning & setting 24 ft of 30 in. diam casing

Spudding & jetting used for cleaning & setting 36 ft of 30 in. diam casing

Spudding used

Spudding & jetting used to depth of 46 ft before driving

Spudding & jetting used to depth of 39 ft before driving

(Continued) (15 of 24 sheets)

Dis­trictor Date

Divi- ofsion Project Tests

SPL Seventh Ave. Bridge, San Jose Creek Channel

1966

UPKR Bridge, San Jose Creek Channel

Turnbull Canyon Road Bridge, San Jose Creek Channel

SPRR Bridge, No. 2, San Jose Creek Channel

SPRR Bridge No. 1, San Jose Creek Channel

Lemont Ave. Bridge, Coyote Creek Pier 3

Valley View Ave. Bridge,Coyote Creek Channel, upstream from North Fork Channel ' 1

Merrill Ave. Bridge, San 1961Antonio & Chino Creeks Improvement, Chino Creek Channel

Los Serranos Bridge, San i960Antonio & Chino Creeks Improvement, Chino Creek Channel

DepthGWL

Generalized Soil Conditions ft

0-3 ft Sd Cl, 3-6 ft M dense gi Sd, l4.4 6-12 ft St Sd Cl, 12-15 ft M St fat Cl, 15-27 ft'Cl.,some carbonaceous wood, 27-33 ft St Sd Cl, 33-36 ft St Sd Si, 36-39 ft M dense Si Sd

0-6 ft M dense Si Sd, 6-21 ft M St 29.0 Sd Cl, 21-24 ft M dense Si Sd,24-33 ft Si Sd w/G, 33-39 ft St Sd Cl, 39-42 ft M dense Si Sd,42-48 ft Sd Si, 48-51 ft Si Sd,51-57 ft Sd Si, 57-61 ft very dense Si Sd

0-9 ft St Sd Cl, 9-12 ft F Sd Si, 19.8 12-18 ft Si Sd, 18-24 ft F Sd Si, 24-35 ft Sd Cl w/Tr of 0

0-6 ft M dense Si Sd w/G, 6-27 ft St 5.0 Sd Cl, 27-33 ft fat Sd Cl, 33-42 ft very St Sd Cl, 42-45 ft dense Si Sd, 45-58 ft hard Sd Cl, 48-51 ft Si Sd w/G, 51-57 ft dense Cl Sd, 57-60 ft dense Cl Sd w/G

0-3 ft M dense Si Sd w/G, 3-6 ft 34.0 M dense Cl Sd w/G, 6-9 ft St Sd Cl, 9-12 ft Sd G, w/cobbles,12-18 ft G Si Sd w/G, 18-30 ft M dense Si Sd, 30-33 ft Sd Si,33-36‘ft Si Sd w/G, 36-39 ft dense Si Sd w/G, 39-42 ft Sd Si,42-45 ft dense Si Sd, 45-51 ft Sd Si w/Sd lenses, 51-54 ft M dense Si Sd, 54-69 ft St Sd Cl

0-6 ft M dense Si Sd, 6-9 ft Sd 24.0Si, 9-15 ft M St Sd Cl, 15-19 ft M dense Si Sd, 19-31 ft St Sd Cl

0-9 ft Sd Si, 9-13 ft Si, 13-17 ft 36.5Sd Si, 17-25 ft Si

0-49 ft Cl,, *+9-55 ft Si Sd,, 55-59 7.0ft Si

0-15 ft St Sd ci, 15--19 ft Si Sd, 10.019-22 ft Si, 22--26 ft Si Sd,26-41 ft Cl

Test Pile No.

Table A2 (Continued)

Driving DataLength of Pile, ft

Em- Type ofType of Pile Driven bedded Hammer

12 BP 53 — 27.0 Raymond 65C

RST, butt-l4 -50.0 37.0in. Tip-10 in.-

12 BP 53 — 38.O

RST, Tip- — 26.010 in.

RST , butt-l4 — 46.0in. Tip-10 in.

RST, butt-l4 -- 32.O Raymond 1in. Tip-10 in.

RST, type Z — 36.5 Vulcan 1butt-l4 in.Tip-10 in.

RST, CIP 12 in. — 39.O Vulcan 1diam

RST, l6 in. — 39*0 Raymond Inner-diam core 15M

Blows Max Fail­per Test ure

Energy Last Load Loadft-lb ft tons tons Remarks19,500 38 90 __ Lugs ten ft above tip

80 150

43 90

80 150

60 150

15,000 55 90

15.000 100 90

15.000 88 126 106

15,060 6l 265 154

(Continued) (l6 of 24 sheets)

T ab le A2 (C on tin ued)

D is ­t r i c t

orD iv i-s io n P r o je c t

D ateo f

T e s t s G e n e ra liz e d S o i l C o n d itio n s

DepthGWLf t

SPL G arey Ave B r id g e , WorthSan A ntonio & Chino C reeks Im provem ent, Chino C reek Channel

1959 0 -4 f t f a t C l , 4 -39 f t C l , S i , 45-57 f t s i l t s t o n e &

39-45 f t sh a le

D r iv in g D ata

T e st P i le

Length o f P i l e , f t

Em- Type o f Energy

Blowsp e r

L a s t

MaxT e stLoad

F a i l ­u re

LoadWo. Type o f P i le D riven bedded Hammer f t - l b f t to n s to n s

1 RST CIP 12 in . diam

30.0 29.0 Vulcan 1 15,000 44 85

ooV P lu n g in g

G arey Ave B r id g e , Worth 1959Sarp A ntonio & Chino C reeks Im provem ent, Chino C reek Channel

0-13 f t C l , 13-14 f t C l S d , 14-17 f t S i Sd , 17-18 f t C l , 18-28 f t S i , 2 8 -37 f t C l , 3 7 - 4 l f t S i Sd 4 l-4 2 f t S i G

2 RST CIP 12 i n . 2 5 .0 2 5 .0 Vulcan 1 1 5 ,0 0 0 35 90 90 P lu n g in gdiam

Remarks

C e n t in e la B lv d . B r id g e , 1961C e n t in e la C reek Channel Improvement

P a c i f i c E l e c t r ic R ailw ay 1961B r id g e , C e n t in e la C reek Channel Improvement

O -I6 f t S i C l, 16-34 f t C Sd & G w /c o b b le s , 34-44 f t M S d ,¿4-4-48 f t C Sd w /pea G

O-7.5 f t C l S i , 7 . 5-IO f t Sd C l, IO -I8 f t Sd S i , 18-24 f t S i , 2 4 -29 f t Sd S i , 29-3 4 f t S i Sd w/G, 34-37 f t Sd S i , 37-39 f t S i S d , 3 9 -5 6 .5 f t S i Sd w/G, 56. 5- 6O f t S i Sd

1 1 .8 D3

T ra in in g F a c i l i t y 75-2 1959Launcher S i t e 4 , Cooke AFB, C a l i f .

S IV B T e st Complex 1963Sacram en to , C a l i f .

0 -4 0 f t p o o r ly g rad ed Sd , 40-60 f t p o o r ly graded Sd w/some S i ,6 0 - ? f t C l Sd

0-10 f t C l Sd G w /c o b b le s , 10-17 f t C l Sd G w /cfcb bles, 17-28 f t C l G, 2 8 -9 7 .3 f t no sam p le , 97. 3- 1 1 1 .3 f t san d sto n e h ig h ly cem ented, 1 1 1 . 3- 119-9 f t w eathered sa n d sto n e , 1 1 9 .9 - 1 2 4 .5 Sd

B e ta -1

0 -4 f t G w/Sd S i & C l , 4 -1 2 .5 f t 2 4 .0 B e ta -3 C l Sd G w /c o b b le s , 1 2 .5 - 4 1 .6 ft C l G, 4 1 .6 - 4 7 .0 f t S i & Sd ,4 7 .0 - 5 7 .0 f t C l S d , 57-60 f t F Sd w /c i l a y e r s , 60-71 f t Sd w / p ea G, 71-80 f t t a i l i n g s w /le n se o f C l, pea G, & S d , 8 0 -8 0 .6 f t C l S d , 8 0 .6 -9 0 f t no sam p le ,90-92 f t S i Sd w/G, 92-95 f t S t C l, 95-99-5 f t c o b b le s G w/fr-M Sd & C l, 9 9 .5 -1 0 5 f t C l , Sd , &G, 105-107 f t Sd C l , 107-116 f t Sd C l, 116-121 f t C l S d , 121-124 f t san d sto n e

Armed F o r c e s R eserv e C e n te r , S an ta Ana, C a l i f .

0 - 2 7 .4 f t C l Sd G, 2 7 .4 - 3 9 .5 f t C l 1 1 .0 C o n tro l Sd , 3 9 .5 -4 4 f t C l Sd w/G, 44-50 c e n te rf t C l Sd

0-8 f t C l, 8 -27 f t Sd C l, 27-30 f t 5 .3 S i C l, 30-33 f t C l S i , 33-43 f t Sd ci, 43-46 f t S i Sd , 46-49 f t S i S d , 49-55 f t Sd S i C l, 55-58 f t Sd S i , 58-60 f t C l

RST, b u tt- . 1 4 -1 /2 in . T ip - lO - l /2 in .

32.0 Vulcan 65C

RST, b u t t - 1 4 -3 /4 in . T ip - lO - l /2 i n .

4 5 .0 Vulcan 65C

14 BP 73 — 3 2 .0 Vulcan 1

10 BP 42 - - 9O.O V ulcan 8OC

10 BP 42 — I I 8 .O V ulcan 8OC

10 BP 42 - - IO3 .O Vulcan 8OC

Wd, b u t t - l 4 3 5 .0 35 .O V ulcan 1i n . T ip - 10 in .

19.200 101 140 140

1 9 .2 0 0 136 255- 255 Anchor s t r a p f a i l e d a t 255 to n s0 . 28' n e t b u tt se tt le m e n t

15,000 32 220 200

2 4 .4 5 0 108 160

2 4 .4 5 0 90 170 — F la n g e bu ck led

2 4 ,4 5 0 44 180 — F la n g e bu ck led

15,000 13 85 85

(C ontinued) (17 o f 24 s h e e t s )

T ab le A2 (C o n tin u e d )

D is ­t r i c t

o rD iv i-s io n

SPL

D r iv in g D ataL en g th o f Blows Max F a i l ­

D a te D epth P i l e , , f t p e r T e s t u reo f GWL T e s t P i l e Em­ Type o f E nergy L a s t Load Load

.P ro je c t T e s ts G e n e ra liz e d S o i l C o n d it io n s f t No. Type o f P i l e D riv e n bedded Hammer f t - ■lb f t to n s to n s

P a c i f i c E l e c t r i c R ailw ay 196b 0 -6 f t S i Sd , 6 -1 0 f t no sam ple , __ 1 r t a b u t ­ CIP 12 i n . __ 50.0 V ulcan 50C 15:,100 55 150B rid g e , C oyote C reek 1 0-20 f t Sd S i , 20-21+ f t S i , mentC hannel 2 ^ -2 7 f t Sd S i , 2 7-3 1 f t C l Sd,

3 1 -3 ^ f t no sam p le , 3 ^ -3 7 f t C l, 37-^5 f t S i , 1+5-55 f t Sd S i , 55-58 f t C l , 58 -71 f t S i Sd

