10 I?9A

- 5«)l CIVIL ENGINEERING STUDIES

C • ?.- Structural Research Series No. 581

UILU-ENG-93-2004

ISSN: 0069-4274

SEISMIC RESPONSE OF TILT-UP CONSTRUCTION

By

James w. Carter III Neil M. Hawkins and Sharon L. Wood

A Report to the National Science Foundation Research Grant BCS 91-20281

Department of Civil Engineering University of Illinois at Urbana-Champaigr:-a Urbana, Illinois

December 1993 l£:etg RS:~~~ I{c.~m U:rriveZ'stty (;f miz~(;i3

B10S NZ'Rl:r~"l'k CE L~ 2C6 North Mati~ews Avenue

Urbrola, Illinois 618D1

50272-101

REPORT DOCUMENTATION I 1. REPORT NO.

PAGE UILU-ENG-93-2004

4. Title and Subtitle

Seismic Response of Tilt- Up Construction

7. Author(s)

James W Carter III, Neil M. Hawkins, and Sharon L. Wood

9. Per10rming Organization Name and Address

University of Illinois at Urbana-Champaign Department of Civil Engineering 205 North Mathews Avenue Urbana, Illinois 61801

12. SponSDring Or\1&l1i.z.at;on Name and Addre~

National Science Foundation 4201 Wilson Boulevard, Room 545 Arlington, Virginia 22230

15. Supplementary Notes

16. Abstract (Limit: 200 words)

2. 3. Recipient's Accession No.

s. Report Oat.

De~ember 1993 6.

8. Per10rming Organization Report No.

SRS 581 10. Project/TaskiWorX Unit No.

11. Contrsct(C) or Grant(G) No.

BCS 91-20281

13. Type 01 Report & Period Covered

14.

Tilt-up construction is an economical method of constructing low-rise industrial and commercial buildings.

However, the structures are susceptible to seismic damage. The seismic response of tilt-up construction is

reviewed in this report. Construction techniques are described, design procedures are summarized, the results

of previous experimental and analytical research are compared, and observed damage during recent earth­

quakes is discussed. The measured response of three tilt-up buildings in California is also presented.

17. Document Analysis B. Descri;:>tors

Low- Rise Buildings, Precast Concrete Wall Panels, Roof Diaphragms, Connections, Earthquakes, Design Provisions, Observed Damage, Measured Seismic Response

b. Identifiers./O;:>en - Ended Terms

c. COSATI Field/Group

is. Availability Statement t 9. Security Class (Thi. Reportj

UNCLASSIFIED

20.SCCtJrity Class (This P~)

UNCLASSIFIED

224

22.Pri~

i (See ANSI-239.18) OPTIONAL FORM 272 (4-77)

Department 01 Commerce

SEISMIC RESPONSE OF TILT-UP CONSTRUCfION

1. Introduction.................................................................. 1 1.1 Past Seismic Perfonnance ................................................ - . 2 1.2 Research Needs. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2 1.3 Objective and Scope ........... _ . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3 1.4 Acknowledgements. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3

2. Construction of Tilt-Up Structures. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4 2.1 Wall Panel Construction ..... . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4 2.2 Roof Construction. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5

2.2.1 Wood Diaphragms .................... , . . . .. ... .. . . . ... . .. ... . .. . . . . 5 2.2.2 Metal Deck Diaphragms ................. _ . . . . . . . . . . . . . . . . . .. . . . .. . .. . 6 2.2.3 Composite Diaphragms .................................. _ . . . . . . . . . . . 6

3. Design of Tilt-U p Structures to Resist Lateral UJads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7 3.1 Wall Panels ............................................................. 7 3.2 Diaphragms ............................................... _ . . . . . . . . . . . . . 9

3.2.1 Diaphragm Strength and Stiffness ...................................... 10 3.2.2 Diaphragm Fasteners ................................................ 13 3.2.3 Design of Non-Rectangular Diaphragms and Diaphragms with Openings ..... '. 15

3.3 CanneL'jons in Tile-Up Systems ............................................ 16 3.3.1 Panel-la-Foundation Connections. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 16 3.3.2 Panel-ta-Panel Connections .......................................... 16 3.33 Panel-ta-Roof Connections .............. _ .................. _ ...... _ . . 17

3.4 Summary ............... _ .. __ . _ ..... _ .... _ . _ ................ = = • • • • • • • • • • 19

4. Summary of Previous Investigations of the Seismic Behavior of rtlt-Up Construction .... 20 4.1 Cyclic Response of Diaphragms _ . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 20

4.1.1 Experimental Tests of Wood Panels and Diaphragms ..................... '. 21 4.1.2 Ex perimental Tests of Metal-Deck Diaphragms ........................... 22 4.1.3 :~.nJlytical Models of Diaphragms ............. _ ...................... " 22

4.2 C~'clic Response of Tilt-Up Wall Panels ...................................... 25 4.3 AnJlytJcal \1odels of Tilt-Up Systems ........... _ . ..... .. . ... . . . . . .... . . .... 25 4.4 SummJr\' 28

5. Measured Stro~)\lotion Response Of Tilt-Up Buildings. . . . . . . . . . . . . . . . . . . . . . . . . . . . 30 5.1 Hollister \1.arehouse ............................ _ .. _ .. _ .. ' .. _ . . . . . . . . . . . . . 30 5.2 Red13nds V,';,trchouse ........................... _ .. _ ...................... . 5,3 l\1ilpli...1~ Indu.\tnJI Building ..................... _ . _ ...... _ ..... _ .......... .

5.4 SummJr\

32 34 36

6. Observed Performance Of Tilt-Up Construction During Earthquakes ................. 37 6.1 The 1964 Alaska Earthquake ..................... __ ..... _