P a c i f i c E l e c t r i c R ailw ay 19 6b 0-6 f t S i Sd, 6- I 8 f t C l , I 8- 3O f t __ 2 l e f t p i e r CIP 12 i n . __ 1+7.0 V ulcan 50c 15:,100 133 185 __B rid g e , C oyote C reek Sd S i , 30-33 f t S i Sd, 3 3 -36 f tC hannel S i , 36-39 f t Sd C l , 39-b2 f t S i ,

1+2-1*8 f t Sd S i , I+8-56 f t S i Sd, 56-60 f t f a t C l , 60-61+ f t S i ,61+-68 f t S i Sd

D el Amo B rid g e , C oyote 19 6b 0 -3 f t Sd C l, 3 -6 f t Sd S i , 6 -1 0 __ P ie r 3 CIP ll+ i n . diam __ 1+5.0 V ulcan 1 15:,000 80 90 __C reek C hannel f t S i , 10-13 f t Sd C l , 13 -16 f t

Sd S i , I 6 - I 9 f t S i Sd, 19 -3 1 f t C l , 31-1+3 f t Sd S i , 1+3-1+6 f t S i Sd, 1+6-1+9 f t S i , 1+9-52 f t Sd C l., 52- 5I+ f t S i , 5 ^ -6 0 f t Sd S i , 6O-69 f t S i Sd

H anson Dam 1938 __ __ 5aL 12 BP 53 - 3 6 .0 2 1 .0 McK-T 9B3 8,750 -250 100 __B rea Dam O u tle t Tower 191+0 - - — IT 12 BP 53 - 58.0 3 5 -0 McK-T 9B3 8 ,,750 -- 120

F o u n d a tio n

B rea Dam O u tle t Tower 191+0 __ __ 2T 12 BP 53 - 65.0 3 6 .0 McK-T 9B3 8 , 750 __ 120 __F o u n d a tio n

A tc h iso n Topeka & S a n ta 1950 0 -2 1 f t S i Sd, 2 1 -3 1 f t C l S i , 1 1 .0 2 P i e r H Wd, b u t t - l l+ 5 2 .5 3 8 .0 V ulcan 1 1 5 , 000 25 90 __Fe R ailw ay B rid g e 31-36 f t C l Sd, 36-1+1 f t C l S i , i n . T ip -8

I+I-6I f t S i Sd, 6I -65 f t C l S i in .

A tc h iso n Topeka & S a n ta 1950 0-22 f t no sam p le , 22-32 f t 5.0 1 P ie r C Wd, b u t t - l l+ 3 5 .0 ' 26.0 V ulcan 1 1 5 , 000 8 55Fe R ailw ay B rid g e S i , 3 2 -60 f t S i i n . T ip -8

i n .

S o u th e rn P a c i f i c R a i l ro a d 1951 0 -1 0 f t c le a n Sd, 10-12 f t S i Sd, Mono, T ip -8 i n . 2 5 .O 23.0 V ulcan 2 7,260 32 100 100C o a s t l in e B r id g e , T ujunga 12 -16 f t Sd C l I 6 - I 7 f t S i Sd,Wash Im provem ent 1 7 -2 1 f t C l S i , 21-29 f t Sd S i ,

29-35 f t C l S i

Union P a c i f i c R a i l ro a d 1951 0 -1 7 f t S i Sd, 17-2 7 f t S i , 2 7 -2 9 13 .3 2 P ie r 2 CCC-16 OD i n . 30 .O 29.0 V ulcan 1 1 5 ,

000

66 100 __B rid g e , Los A nge les R iv e r Im provem ent

f t S i Sd, 29-1+7 f t S i 9 -1 0 i n . ID

C o n tr o l Tower & R ead in e ss 1952 0 -7 f t C l, 7 -1 2 .5 f t S i Sd, 7 .0 1 C o n tr o l RST, b u t t - 1 5 .5 1+0.0 1+0.0 1+7 98H angar, Oxnard A ir F o rce I 2 . 5 - I 7 f t C l, 1 7 -2 0 f t S i Sd, to w er i n . T ip -1 0 .5Base 2 0 -23 f t S i , 2 3 -3 3 f t C l, 33-1+2

f t S i , 1+2-1+3 f t C l, 1+3-1+8 f t S i ,i n .

1+8-50 f t C l

C o n tr o l Tower & R ead in e ss 1952 0- 1+ f t C l, 1+-10 f t S i , 10-15 f t 8 .0 2 R ead in e ss RST, b u t t -1 3 .5 28 .O 26.0 26 70 70H a n g ar, Oxnard A ir F o rce ci, 15 -19 f t s i , 19-26 ft Cl, h an g a r i n . T ip -1 0 .5Base 26-1+3 f t S i , 1+3-1+6 f t C l,

1+6-1+8 f t S i , 1+8-1+9 f t c i , 1+9-50 f t S i

i n .

Remarks

H y d ra u lic pump f a i l e d and t e s t was d i s c o n t in u e d

H y d ra u lic j a c k f a i l u r e a t 90 to n lo a d

Tw enty th r e e f t , deep t e s t p i t s w ere d r i l l e d b e f o r e p i l e s w ere d r iv e n an d o v e rb u rd e n was l a t e r rem oved

T ap e re d s e c t io n

Gauge s l ip p e d a t 98 to n s

(C o n tin u ed ) ( l 8 o f 2b s h e e ts )

Table A2 (Continued)

Dis­trictor

Divi-sion

SPL

Driving Data

Project

Dateof

Tests Generalized Soil Conditions

DepthGWL Test Pile ft No. Type of Pile

Length of Pile, ft

Em-Driven bedded

Type of Hammer

Energyft-lb

Blows per Last . ft

MaxTestLoadtons

Fail­ure

Loadtons Remarks

Elevated Steel Water Tank Ornard Air Force Base

1952 0-7 ft very So Cl, 7-20 ft So Cl, 20-29 ft Sd Si, 29-5O ft M Cl

6.0 C 16 in. 35.0 3^.0 Vulcan 1 15,000 25 62

Southern Pacific Railroad Bridge, Rio Hondo Channel

1953 0-2 ft Si Sd, 2-5 ft Si, 5-ll+ ft Si Sd, lU-1+0 ft Si

+6.0 CCC 16 in. diam 15.0 12.0 Vulcan 2 7,260 100 68 60

25O-Bed permanent-type 1956 0-8 ft Cl, 8-9 ft poorly graded Sd, 1 CIP 16 in. diam — 25.O Drilled — -- 100 —hospital, March Air Force 9-21+ ft Cl, 21+-25 ft Cl SdBase

250-Bed permanent-type 1956 0-20.5 ft Cl, 20.5-25 ft Cl Sd -- 2 CIP 16 in. diam -- 25.0 Drilled -- -- 100hospital, March Air Force Base

Pacific Electric Railway 1957Bridge, Los Angeles River Improvement, Santa Ana Branch to Pacific Ocean,20th St to 7th St

Pacific Electric Railway Bridge, Los Angeles River Improvement, Santa Ana Branch to Pacific Ocean below Dominguez to Carson Street

0-11 ft Si Sd, 11-20 ft Cl, 20-23 ft fat Cl, 23-32 ft Cl, 32-59 ft Si Sd, 59-71 ft Si

0-ll+ ft Si Sd, 114-20 ft poorly graded Sd, 20-22 ft Cl, 22-27 ft Si, 27-1+2 ft Cl, 1+2-1+1+ ft Si Sd, 1+1+-1+7 ft Cl, 1+7-55 ft Si Sd, 55-61 ft Cl, 61-6U ft Si, 6b- 71 ft Cl

0-1+ ft no sample, 1+-9 ft Cl, 9-13 ft no sample, 13-16 ft Cl,16-18 ft Si Sd, 18-23 ft no sample, 23-32 ft Si Sd, 32-3^ ft Si, 3^-39 ft Cl, 39-1+0 ft Si,1+0—1+2 ft Cl, 1+2-1+6 ft Si, 1+6—1+7 ft Cl, 1+7-50 ft Si Sd, 50-61 ft Si, 6l-66 ft well-graded Sd,66-69 ft Si

0-5 ft no sample, 5-1^ ft Cl,11+-29 ft Si Sd, 29-31 ft fat Cl, 31-36 ft Si Sd, 36-39 ft Cl, 39-I+O ft Si, 1+0-1+2 ft fat Cl, 1+2-1+1+ ft Si, 1+1+-1+7 ft Si Sd, 1+7-50 ft no sample, 50-51 ft Si Sd, 51-52 ft Si, 52-55 ft Si Sd, 55-59 ft Cl, 59-62 ft Si Sd, 62-70 ft Sd

0-5 ft Si, 5-13.5 ft Sd & Si Sd, 13.5-29 ft Si, 29-33 ft Si Sd, 33-35 ft Si, 35-^0 ft Si Sd,1+0-1+5 ft Si Sd, 1+5-50 ft Si Sd

1 east abut- 1I+ BP 73 ment

2 west abut- 1I+ BP 73 ment

>16.0 3 Pier 1+ 1I+ BP 73

>16.0 1+ Pier 1 1I+ BP 73

10.0 Pier 2 RST, butt-I2-I/I+ in. Tip- 8-I/2 in,

1+2.0 Vulcan 1 15,000 1+7 150

1+6.0 Vulcan 1 15,000 38 75 — Pile was not vertical and jacksslipped

33.0 Vulcan 1 15,000 1+9 11+1+ 121+

1+5.0 Vulcan 1 15,000 IO5 I60

1+0.0 Vulcan 65C 19,200 85 I50 I50 Jetting used

(Continued) (19 of 2I+ sheets)

Table A2 (Continued)

Driving Datatrict Length of Blows Max Fail­or Date Depth Pile, ft per Test ureDivi- of GWL Test Pile Em- Type of Energy Last Load Loadsion Project Tests Generalized Soil Conditions ft No. Type of Pile Driven bedded Hammer ft-lb ft tons tons

SPL Pacific Electric Railway 1957 0-6 ft Si, 6-9 ft Si Sd, 9-12 ft 25.O West abut- DST, butt-l4 50.O Vulcan 65C 19,200 110 170 170Bridge, Los Angeles River Improvement, Santa Ana Branch to Pacific Ocean below Dominguez to Carson St

Base Maintenance Hangar, George Air Force Base Victorville, Calif.

Si, 22-21 ft Si Sd, 21-21+ ft Sd or Si Sd, 2I+-3I+ ft Si Sd, 34-40 ft Si, I+O-56 ft Si Sd

1955 0-29 ft well-graded Sd & Si Sd,29-31 ft Si, 31-39 ft well-graded Sd & Si Sd

’in. Tip-8-1/2 in.

1 Column Bid l4 BP 73 12.0 Diesel free 11,000- I+9piston opera- 18,000 tion

Jetting used

Probable failure at 200 tons

Base Maintenance Hangar 2 Edwards Air Force Base

Jet fuel-storage tank, Ornard Air Force Base

1956 O-I+.5 ft no sample, 4.5-10.5 ft Sd, -- 110.5- 1 1 ft Si, II-I2.5 ft Cl Sd,12.5- 14 ft Si Sd, 14-16.5 ft Cl Sd, 16.5-18.5 ft Si Sd, I8.5- 19.5 ft Sd, 19.5-21 ft Cl sd,21-22 ft Sd, 22-23.5 ft Cl Sd,23.5- 25 ft Si Sd

1956 0-2 ft Si Sd, 2-5.5 ft Cl, 5.5-7 10.0 1ft Si & Si Sd, 7-8.5 ft Si Sd,8.5- 13 ft Cl, 13-16 ft Si,I6-I9 ft Cl, 19-22 ft Si, 22-28 ft fat Cl, 28-31 ft Cl, 31-46 ft Si

CIP 16 in. — 22.0 Drilled -- -- l80 l80diam

CIP EST, butt- — 35.0 Raymond 1 15,000 l6 87 8715.3 in. Tip- 10.5 in.

Swimming Pool Ornard Air Force Base

1957 0-2 ft Si Sd, 2-II.5 ft Cl,11.5-22 ft fat Cl, 22-32 ft Si

1 CIP RST, butt- — 34.0 Raymond 1 15,000 l6 89 8915.'3 in.Tip-10.5 in.

Airmen Dormitories, George 1957 Air Force Base, Victor­ville, Calif.

1 CIP l8 in. diam — 20.0 Drilled l40 — Top of concrete pile failed incompression at l4o tons

Sepulveda Dam 1939 Br Cl loam at surface, gradingto Sd loam at 10 ft which con­tained small amounts of l/2 in. G at 20 ft. G lenses occurred at 50 and 85 ft

8.0 1 C2 12 BP 53 4o.o McK-T 9B3

_ 2 F2 .. 4o.o McK-T 9B38.0 3 A17 __ 56.0 McK-T 10B3r 4 A9 -- 4o.o 1i 5 A5 — 4o.o T! 6 F9 — 4o.o McK-T 10B3— 8 C2 " 4o.o McK-T 9B37.0 10 H5 RST, Butt-14.9 __ 35.0 Vulcan 1

in. Tip-10.6in.

12 H25 RST, Butt-l4.9 -- 35.0 Vulcan 1in. Tip-10.6 :in.

13 L5 Union concrete 4o.o 37.0 McK-T 9B3Butt-17.5 in.

Tip-8 in.

8,750 -35 78. 78

-35 80 Pool of water maintained abouttop of pile

13,100 ~39 119 --I —20 60 - -

-34 60 --1f -23 70 --

8,750 35 80 — Pool of water maintained abouttop of pile

15 5000 -71 l4o Pile settled rapidly at l40 tons

15,000 -92 125 -- Jack fully extended

8,750 -300 175 —

(Continued) (20 of 24 sheets)

Table A2 (Continued)

Dis­trictorDivi-sion Project

Dateof

Tests Generalized Soil ConditionsDepthGWLft

Test Pile No.

SPL Sepulveda Dam 1939 Br Cl loam at surface, grading to Sd loam at 10 ft which con­tained small amounts of l/2 in. G at 20 ft. G lenses occurred at 50 and 85 ft

7.0 16 D30

Driving DataLength of Blows Max Fail­Pile, ft I per Test ure

Em- Type of Energy Last Load LoadType of Pile Driven bedded Hammer ft-lb ft tons tons

C l8 in. oct. 35.0 McK-T 10B3 13,100 ~4oo 120 _top-l8 in. point-8 in.

SPN Rodeo Creek Flood ControlSWL Lock & Dam 1 Arkansas

River Project

Lock 2, Arkansas River Proj ect

Lock & Dam 3, Arkansas River Navigation Project

1965 0-8 ft Sd Cl, 8-6O ft So Cl,60-75 ft M Cl, 75-IOO ft St Cl

1964

1965 0-36 ft select fill material,36-56 ft hard fat Cl, 56-62 ft dense Sd

196519641964

1965 0-10 ft Sd, 10-11 ft fatCl, II-I5 ft Sd, 15- 22 ft Sd Cl, 22-34 ft Sd, 34-42 ft Sd w/G,42 -45 ft Cl & Sd,45-50 ft Sd, 50-56 ft Sd & G

0-15 ft Sd, 15-21 ft Sd & fat Cl, 21-55 ft Sd w/Cl lenses

0-15 ft Sd, 15-21 ft Sd & fat Cl, 21-55 ft Sd w/Cl lenses

0-3 ft Sd, 3-9 ft Sd w/fat Cl layers, 9-11 ft Sd & Si Sd, 11-23 ft Sd, 23-59 ft Sd &Si Sd, 59-66 ft Sd,66-69 ft Sd & Si Sd, 69-79 ft Sd, 79-84 ft Si & Cl, 84-88 ft Fat Cl

7.5 - C 12 j sq

21.0 c-8 PC 18 in. sq6.0 G-8 PC 20 in. sq7.5 G-2 PC 18 in. sq

-- B-5 12 BP 53

2.5 B-5 PC 18 in. sq

9.0 G-2 PC l4 in. sq7.0 5 .PC l4 in. sq6.0 J-2 Wd , Cl.A

5.2 G-8 l4 BP 73

7.5 E-ll first 14 BP 73test

9.9 E-ll secondtest

8.9 J-3 firsttest

8.3 J-3 second test

8.5 J-3 thirdtest

98.3 97.0 Vulcan 010 32,500 4 110 100

56. O' 50.0 Vulcan 200C 50,000 i4o 425. 445*

0 0

-C? 0 45.048.0

Vulcan 200C Vulcan 200C

50,00050,000

409135

490475

4oo352

Jetted to Jetted to

32 feet 40 feet

65.O 62.8 Delmay Dt-12 22,500 276 195 95** Driven in lateral

pilot holes to reduce movement

45.O 43.0 Vulcan 0l6 48,750 38 325 280

47.0 42.8 Vulcan l40C 36,000 67 300 296

47.0 41.7 Vulcan 140C 36,000 52 I60 129 Jetted to -36 ft40.0' 35.0 Vulcan 65C 19,200 70 120. 112 Jetted to 27 ft45.0 43.0 2-50 Foster

Vibrator77 0cpm

130 85 Driven in 90 seconds

45.O 42.8 Vulcan 140C 36,000 10 I80 134

65.O 61.8 Vulcan 140C 36,000 18 230 185 One foot removed and a 21 ft section was spliced to the first test pile

50.0 46.7 2-50 Foster Vibrator

770cpm

I80 105

5O.O 46.7 2-50 Foster Vibrator

770 210 cpm

120 Concrete vibrator was used to density the sands between the flanges of the preceeding pile

65.O 61.8 2-50 Foster Vibrator

200 145 Fifteen feet section was spliced vibrator was used again to 21 j

(Continued)* Extrapolation of load deformation curve,** Pile failed structurally at 95 tons. (21 of 24 sheets)

T ab le A2 (C o n tin u e d )

D is ­ D r iv in g Data.t r i c t L eng th o f

o r D ate D epth P i l e , f tD iv i­ o f GWL T e s t P i l e Em- Type o f Energys io n P r o je c t T e s ts G e n e ra liz e d S o i l C o n d it io n s f t . No. Type o f P i l e D riv e n bedded Hammer f t - l b

SWL Lock & Dam 3 , A rk an sas R iv e r N a v ig a tio n P r o j e c t

Lock & Dam A rk a n sa s R iv e r N a v ig a tio n P r o j e c t

Lock & Dam 6 , A rk an sas R iv e r P r o je c t

1965 0 -3 f t Sd, S i Sd & f a t C l l e n s e s ,3-30 f t Sd, 30-43 f t Sd & S i Sd, 4 3 -6 0 f t Sd

4 .9 R -19-55 l4 BP 73f i r s t o f s e r i e s

4.9 R-19-65second o f s e r i e s

4 .9 R -19-75t h i r d o f s e r i e s

— 6 .5 G-7 Wd, l 6 i n . X9- I /2 i n .

0-6 f t Sd, 6-9 f t C l, 9-52 f t Sd, 1.2 k-8 l 4 BP 895 2 -? f t t e r t i a r y C l

7 .7 K-9 14 BP 890-76 f t Sd & S i , 76-90 f t C l 2 .5 C-3 C l 6 i n . sq

0 -1 7 f t Sd, 1 7 -21 f t Sd & S i Sd 2 1 -3 7 f t Sd , 37 -40 f t Sd & S i ' 40 -4 7 f t Sd , 47-53 f t Sd & S i

’s d ,Sd,

5 .5 L-10 C l 6 i n . sq

5 3 -7 6 f t Sd, 76-90 f t C l

0 -22 f t Sd, 22-32 f t Sd & S i Sd. 32-53 f t Sd, 53-63 f t Sd & S i '’s d ,

6 .4 H-3 Wd, 16 i n . X 10 i n .

63-70 f t C l

0 -3 f t a l t e r n a t e l e n s e s o f S i & 3 - 5O f t Sd

Sd, 3 .6 B-2 Wd, 14 in . X 8 i n .

5 5 .0 5 2 .8

65.0 63.O

7 5 .0 7 3 .0

4 1 .0 39 .O

37.0 3 5 .3

37.0 3 5 .5

4 2 .0 39.6

4 2 .0 38.8

5 5 .0 5 0 .3

5 5 .0 4 7 .3

V ulcan 140C

V ulcan 140C

V ulcan 140C

McK-T C-5

McK-T S8

McK-T S8 V ulcan 016

V ulcan 016

V ulcan 65c

V u lcan 65c

36,000

36,000

36,000

16,000

26,000

26,000

4 8 .7 5 0

4 8 .7 5 0

19,200

19,200

1965

0 -1 7 f t Sd, 17-19 f t G & S i G, 19-38 f t Sd, 38-50 f t Sd S= S i Sd

0 .0 C-2 f i r s t PC l 8 i n . sqt e s t

4 .7 C-2 secondt e s t

4 .8 B-4 f i r s tt e s t

0 .0 B-4 sec o n dt e s t '

4 8 .0 4 4 .9 V u lcan 0 l6 4 8 ,7 5 0

4 4 .9

4 5 .4

4 5 .4

0 -6 f t S i Sd, 6 - l4 f t Sd & S i Sd, 6 .4 K-8 14-42 f t Sd, 42 -50 f t Sd & S i Sd, 5O-6I f t Sd, 6I -89 f t Sd & S i Sd

14 BP 73 4 2 .5 3 9 -3 W a rr in g to n - 3 2 ,5 0 0V ulcan 010

D avid D. T e r ry Lock & Dam 6 , A rkansan R iv e r P r o j e c t

1965 0 -1 3 f t Sd, 13-16 f t f a t C l, 1 6 -19 f t Sd & S i Sd, 19-38 f t Sd,3 8 -4 1 f t Sd & S i Sd, 41-47 f t Sd , 47 -57 f t Sd & S i Sd, 57-73 f t C l , 7 3 -77 f t Sd & S i Sd,7 7 -79 f t C l

1 .8 - B -l44 .3

14 BP 73 4 2 .5 3 9 -4 W a rrin g to n - V u lcan 010

32,500

Lock & Dam 4 , A rk an sas R iv e r & T r i b u t a r i e s , A rk an sas & Oklahoma

1963 Medium to f i n e san d s and s i l t y san d s

3 .0 1 P ip e , 1 2 .7 5 i n . OD (CE)

5 5 .0 5 3 .1 V ulcan l40C 3 6 ,0 0 0

Blowsp e r

La,stf t

Ma,xT e s tLoadto n s

F a i l ­u re

Loadto n s Remarks

13 200 150

17 244 175

19 250 215 T ip o f p i l e th o u g h t t o be in t e r t i a r y c l a y

42 120 97

12 I 80 129

12 I 80 15947 316 273.

44 272 216 J e t t e d t o 33 f t

50 120 100 J e t t e d t o 44 f t

92 120 n 4

102 5OO 389 J e t t e d t o 36 f t

102 491 431

100 578 506 J e t t e d t o 36 f t

100 6OO 504

9 300. 201.

20 310 255

16 173 i4 o W all th i c k n e s s = .3 3 i n >

(C o n tin u ed ) (22 o f 24 s h e e ts )

Table A2 (Continued)

Dis­trictor

Divi­s i o n Project

Dateof

Tests Generalized Soil Conditions

DepthGWL Test Pile ft No.

SWL Lock & Dam 1+, Arkansas River & Tributaries, Arkansas &

1963 Medium to fine sands

sands and silty 2.5 2 test 1

Oklahoma

Driving .DataLength of Blows Max Fail­Pile, ■ ft per Test ure

Em- Type of Energy Last Load LoadType of Pile Driven bedded Hammer ft-lb ft tons tons RemarksPipe, l6 in. OD 55.0 52.8 Vulcan ll+OC 36,000 38 250 195 Wall thickness = .312 in.

(CE)

2.5 3 Pipe, 20 in. OD 55.0 53.0 Vulcan ll+OC 36,000 1+1+ 260 215 Wall thickness = .375 in.(CE)

2.5 10 Pipe, 16 in. OD 55.0 53.1 Bodine -- — 227 l80. Wall thickness = .312 in.(CE)

2.5 16 Pipe, 16 in. OD 55.0 52.7 Vulcan ll+OC 36,000 2h 168 1U0.. Wall thickness = .312 in.(CE)

2.5 1+ C l6 in. sq i+5.0 1+0.2 Vulcan ll+OC 36,000 1+2 197 170

2.5 5 C l6 in. sq 55.0 51.0 Vulcan ll+OC 36,000 1+8 285 21+0

2.5 11 C l6 in. sq 55.0 38.8 Bodine -- 153 150

2.5 6 ik BP 73 U2.0 1+0.0 Vulcan 8OC 2l+,l+50 17 180 ll+O

2.5 7 Ik BP 73 55.0 52.1 Vulcan 8OC 2l+,l+50 31 220 190

2.5 9 lU BP 73 55.0 53.1 Bodine 25O 210

(Continued) (23 of 2k sheets)

Table A2 (Concluded)

Dis­trictor DateDivi­ ofsion Project Tests Generalized Soil Conditions

SWL Lock & Dam 1+, Arkansas River 1963 Medium to fine sands and silty& Tributaries, Arkansas & sandsOklahoma

IMK Jonesville Lock & Dam 1967 0-56 ft, F to M Si Sd w/Tr G

Driving DataLength of Blows Max Fail­

Depth Pile,, ft per Test ureGWL Test Pile Em­ Type of Energy Last Load Loadft No. Type of Pile Driven bedded Hammer ft-lb ft tons tons Remarks2.5 2 test 2 Pipe, l6 in. OD 55.0 52.8 Vulcan ll+OC 36,000 38 21+0 210 Wall thickness = .312 in.

(CE)

2.5 8 Wd butt 1 5 .2- in. tip 10.7-

1+0.0 38.6 Vulcan 65C 19,200 23 125 80

Supplemental Data

7 .O 1 PC l8 in. square 39-5 38.O Vulcan 016 1+8,750 66 1+00 356

2 1+6.5 I+5 .O 27 31+0 303

3 55-5 51+.0 1+1 381+ 3^72A 1+6.5 I+5 .O 19 21+1+ 196 Jetted 39 ft

(2l+ of 2k sheets)

Table AR

Tension Load Tests on Single Vertical Piles

trict Length of Blows Max Fail­or Date Depth Pile , ft per Test ureDivi- of GWL Test Pile Em­ Type of Energy Last Load Loadsion Project Tests Generalized Soil Conditions ft No. Type of Pile Driven bedded Hammer ft-lb ft tons tons

LMN Morganza Floodway Control 19*+9 O-26 ft Cl, 26-29 ft Sd Si, 29- 7.0 T-l Pipe, 2b in. OD (con- 91.2 6b.6 Vulcan OR 30,225 — 33*+ 136 963b ft Si Cl, 3 *+-53 ft Cl, 53- 69 ft Si Cl, 69-75 ft Sd, 75- 79 ft Si Cl, 79-85 ft Sd

crete filled)Pile set in 8.5 ft excavation

0-20 ft Cl, 20-26 ft Cl Si, 26- T-2 Mono FNI8, butt-18 in., 9b. 0 8O .2 Vulcan 1 15:,000 -3 2 5 128 9031 ft Sd Si, 31-37 ft Si Cl, tip-8 in.37-61 ft Cl, 61-72 ft Cl Si,72-78 ft Cl, 78-102 ft Sd

T-3 Mono N12, 12 in. diam 10b. 0 75.9 60 88 59

T-*4- Pipe, l8 in. OD (con­ 87.4 7*+-9 ~ I77 128 80crete filled)

T-5 Pipe, 2*4 in. OD (con­ 90.8 79-9 Vulcan OR 30,225 *+5 200 __crete filled)

O-3I ft Cl, 3¡L-líl ft Sd Si, *4-1- t -6 II9.8 93.3 Vulcan OR 30,225 -356 150 —65 ft Cl, 65-75 ft Si Cl, 75- 79 ft Cl, 79-96 ft Sd, 96- 100 ft Si Sd, 100-103 ft Sd

Pile set in 9 ff excavation

VA Hospital (Group 2)

Old River Control Structure

LMK Columbia Lock and Dam

19^7 0-8 ft fill, 8-22 ft 0 Cl, 22-39 ft plastic Cl, 39-67 ft Sd,67-8O ft plastic Cl, 8O-83 ft Si & Sd, 83-91 ft hard Cl, 91- 97 ft Si & Sd

1955 0-h0 ft Sd Si w/ci strata, hO- 9-*+b2 ft Cl w/Sa strata, *42-52 ft Sd Si, 52-8O ft F to M Sd w/tr G

1965 O-I8 ft Cl, I8-32 ft F to M 3.0Sd, 32-88 ft Cl (Tertiary)

D Wd, butt-15 in,tip-9.5 in.

,6 8I .0 Vulcan 1 15,000 3 b 150 Jetted to 65 ft

2 Pipe, 21 in. diam 66.6 65.1 Vulcan OR 30,225 ¿+0 200 135

3 l*4-H-73 beam (with bottom plate)

75.2 71.0 20 100 50

k Pipe, 17-in. diam 75.0 66.0 96 200 162

5 Pipe, 17-in. diam 50.0 *+5.0 3 8b 55

6 Pipe, 19-in. diam 75.1 65.0 65 18b 1*40

l lb BP 73 H-beam 68.0 63.0 Vulcan 1*40C 36,000 30 200 90

2 78.O 51.0 Vulcan 1*40C 36,000 6025 ibo No movement at 6025 blows

3 9O.O 81.6 Raymond0000

*+8,750 8OO 110

b Pipe, l8 in. diam 68.0 62.6 Vulcan 1*40C 36,000 50 160

5 Pipe, 18 in. diam 93.0 81.8 Vulcan 1*40C 36,000 551 l*+5 No movement at 551 blows

(Continued) (Sheet 1 of b)

Table A3 (Continued)

Dis-' ______ Driving Datat r i c t Length of Blows Max Fail­o r Date Depth Pile,, ft per Test ure

D i v i - of GWL Test Pile Em­ Type of Energy Last Load Loads i o n Proj ect Tests Generalized Soil Conditions ft No. Type of Pile Driven bedded Hammer ! ft-lb ' ft tons tons

LMK Jonesville Lock and Dam 1967 O-56 ft, F to M Si Sd w/tr G 7 1 PC l8 in. square 3 9 .5 38.0 Vulcan 016 1+8 ,7 5 0 66 I30 88

2 4 6 .5 1+5 .0 27 ll+O 115

3 55-5 5I+.O 1+1 130 112

2A 1+6.5 1+5 .0 19 80 69

10 4 1+8.9 4 5 .0 1+0 120 97

NED Fox Point Hurricane 1963 O-3O ft M Sd, 3O-5O ft Si Sd, _ __ 111 BP 89 __ 78.0 Bodine __ 90 52Barrier 5 0 -8 0 ft Si, 8D- Si Sd Sonic

POF 660 EM Barracks i960 - - 3 CCC 3OO mm diam 3 2 .8 32.0 - 26, 00

17 21 -

AFSS Operations Building I96I+ 0-5 ft 0 si, 5-23 ft ci, 23- 6 C-2 H-Pile __ 27.0 Diesel 22,000 1I+ 3529 ft Sd, 2 9 -4 0 ft Cl, 4o - IDH 1266 ft Sd

C-26 C 36.0 I 1 2 9 32

AB-90 BP - 61.0 1\ 1\ 1 3 b 1+0

SPL Sepulveda Dam 1939 Br Cl loam at surface, grading 0 C-2 12 BP 53 __ 1+0.0 McK-T 9B3 8 ,7 5 0 35 38to Sd loam at 10 ft whichcontains small amount of1/2 in. gravel at 20 ft. Gravel lenses occurred at 50and 85 ft

C) H-2 - 4o .o McK-T 37 38 —

£ A-lk -- 1+0.0 I 26 58 -A-2 1 - 1+0.0 1\ 31 58 --

A-7 - 1+0.0 McK-T 10B3 8,100 20 28 -

A-13 1+0.0 McK-T 9B3 8,750 32 28 -

A -3 -- 1+0.0 Mck-T 9B3 8,750 '38' 28

A-7 - 1+0.0 McK-T 10B3 8,100 20 28 —

F-7 - 1+0.0 McK-T 9B3 8,750 34 32 -F-13 ~ 1+0.0 120 32 -

A -3 -- 1+0.0 38 39 -

F-3 -- 1+0.0 39 39 -C C-l 1+0.0 52 48 -

C-3 - 1+0.0 McK-T 10B3 8,100 2I+ 48 -H-4 RST butt diam l b . 9 in. _ 31+.0 Vulcan 1 80 66

tip diam 10.6 in.

H-6 - 3 5 -0 75 66 -f -4 - 3 b . 0 86 — —f-6 __ 3 5 .0 48 _ _

Remarks

Pile jetted to within 6 ft of finished grade

(Continued) (Sheet 2 of 4 )

Dis­trictor Date DepthDivi­ of GWL Test Pilesion Proj ect Tests Generalized Soil Conditions ft No.SPL Sepulveda Dam 1939 Br Cl loam at surface, grading 8

to Sd loam at 10 ft which con­tains small amount of l/2 in. gravel at 20 ft. Gravel lenses occurred at 50 and 85 ft

L-25

F-25

L-1+L-6

0 -1+

0 -6

E-30

B-30

8" 8

0-10 ft Sd, 10-H ft fat Cl, 11- 5.015 ft Sd, 15-22 ft Sd Cl, 22- 3b ft Sd w/c, 1+2-1+5 ft Cl and Sd, 1+5-50 ft Sd, 50-56 ft Sd and. G

0-15 ft Sd, 15-21 ft Sd and fat 7.5 Cl, 21-55 ft Sd w/ci lenses

0-3 ft Sd, 3-9 ft Sd w/fat Cl 9.2layers, 9-11 ft Sd and Si Sd,11-23 ft Sd, 23-59 ft Sd and Si Sd, 59-66 ft Sd, 66-69 ft Sd and Si Sd, 69-79 ft Sd, 79-81+ ft Si and Cl, 81+-88 ft fat Cl

0-3 ft Sd Si and fat Cl lenses, 1+-9 3-30 ft Sd, 30-1+3 ft Sd and Sisd, 1+3-60 ft Sd

5-1

* t 7.6Lock and Dam No. 1, 196k — 21Arkansas River Naviga-

'' tion ProjectExtrapolated from Gross Pile Head Rise Curve.

SWL Lock and Dam No. 3, 1965Arkansas River Naviga­tion Project

B-29

E-29

A-2

F-2

E-ll

J-3

R-19-55First of Series

R-I9-65 Second of Series

G-7C-8

Table A3 (Continued)

_______Driving Data_______Length of Blows Max Fail-Pile, ft per Test ure

Em- Type of Energy Last Load LoadType of Pile______ Driven bedded Hammer ft-lb ft tons tons

12 BP 53

12 BP 53 Union 7J8x 1+0

18 in. C oct tip diam 8 in.

12 BP 53

12 BP 53 ll+ BP 73

1+0.0 Vulcan 1 8,:100 19 32

_ 1+0.0 McK-T-10B3 16 . 55

- 38.0 1+00 82 -

- 35.0 109 85 -200 - -

McK-T-9B3 8,750 21+0 -- -McK-T-10B3 8,100 1+00 58

- 21.1+ McK-T-10B3 8,100 - 58 -0 Vulcan 2l+,370

- 35 .0 McK-T-10B3 8,100 3OO 32 —

35 .0 McK-T-10B3 8,100 150 32 -1+0.0 McK-T-9B3 8,750 31 36 -

- 1+0.0 McK-T-9B3 8,750 35 36 —1+5.0 b2.3 2-50 Foster

Vibrator58 25

Wd, l6 x 9-I/2 in.PC 18 in. sq.

(Continued)______

1+5.0 1+2.8 Vulcan ll+OC 36,000 10 69 3I+

50.O 1+6.7 2-50 Foster Vibrator - 80 31

55.0 52.8 Vulcan ll+OC 36,000 13 55 1+0

65.0 63.0 Vulcan ll+OC 36,000 17 80 51

1+1.0 39.0 McK-T-C-5 16,000 1+2 50 3556.O 50.O Vulcan 50,000 ll+O 175 210*

200 C

Remarks

Tapers from 18 in. to 3 in.

0 Vulcan used from 15 to 16.5 ft; pile shattered

Driven in 90 seconds

(Sheet 3 of 1+)

Table A 3 (Concluded)

Dis­trictor Date

Divi­ ofsion Proj ect Tests

SWL Dam N o . 2 , Arkansas River 1965Project

I 1964

1 1964

Lock No. 2 , Arkansas River Project

1965

Lock and Dam No. 4 , Arkansas River Naviga­tion Project

1966

Depth

Generalized Soil ConditionsGWLft

Test Pile No. Type of

O-36 ft select fill material, 36- 56 ft fat Cl, 56-62 ft dense Sd

60 J -2 Wd Class "A"

- 7 .0 5 PC 14 in. sq

9.O G -2 PC l4 in. sq

2 .5 B -5 PC 18 in. sq

Medium to fine sands and silty- sands

2.1 C -3 C l6 in. sq

5.I L-1 0 C 16 i n . sq

6 -3 H -3 Wd l6 in. butt 10 in. tip

4 .7 B- 2 Wd 14 in. butt 8-in. tip

2-5 1 12.75-in. pipe

2 test 1 16-in. pipe

3 20-in. pipe

4 l6-in. concreti

7 14 BP 73

8 Timber10 l6-in. pipe

H I - " I f 16 16 -in. pipe

Lock and Dam No. 6 1965 ,0 -6 ft Si Sd, 6-l4 ft Sd and Si 9.7 K-8 14 BP 73Arkansas River Project Sd, 1 4 -4 2 ft Sd, 4 2 -5 0 ft Sd

and Si Sd, 50-6l ft Sd, 6l- 89 ft Sd and Si Sd

0 -13 ft Sd, 1 3-16 ft fat Cl, 16- 4 .019 ft Sd and Si Sd, 19-38 ft Sd,3 8 -4 1 ft Sd and Si Sd

Pile

19 66 B-1 4 14 BP 73

Driving DataLength of Pile, ft

Em-Driven bedded

Type of Hammer

Energy ft-lb

Blowsper

Lastft

MaxTestLoadtons

Fail­ure

Loadtons Remarks

4 0 .0 3 5 .0 Vulcan 65C 19,200 70 60 4o Prejetted to depth of 27 ft

4 7 .0 4 1 .7 Vulcan l40C 36,000 52 72 53 Jetted to ^ 3 6 ft

4 7 .0 4 2 .8 Vulcan l40C 36,000 67 122 105

45.O 4 3 .0 VulcarL OI6 CO '50 38 I50 100

4 2 .0 3 9 -6 47 139 113

42.0 38.8 44 IO6 93 Jetted to 33 ft

55.O 50.3 Vulcan 65C 19,200 50 39 24 Jetted to 44 ft

55.O 4 7 .3 Vulcan 65C 19,200 92 60 55 Jetted to 39 ft

55.O 53.1 i4oc 36,000 16 92 70

5 5 .0 52.8 38 115 91

55.O 53.0 44 119 90

45.O 4 0 .2 42 96 71

55.O 52.1 8OC 24,500 31 75 41

4o.o 38.6 65C 19,500 23 35 2555.0 53.1 Bodine -- -- 109 8755.0 52.7 140C 36,000 24 79 63 Jetted to 40 ft42.5 39.3 Warrington- 32,500 9 120 70

Vulcan010

4 2 .5 39.4 Warrington- 3 2 ,5 0 0 20 l80 127Vulcan 010

(Sheet 4 of 4)

Lateral Load Tests on Single Vertical Files

Driving DataLength of Blows Max Total Net

Date Depth Test Pile,, ft per Test Deflec­ Deflec­Dis­ of GWL Pile Em­ Type of Energy Last Load tion tiontrict Project Test Generalized Soil Conditions ft No. ■ Type of Pile Driven bedded Hammer ft-lb ft tons in. in. Remarks

LMK Columbia Lock and Dam, La. 1965 0-17 ft fat Cl, 17-30 ft Sd, 30- 7 Ik BP 73 68 54 Vulcan 140C 36 ,0 0 0 io4o 10 0 .6 2 Load applied on web86 ft fat Cl (Ter.), 86-96 F Sd 8 Ik BP 73 68 5k Raymond 0000 48,750 1017 10 0.35 Load applied on flange

MRK N. Kansas City Levee 19 46 0-17 ft lean Cl, 17-46 ft Sd 1 5 .0 i4-8a Prec Cone l8-in. sq 34 3k Vulcan No. 1 1 5 ,0 0 0 48 6 0 .1 0 0 .0 6and Floodwall w/small Gr

ORP Emsworth Dam, Pa. 1936 0-30 ft Sd Gr, Rock at 30 ft 10 BP k2 30 2 8 McK-T 9 B 2 8 ,20 0 27 60 0.05 0.03 Load test on 4-pile group10 BP k2 30 2 8 McK-T 9 B 2 8 ,20 0 27 60 0 .0 6 0 .0 3 Load test on 4-pile group

Blows are for last inch

SPL Sepulveda Dam, Calif. 1939 0 -1 0 ft Cl Sd, IO-8 5 ft Sd Cl C-2 5 1 2 BP 53 ko McK-T 9 B 3 8,750 33 7OH 1 .2 6 0.05w/Gr lenses at 50 and 85 ft 270V

J-3 0 Prec Cone l8-in. 33 McK-T 10 B 3 1 3 ,1 0 0 240 7OH O .76 o.o4Oct 28OV

SWL Arkansas River Navigation Project, Ark.Lock and Dam No. 1 19 6 k 6.0 G-7 Prestress ConeMono L-l8 g - 8 20-in. sq

Lock and Dam No. 3 1965 O-9 ft alternate strain Sd andMono 23 Cl, 9-52 ft fat Cl 0.0 n -6 14 BP 89

0.0 N-7 I8 .0 M -98 .0 M-10 1

Mono 10 1965 5 .5 F-1+ ll+ BP 73I 5 .5 F-5

7 .0 E-33i

7 .0 E-3I+1

Mono1

L-7 1965 1 3 .0 A-31 3 .0 A-1+

13.1* f-61 3.1+ F-71 0 .5 N-ll1 0 .5 N-12

Mono r - 8 1965 9.1* C-l9.1+ C-210.8 E-l10.8 E-211.9 G-l12.0 G-2

Lock and Dam No. 4 1965 3-6 D -3 1 Prec Cone l8-in.3-6 E-3 1

Arkansas River & 1964 O-26 ft Sd, 26-34 ft Si Sd, 3-2 13A 11* BP 73Tributaries 34-38 ft Sd, 38-I+O ft Si Sd,Lock & Dam No. 1+ 40-46 ft Sd, 1+6-1+8 ft Si Sd,

1*8-60 ft Sd, 6O-66 ft Si Sd

47 45 Super Vulcan 200C 5 0 ,2 0 0 327 30 0.53 o.l4 Jetted to 38 ft47 45 Super Vulcan 200C 50 ,20 0 409 30 0.42 0 .1 2 Jetted to 38 ft

37 35 McK-T S 8 26 ,00 0 16 18 0.8437 35 17 18 0 .9 437 36 13 18 0 .5 837 35 17 18 0 .8 9

45 43 Super Vulcan l40C 36 ,0 0 0 17 0 .6 045 43 I I 17 0 .6 045 43 16 0 .6 045 43 i\ \Ï 18 0 .4 3

45 43 2-50 Foster 25 0.5945 43 Vibratory 25 0 .6 145 43 25 0.4845 43 25 0 .5 145 43 45 0 .9 8 0 .2 745 43 45 0 .9 1 0 .1 7

50 47 19 0 .7 850 47 18 0 .8 950 46 25 0.4l50 46 25 0 .6 350 47 25 0 .3 650 47 25 0.55

39 37 Vulcan 0l6 148 50 0 .3 2 Jetted to 31 ft39 36 i4o 50 0 .3 8 Jetted to 30 ft

45 — Vulcan 80C 24,450 20 25 0.60

(Continued)

Table Ah (Concluded)

D riving DataLength o f Blows Max T o tal Net

Date Depth Test P i l e , f t per Test D eflec- D eflec-D is­ o f GWL P ile Em- Type o f Energy L ast Load tio n tio n

t r i c t Pro,i ect Test G eneralized S o il Conditions f t No. Type o f P ile Driven bedded Hammer • f t - lb f t tons in . in . Remarks

Supplemental Data

LMN Morgan C ity Floodw all 1966 0-8 f t f i l l , 8-20 f t C l w /Si s t r a - - 1A Timber Butt lU -in . 44 36.7 - - - 14.5 4 .7 - D eflectio n a t 10 Tt a , 2O-3 2 f t S i , 32-^6 f t S i Tip 7 - in .w/Cl & Sd S i S tra ta

IB 36 .7 - - 4 .2

2A 36.5 2.8

' 2B 36.5 3-b

Table A5

Vertical Load Test on Single Battered Piles

Driving DataBatter Length Vert Blows Max Fail­

Date Depth Test Degrees of Embed­ per Test ureDis- of GWL Pile from Pile ment Type of Energy Last Load Loadtrict Project Test Generalized Soil Conditions ft No. Type of Pile Vertical ft ft Hammer ft-lb ft tons tons RemarksLMK Steele Bayou Drainage 1966 0-29 fat Cl, 29-31 Si Cl, 31-120 1-10 8 12 WF 99 30.5 112 96 Vulcan OR 30,225 38 3OO 280

Structure, Miss. ft F M C Sd l-io 8 12 WF 99 30.5 130 1 1 1 Vulcan. OR 30,225 59 320 3OO Pile redriven w/l8 ft extension1-10 11 12 WF 99 3O .5 97 82 Vulcan. OR 30,225 22 180 160

MRK N. Kansas City Levee 19 16 O-I8 ft Si Sd, 18-28 ft Sd, 16 6-2C Prec Cone l8 in. sq 27 25 Vulcan No. 1 15,000 95 33 _ 15 ton H load applied in directionand Floodwall 28-32 ft Si Sd, 32-16 Sd of batter. 0.18 in. H deflect.

0.07 in. V deflect.SPL Sepulveda Dam, Calif. 1939 0-10 ft Cl Sd, IO-85 ft Sd Cl C-3 12 BP 53 18.5 lo 38 McK-T 10 B 3 13,100 2l 85 85 30 ton H load applied in direction

w/Gr lenses at. 50 and 85 ft batterC-3 12 BP 53 18.5 lo 38 13,100 2-1 18 — Upward loadC-l 12 BP 53 I lo 38 McK-T 9 B 3 8,750 52 18 — Upward loadF-5 Raymond Step Taper 31 32 Vulcan No. 1 15,000 52 70 — 25 ton H load applied in direction

1 batter0-5 Union 7J0XIO t 36 31 McK-T 10 B 3 13,100 3OO 70 — 25 ton H load applied in direction

batterD-29 Prec Cone l8 in. Oct 11 35 31 2l0 70 - - 20 ton H load applied in direction

batterC-ll 12 BP 53 18.5 lo 38 McK-T 9 B 3 8,750 92 H o Load test on 2-pile groupC-15 18.5 1 38 75 120 35 ton H load applied in directioni batterC-28 18.5 T 38 30 280 — Load test on 1-pile groupC-27 16 39 37 16 — V load is total for groupC-26 0 lo 11 — H load varied 0-70 tonsC-25 0 lo 33N-30 Prec Cone l8 in. Oct 16.5 35 33 McK-T 10 B 3 13,100 172 280 __ Load test on 1-pile groupM-30 II 16 35 3l II I 172 — V load is total for groupK-30 0 38 2l0 — H load varied 0-70 tonsJ-30 1f 0 33 F 2l0 —

Note: All pile tests are compression tests, loads applied vertically downward, except as otherwise noted.

T ab le A6

A x ia l Load T e s ts on S in g le B a t t e r e d P i l e s

D r iv in g D a ta_________ T en s io n T e s ts Com pres- Com pres-

D is­t r i c t P ro je c t

Dateof

Test G eneralized S o il Conditions

DfepthGWLf t

TestP ileNo. Type o f P ile

B a tte rDegrees

fromV e r tic a l

Lengthof

P ilef t

VertEmbed­ment

f tType of Hammer

Energy f t - l b

BlowsperLast

f t

Load a t l /l+ -in .

Rise tons

Load, a t 0 .1 - in .

Rise tons

s io n Test Maximum

Load tons

s io n Test F a ilu re

Load tons Remarks

IMM GM & 0 RR B ridge, S. Fork, 1967 0-15 f t S i C l, 15-60 f t F Sd, 7-10 II+ BP I I 7 6.5 7^ 73 Vulcan 06 19,500 51 I 50 _Obion R iver, Union C ity , 60-75 f t Sd and C l, 75-85Tenn. Sd Cl

NAP P oin t P leasan t Canal, 1966 0-26 f t C M F Sd w/F Gr, 0 1 10 BP 1+2 h5 32 19 McK-T 10 B 3 13,100 32 36 26N. J . , S i te 1 26-3^ f t S i Sd and S i C l, 3 ^ 0 2 1+5 32 19 51 38 29

f t C M F Sd

S i te 2 0-18 f t Sd, 18-26 f t S i Sd and Sd, 0 3 30 31 2k 12 1+2 3526-1+1+ f t S i C l, M+-1+5 f t S i Sd 1+ 30 31 21+ 8 1+0 32

5 i+5 36 20 15 39 306 30 20 11 2k 137 1+5 30 20 II+ 28 16

SÜbe 3 0 - lk f t C M F Sd w /s i and p ea t 2 8 30 31 2l+ 16 32 23I le n s e s , 1 4-21+ f t S i C l, 9 30 31 2k 23 1+0 31

2U-29 S i Sd, 29-1+1 f t S i C l, 10 25 1+0 37 12 k9 1+31\ 1+1-1+3 F Sd w/Si len ses 11 25 Uo 37 3 33 25

POF APSS O perations B u ild ing , 196h C 2 Concrete 23 26 Drop Hammer 33,000 2I+0 <30 28 100 — Hammer dropMisawa, Honshu, Japan 25 ton 6.5 f t

C-26 23 36 50,000 300 <30 23 100 — Hammer drop9.8 f t

SPL Sepulveda Dam, C a lif . 1939 0-10 f t Cl Sd, 10-85 f t Sd Cl C-3 12 BP 53 18 l+o 38 McK-T 10 B 3 13,100 2k 90 __

w/Gr len se s a t 50 and 85 f t C-5 Union 7J8xUO 18 36 3k 3OO 65 —D-29 Prec Cone l 8 in . Oct ll+ 35 3k 2I+O 85 "

SVJL Arkansas River N avigation 1965 0-16 f t s p , 16-25 f t C l, 8 .3 F-9 ll+ BP 73 18 68 59 Super Vulcan 36,000 10 225 225P ro je c t , Ark. 25-75 f t Sd, 75-77 f t Cl ll+OCLock and Dam No. 3 Mono L -lL

Supplemental Data

Dis­trict P ro je c t

D riv ing Data Tension T ests Compres­ Compres­s io n T estB a tte r Length V ert Blows Load a t Load a t s io n Test

Date Depth Test Degrees o f Embed­ per l /l+ -in . 0.1 in . Maximum F a ilu reo f G eneralized GWL P ile from P ile ment Type o f Energy L ast R ise R ise Load Load

Test- S o il Conditions f t No. Type o f P i le V e r tic a l f t f t Hammer f t / l b f t tons tons tons tons Remarks

1968 0-56 f t F to M 9-7 58 PC 18 in . square 21.8 52 k5 Vulcan 0 l6 1+8,750 1+0 P ile was not loaded to f a i l u r e -S i Sd w/Tr G p ro je c te d f a i l u r e load 196 to n s .

9 .7 60 21.8 52 kk 35 — __ _ __ P i le was je t t e d to w ith in 5 f t o fp lanned p e n e tra tio n . P i le wasnot loaded to f a i l u r e - p ro je c te d f a i l u r e load 182 to n s .

6.8 5 26.6 52 1+6 79 135 __ __ _ P ile was j e t t e d to w ith in 5 f t o fp lanned p e n e tra tio n . F a ilu reload computed as average o f four methods - ll+l to n s .

• 6.8 6 26.6 52 1+6 65 130 - - — F a i lu re load computed as averageo f fo u r methods - 137 to n s .

Table A7Vertical Load Tests on Pile Groups

Date GeneralizedDis­ of Soiltrict Project Test Conditions

Depth PileNo.Pile

AvgLengthof

AvgEmbed­

GWL Test in Pile mentft No. Type of Pile Group ft ft

LMN Morganza Floodway, New Orleans, Texas and Mexico Railway Co.

19Ì+0 0-5 ft Cl, 5- 010 ft Cl Si,10-15 ft Sici, 15-30 ft ci, 30-U0 ftSi Cl, UO-55ft Cl

T-2 Wd-butt 16 toto 18 in. diam

T-9 Tip 7-5/8 toincl 12 in. diam

56 50

0-30 ft ci, 30-kO ft Si Cl, U0-60 ft Cl

0-35 ft Cl, 35- ^5 ft Si Sd,1*5-75 ft Cl, 75-78 ft veg matter, 78- 92 ft Sd

T-ll Wd-butt 17 to 9 65 60to I9-I/2 in.

T-19 diam Tip9-I/2 to12 in. diam

T-21 Wd-butt I7-I/2 9 65 60to to 19 in.

T-29 diam Tip 9to 11-3A in. diam

T-37 Wd-butt 15 to to I9-I/2 in.T-52 diam Tip

7-I/2 to9 in. diam

16 65 60

Driving Data

Spacing

2-groups at 12 ft OC Piles in group 3 ft OC

k ft OC

b ft OC

b ft oc

Avg Max Fail­Type Blow per Test ureof Energy Last Load Load

Hammer ft-lb ft tons tonsVulcan No. 1 15,000 11 3b 0 _

lb 508 508

11 Ij-98 J498

11 807 807

Table A8

District

LMN

LMK

SWL

Instrumented Pile Load Tests

Date Type Instrumentationof Test Pile Type Load Tests I-Mechanical Strain Rods

Project ' Tests No. Type of Pile Compression Tension Lateral Dynamic II-Bonded Electrical Strain GagesOld River Control Low-Sill 1955 1 1 BP 73 X _ IStructure

2 Pipe 21 in. diam (CE) X X - -3 l b BP 73 X X - -b Pipe 17 in. diam (CE) X X - -5 Pipe 17 in. diam (CE) X X - -6 Pipe 19 in. diam (CE) X X - -

Morgan City Floodwall 1966 A Wd, Butt-lU-in. - - X - IITip-7 in.

B - - X -A - - X -B - - X -

Columbia Lock & Dam 1965 1 l b BP 73 X X - - II2 X X - -

3 X X - -

1+ Pipe 18 in. 0D (OE) X X - -5 Pipe 18 in. 0D (OE) - X - -

7 14 BP 73 - - X -8 - - X -C X - - -

Arkansas River & Tributaries 196^ l Pipe 12.75 in. 0D (CE) X X _ _ CLock & Dam U

2 Pipe 16 in. 0D (CE) X X X - I & II3 Pipe 20 in. OD (CE) X X - - I6 l b BP 73 X - X - I7 X X - - I & II9 X - - - I

10 Pipe 16 in. OD (CE) X X X - I & II12 I k BP 73 - - X - II13 - - X -13A - - X -16 Pipe 16 in. OD (CE) X X X -

Remarks

Dynamic strains measured during driving

Dynamic strains measured during driving

Dynamic strains measured during driving

(Continued)

DistrictSAM

Date. of Test Pile

Project___________ Tests No. _____ Type of PileNASA 1962Mississippi Test Facility 1963 S-II Test Stand

S-l 14 BP 73S-2 Pipe l k in. OD (CE)

Concrete filledD-l Pipe l b in. OD (CE)

Concrete filledD -2 lU BP 73

D-3

D -k

Table A8 (Concluded)

________Type Instrumentation_______________ Type Load Tests____________ I-Mechanical Strain RodsCompression Tension Lateral Dynamic II-Bonded Electrical Strain Gages Remarks

XX

X

XX

X

II

X

X

X

X

Pile D-3 was in a four-pile group with a common pile cap

APPEND IK B: INSTRUMENTATION FOR PILE LOAD TESTS

Introduction

1. The type of instrumentation used in a pile load test depends on the type of test being conducted and the information desired from the test. In the most common type of pile load test, the vertical load is placed on the top of a driven pile and the vertical movement of the top of the pile is measured. In more sophisticated tests, the distribution of load along the pile length is determined. In lateral load tests, the lateral load is applied to the pile head, and the bending moments and bending stresses in the pile are determined. Therefore, depending on the type test performed, instrumentation may be needed to measure:

a . Vertical or lateral load applied to the pile.b . Vertical or lateral movement of the top of the pile.c. Axial stresses along the length of the pile.d. Bending stress in the pile.

This appendix describes instrumentation capable of obtaining the above measurements.

Pile Load Measurements

2. In a compression load test, a vertical load is applied to the top of the pile with a hydraulic jack which reacts against a reaction beam. The reaction beam, in turn, acts against a loading platform, or it may be attached to reaction piles driven some distance away from the test pile. A loading platform is preferred over reaction piles, since there is always a question of how far away from the test pile the re­action piles must be driven to prevent influence of the reaction piles on the test pile behavior. If reaction piles are used, they are nor­mally driven at least 5 pile diameters away from the test pile. Re­action piles driven in cohesionless soils should be located farther from the test pile than those driven in cohesive soils, since the driving of the reaction piles can affect the relative density of the

B1

material surrounding the test pile. The load induced in the soil by the reaction pile can also affect the pressure distribution in the vicinity of the test pile. Various investigations have pointed out that piles interact even beyond spacings of 8 diameters. Anchor piles should therefore be located a sufficient distance from the test piles so that interaction -will not occur.Hydraulic jack

3. Load transmitted to the pile can be determined by measuringthe pressure in the hydraulic jack with a Bourdon gage. Although it isusually required that the jacks and gages be calibrated both prior toand after the load test, it is apparent that significant errors may be

11introduced by means of this type of load measurements. Davisson haspointed out that eccentric loading produces friction in the rim of thehydraulic jacks which can result in errors on the order of 10 to 20 per-

52cent of the applied load. Cole states that errors of as much as 20 percent are possible. The use of instrumented piles provides a means for checking the accuracy of the applied load. Particularly on piles in cohesionless soils, the frictional resistance near the top of the pile will be relatively small; thus, the uppermost strain rod or strain gage should serve as a close check on the applied load. A com­parison of loads determined from the hydraulic jack pressure gage with loads computed from strain measurements for several CE projects is shown in table Bl. It appears that appreciable errors are introduced by rely­ing on the hydraulic jack pressure gage values. These errors can amount to as much as 29 percent of the applied load in some instances.

k. For Important pile test programs, it is essential that accu­rate load measurements be obtained. The use of a ball or swivel arrange ment to reduce the effects of eccentric loading can greatly minimize theerror from misalignments. A typical load test arrangement used at

I4.I4.Jonesville Lock employing a swivel connection is shown in fig. Bl.Load cell

5. To obtain better accuracy in measuring the load applied to the pile, a calibrated load cell can be placed between the hydraulic jack and the pile or between the jack and the reaction platform. Load cells

B2

5-TON LOAD LOGS (82)

Fig. Bl. Compression loading test

used in pile load tests generally employ electrical strain gages bonded to internal steel members so that load on the device can be determined by measuring the strain in the steel members. Such cells make it pos­sible to measure applied loads to an accuracy of better than one

9percent.

Movement of Pile Butt

Dial gages6. Vertical movement of the top of the pile can be measured with

dial gages (extensometers) attached to a reference beam and having their stems in contact with a steel bracket attached to the pile butt. At least two dial gages should be used, one on either side of the pile at equal distances from the center of the pile. If more than two are used, they should be equally spaced around the pile and at equal radial

B3

distances so that their readings can be averaged to obtain the settle­ment of the center of the pile. Dial gages should have a minimum ac­curacy of at least 0.001 in. Every effort should be made to insure that the dial gages are vertical and to insure that they are measuring ver­tical movements.

7- The reference beam for the dial gage should consist of a steel beam extended across two steel anchors driven into the ground as illus­trated in fig. B2. The anchors should be located at a minimum distance

Fig. B2. Reference beam for dial gagesof 5 pile diameters from the test pile to eliminate drag effects which would cause settlement of the reference beam. One end of the reference beam should be bolted or welded to one anchor; the other end should be cradled in a smooth saddle over the second anchor to permit the beam to expand and contract without bending due to temperature changes. The reference beam should be protected from sunlight and undue temperature changes and protected against disturbance. The elevation of the top of

Bh

the reference beam should be checked with an engineer’s level at regular intervals during the load test to insure that the position of the beam has not been disturbed during the test. The benchmark used for the level survey should be at a sufficient distance from the test so that it will not be disturbed by activities in the pile test area.LVDT's

8. In some cases, linear variable-displacement transducers (LVDT) are used to measure vertical movement of the pile butt. The LVDT is basically a miniature transformer with primary and secondary windings coupled by a free-moving cylindrical magnetic core which travels through the center of the instrument. The device has the same accuracy as a dial gage. The main advantage of the LVDT is that readings can be auto­matically recorded; thus, the device is suitable for measuring movement of the pile during a dynamic load test. The devices are installed in the same manner as dial gages.Wire-scale-mirror system

9- The wire-scale-mirror method is described by Davisson."*”*" The system consists of a vertical scale attached to a mirror which in turn is attached to the pile head. A piano wire is extended just in front of the scale. The scale can be read by sighting directly across the wire into the mirror until the extra image of the wire in the mirror is elim­inated. A scale 6 in. long divided into 0.02-in. increments can be read to the nearest 0.01 in. The scale should be in a vertical plane, and the wire should be in a horizontal plane. Stakes for the horizontal wire should be at least 5 pile diameters away from the pile. A turn- buckle should be used on the wire to tighten it sufficiently after it has been tied to the stakes. At least two systems should be used, one directly on the opposite side of the pile from the other. The eleva­tions of the reference wires should be checked with an engineer’s level at regular intervals during the pile load test.

Load Transfer Test

10. In addition to measuring the vertical movement of the pile

butt during a pile load test, the transfer of load from the pile to the surrounding soil can be estimated by determining the distribution of load along the length of the pile. The distribution of load in the pile can be determined indirectly by measuring the vertical deformations of several points along the pile with respect to the top of the pile with strain rods or by measuring strains at various points along its length with bonded electrical strain gages attached to the pile.Strain rods

11. Strain rods or "telltales," as they are sometimes called, consist of small-diameter steel rods with the lower end attached to some point along the length of the pile and the upper end extending to the top of the pile. A dial gage measures the movement of the top of the rod with respect to the top of the pile. The load in the pile is computed by the formula

P = EA (Bl)

whereP = average load between two strain rod anchors, lbE = modulus of elasticity of the pile, psiA = cross-sectional area of the pile, sq in.

■Ze = vertical deformation between the rod anchors, in.ZJj = vertical distance between two rod anchors, in.

Details of strain rods used for pile load tests on steel piles at the12Old River Low-Sill Structure are shown in fig. B3. The rods are iso­

lated from the pile, and a protective cover is provided around the rods.Strain rods developed for concrete piles by Raymond International, as

53described by Snow, are shown in fig. BU. Strain rods are generally unsatisfactory in determining residual loads due to driving or in cases in which significant bending strains may occur.Bonded strain gages on steel piles

12. Bonded strain gages have been used successfully for deter­mining distribution of load in steel piles. The strain measured at a point along the length of a pile can be readily converted into stress

B 6

CONTINUOUS WATERTIGHT WELD

STRAIN ROD ASSEMBLYPIPE PILE SHOWN, H -P IL E SIMILAR

TYPE B LOWER ROD ANCHOR SUPPORT

IN .-D IAM STEEL RODS

<L PILETYPE A LOWER ROD ANCHOR S U P P O R T

TYPE B LOW ER ROD ANCHOR SUPPORT

sec .T,|o n ..Az A

PIPE PILES

SECTION B -B

H -P IL E S

GAGE ASSEMBLY FOR STRAIN RODSNOT TO SCALE

. B3.Fig Mechanical strain rod for steel piles

Top 3" P la te

Fo u r 2” x 2" V e rt ica l Colum ns

REM O TER E FE R E N C E

P la s t ic Tubing

Telltale Rod

3/8" N.C. Thread R. H.

T e llta le R ods

E x t r a Dial Gauge to d e te ct possible crushing at b u tt.

S P E C IA LT E LLTA LE

A S S E M B LY -

T E LLTA LE = A S S E M B LY

- P L A S T IC TUBING - T E L L T A L E ROD

-TRA N SITIO N ROD

L E F T -H A N D THREAD

3/4" P IP E

P L A S T IC TUBING T E L L T A L E ROD

TRANSITION ROD

L E F T -H A N D TH R EAD

T O G G L E SW INGS OUTWARD DUE TO U N B A LA N CE AND

v C O M ES TO R E ST ON SH O U LD ER

Fig, BU. Telltales for concrete piles (from reference 53)

b8

since the modulus of elasticity of steel is a fairly constant value. Stress can then be converted to load knowing the cross-sectional area of the pile and assuming that stress at the gage point is equal to the av­erage stress over the cross section.

13. Various arrangements of bonded strain gages have been used. The most successful have been arrangements in which the strain gages are connected to form a Wheatstone bridge at the gage point on the pile in order to minimize effects of lead wire resistance change. For a short-term test in which no great temperature changes are anticipated in the pile, two strain gages and two resistors can be used at each point, as shown in fig. B5a. For a long-term installation, four gages should be used, two acting as "active" gages and two acting as "dummy" gages (see fig. B5b). The latter arrangement compensates for effects of tem­perature changes in the pile. In reducing data from this type of gage point, a correction factor is applied to compensate for lateral strains measured by the two dummy gages.

lU. Bonded strain gages should be installed only by qualified personnel. Installation requires carefully smoothing the steel surface, bonding the gages and terminal to the pile with epoxy, and then applying waterproofing compound over the gages. Strain gages should be located in such a manner that bending and axial strains can be differentiated during pile driving and load testing.

15. In early tests in which bonded electrical strain gages wereused in pile load tests, protective covers were provided over the gagesand cables to protect against damage during pile driving. Details of

b7cover plates used for the Columbia Lock pile load tests are shown in fig. B6. However, more recent experiences indicate that a strong epoxy- cover over the gages is sufficient to protect the gages mounted on the web of an H-pile.Internal strain gages in concrete

16. Bonded strain gages designed for use in concrete have been used in concrete pipes for pile load tests. One type of gage, Bald­win’s Valore gage, is a bonded strain gage wrapped in a waterproof foil envelope. The "polyester mold gage" manufactured by Tokyo Sokki

B9

C I R C U I T D I A G R A M

D E T A I L O F S T R A I N G A G E S

a. Installation for short-duration test

b. Installation for long-duration test

Fig. B5. Installation of bonded electrical strain gagesfor pile load test

BIO

DETAIL AT TOP OF PROTECTIVE COVER DETAIL AT BOTTOM OF PROTECTIVE COVER

Fig. B6. Protective cover for bonded strain gages for a pile load test

Bll

Kennkyujo Company consists of a banded strain gage hermetically sealedbetween two thin polyester blocks. Although these gages are reliablefor short-term tests, they are not recommended for use over an extended

5k 55period of time. A typical gage installation by Texas A&M is shownin fig. B7. The four gages are connected to form a Wheatstone bridge

to eliminate temperature effects and reduce effects of lead wire resis­tance change.

17• Bonded strain gages can be installed directly on reinforcing bars for concrete piles. Four gages are installed at each gage point;

B12

the arrangement of these gages is the same as that shown in fig. B5.The "stressmeter" manufactured by Structural Behavior Labs., Inc., shown in fig. B8, consists of a I4-—ft length of A-U3 2 reinforcing bar with a

Fig. B8. Stressmeter for embedment in concrete pile (from reference 55)

full U-arm bridge of bonded strain gages. The gages are attached to the bar by resistance welding.

Lateral Load Tests

18. Lateral load tests are conducted to determine the modulus of horizontal subgrade reaction to be used in the pile foundation design.In this type of test, a horizontal load is applied to the pile butt in­stead of a vertical load. The horizontally applied load can be measured with a Bourdon gage on the hydraulic jack or with a calibrated load cell. Horizontal movement of the pile head can be measured with dial gages or LVDTTs.

19. Determination of the subgrade reaction requires the computa­tion of moments in the pile from the measured lateral deflection or measured bending stresses in the pile. Pile instrumentation for a lat­eral load test may include:

a. Strain gages to determine bending stresses in the pile.b. An inclinometer tube to determine the deflected shape of

the pile with an inclinometer.

Strain gages20. Strain gage installation for a lateral load test is

B13

essentially the same as that for a conventional (vertical) load test, except that the gages are installed at the outermost edge of the pile in the plane of bending in order to measure maximum bending strains. Instrumentation for a steel H-pile is shown in fig. B9* Moments in

ELECTRICAL CABLE PROTECTION AT TOP OF PILE

DETAIl AT BOTTOM OF PROTECTIVE COVER

N O TE: STRAIN G AG ES W ERE BALDWIN-LIMA-HAMILTON T Y P E FAB-50-12- 56. RESIST O R S W ERE PREC IS IO N R ES IS T O R CO ., T Y P E TX-176.C A B L E S W ERE B E L D E N T Y P E 8434.

Fig. B9. Gage point installation, lateral load tests

the pile are computed from measured strains using the equation

M alM = —e£EIe (B2)

•wheree = measured strain at the location of the gage a = measured stress at the location of the gage E = modulus of elasticity of the pileI = moment of inertia of the pile about its bending axis e = distance from the bending axis of the pile to the strain gage

Deflection measurements21. In some lateral load tests, the lateral deflection of the

pile is measured with an inclinometer. An inclinometer casing attached5 6to the pile as described by Hanna is shown in fig. BIO. Various types

of commercially available inclinometers are described in EM 1110-2- 1908. Initial inclinometer measurements are made before any load is applied. Subsequent readings are taken as the loading proceeds. The difference between the latter readings and the initial readings permits calculation of the deflection of the pile due to the applied load. Moments and bending stresses are computed from the measured deflections.

Pig. BIO. Inclinometer casing attached to H-pile (from reference 56 )

B15

Table ELComparison of Jack Loads with Loads Computed from Measured Strains

Compression Tests________________ ________________ Tension Tests

ProjectTestPile

JackLoadtons

Strain Rod Strain Gage JackLoadtons

Strain Rod Strain GageComputed Percent

Load, tons ErrorComputed

Load, tonsPercentError

Computed Percent Load, tons Error

Computed Perceni Load, tons Error

Old River Low- 1 3 3 3 286 l 6 .k . . _ _

Sill Structure 2 3 3 3 296 1 2 . 5 - - - - 195 182 7.2 - - - -

3 2 k k 2 2 2 9 . 9 -- -- 60 50 16.7 -- --3bo 322 5 . 6 -- -- 198 * * -- --

5 1 ^ 5 133 9 . 0 -- - - 80 7 6 5 . 0 — --6 3 ^ 5 3^1 1 . 2 -- -- 165 166 0 -- --

Arkansas River 1 1 7 2 159 8 . 2 — 92 * ___ ___ ___

Lock and Dam No. k

2 2 1 3 191 11.5 -x- -- 96 7 7 2 k . 7 7^ 2 9 . 7

3 2 3 5 225 -- -- 115 1 0 0 15.0 -- --7 2 0 1 172 16.9 1 8 3 9 . 9 67 5 k 2 ^ . 0 58 1 5 . 5

1 0 178 139 28.0 1 5 8 1 2 . 7 105 82 28.6 90 1 7 . 2

16 162 -- -- * — 6 6 -- -- 53 2 k . 5

Columbia Lock 1 260 ___ ___ 2 ^ 8 ^ . 9 200 — ___ 185 7 . 5and Dam

2 300 -- — 295 1 . 7 200 -- -- 215 -7 . 0

3 300 -- -- 325 >7 . 7 2 0 0 -- — 185 7 . 5

k 300 -- -- 275 9 . 1 2 0 0 -- — 175 1 2 . 5

5 300 -- -- 275 9 . 1 2 0 0 -- — 175 1 2 . 5

* Data questionable

U n c la s s if ie d Security C lassification

DOCUMENT CONTROL DATA - R & D(S e c u rity c la s s if ic a t io n o f t i t le , body o f a b s tra c t an d in d e x in g an n o ta tio n m ust be e n te re d w hen the o v e ra ll re p o rt is c la s s i f ie d ) ^

l . O R I G I N A T I N G A C T I V I T Y (C o rp o ra te au th o r)U. S. Army E n g in eer W aterways E xperim ent S ta t io n V ick sb u rg , M is s is s ip p i

• 2a . R E P O R T S E C U R I T Y C L A S S I F I C A T I O NU n c la s s if ie d

2 6 . G R O U P

3. R E P O R T T I T L E

ANALYSIS OF PILE TESTS

4 . D E S C R I P T I V E n o t e s ( T y p e o f re p o rt a n d in c lu s iv e d a te s )F in a l r e p o r t

5- A U T H O R I S ) ( F i r s t n am e, m id d le in i t ia l , la s t n a m e )

W alte r C. Sherman, J r .D. M ichael Holloway C h arle s C. Trahan

6 . R E P O R T D A T E

A p r i l 197b7a. T O T A L N O . O F P A G E S

13b7b. N O . O F R E F S

578 a . C O N T R A C T O R G R A N T N O .

6 . P R O J E C T N O .

9 a . O R I G I N A T O R ' S R E P O R T N U f c T B E R ( S )T e c h n ica l R eport S-7^—3

9 6 . o t h e r R E P O R T N O ( S ) (A ny o th er nu m bers th a t m ay be a s s ig n e d th is report)

10. D I S T R I B U T I O N S T A T E M E N T

Approved f o r p u b l ic r e l e a s e ; d i s t r i b u t i o n u n l im i te d .1 1 . S U P P L E M E N T A R Y N O T E S 12 . S P O N S O R I N G M I L I T A R Y A C T I V I T Y

O ff ic e , C h ie f o f E n g in e e rs , U. S. Army

13. A B S T R A C T

W ash ing ton, D. C.

The p u rp o se s o f t h i s s tu d y w ere t o e v a lu a te p i l e lo a d t e s t d a ta o b ta in e d from Corps o f E n g in eers o f f ic e s and to compare th e s e d a ta in l i g h t o f a n a l y t i c a l d e s ig n m ethods f o r p r e d ic t in g p i l e lo a d c a p a c i t i e s .Though many t e s t s were p e rfo rm e d , v e ry few p e rm it te d a d e t a i l e d a n a ly s i s o f th e b e h a v io r o f th e p i l e - s o i l sy stem . C a re fu l ly in s tru m e n te d p i l e lo a d t e s t s , such as th o se p erfo rm ed a t Old R iv e r L o w -S ill S t ru c tu r e and A rkansas R iv e r Lock and Dam Wo. b , p ro v id e d th e o n ly so u rce s o f d a ta f o r w hich th e p i l e - s o i l i n t e r ­a c t io n cou ld be exam ined in s u f f i c i e n t d e t a i l . I t was found t h a t th e c o n v e n tio n a l s t a t i c p i l e c a p a c i ty fo rm u las do n o t a d e q u a te ly d e s c r ib e th e b e h a v io r o f p i l e s in c o h e s io n le s s s o i l s . Load t e s t r e s u l t s fo r p i l e s i n sands in d ic a te t h a t p i l e - s o i l i n t e r a c t io n and s o i l c o m p r e s s ib i l i ty in th e v i c i n i t y o f th e t i p may make th e f r i c t i o n a l r e s i s t a n c e and t i p r e s i s t a n c e in te rd e p e n d e n t . The u n i t s k in f r i c t i o n computed from f i e l d m easurem ents te n d s to in c re a s e l i n e a r l y w ith d ep th o n ly a t sh a llo w d e p th s ; t h e r e a f t e r i t appro aches a l im i t in g v a lu e below a d ep th o f 10 to 20 p i l e d ia m e te rs . For te n s io n p i l e s , t h i s l im i t in g v a lu e rem ains e s s e n t i a l l y c o n s ta n t , w hereas f o r com pression p i l e s , th e u n i t s k in f r i c t i o n d e c re a se s n e a r th e p i l e t i p . O ther i n v e s t ig a to r s have r e p o r te d s im i la r o b s e rv a t io n s . E x tr a p o la t io n o f f i e l d d a ta to d e s ig n p i l e fo u n ­d a t io n s , b a sed upon c o n v e n tio n a l m ethods, may produce s i g n i f i c a n t , u n c o n se rv a tiv e e r r o r s . For p i l e s in s o f t t o medium c la y s , th e lo a d t e s t s in d ic a te t h a t th e c o n v e n tio n a l m ethods o f a n a ly s i s u s in g u n d ra in e d sh e a r s t r e n g th a re g e n e ra l ly s a t i s f a c t o r y and p ro b a b ly c o n s e rv a tiv e f o r lo n g - te rm b e h a v io r . L im ited d a ta f o r p i l e s in s t i f f c la y s su g g e s t many u n c e r t a in t i e s in e v a lu a t in g fo u n d a tio n p e rfo rm a n c e . Time e f f e c t s and r e l a t e d phenomena make th e s e c o n d itio n s m ost d i f f i c u l t t o a n a ly z e . The r e s u l t s o f t h i s s tu d y in d ic a te t h a t f u r th e r re s e a rc h i s n e c e s s a ry t o p ro v id e c le a r e r i n s ig h t i n to th e p i l e - s o i l i n t e r a c t io n p ro b lem s.The b e h a v io r o f p i l e s in c o h e s io n le s s s o i l s d e v ia te s s i g n i f i c a n t l y from t h a t p r e d ic t e d by c o n v e n tio n a l t h e o r i e s . In o rd e r t o d e s ig n p i l e fo u n d a tio n s p ro p e r ly and i n t e r p r e t th e r e s u l t s o f p i l e lo a d t e s t s c o r ­r e c t l y , a more r a t i o n a l m ethod o f a n a ly s i s i s u rg e n t ly need ed .

DD FORM 1 NOV 6 5 1473 REPLACES DD FORM 1473, 1 JAN 64, WHICH IS OBSOLETE FOR ARMY USE. U n c la s s if ie dS ecurity C la s s if ic a tio n

UnclassifiedSecurity C lassification

1 4 .K E Y W O R D S

L I N K A L I N K B L I N K C

R O L E W T R O L E W T R O L E W T

ClaysLoad tests Pile foundations Pile load tests PilesShear strength Soil compacting SoilsSoil strength

Unclassified Security C lassification

In accordance with ER 70-2-3, paragraph 6c(l)(b), dated 15 February 1973 a facsimile catalog card in Library of Congress format is reproduced below.

Sherman, Whiter CharlesAnalysis of pile tests, by W. C. Sherman, Jr., D. M.

Holloway cand^ C. C. Trahan. Vicksburg, U. S. Army Engi­neer Waterways Experiment Station, 197^*

1 v. (various pagings) illus. 27 cm. (U. S. Water­ways Experiment Station. Technical report S-7^-3)

Sponsored by Office, Chief of Engineers, U. S. Army. Includes bibliography.

1. Clays. 2. Load tests. 3» Pile foundations, k. Pile load tests. 5« Piles. 6. Shear strength. 7* Soil com­pacting. 8. Soil strength. 9* Soils. I. Holloway,D. Michael, joint author. II. Trahan, Charles Curtis, joint author. III. U. S. Army. Corps of Engineers. (Series: U. S. Waterways Experiment Station, Vicksburg, Miss. Technical report S-7^-3)TA7.W34 no.S-7^-3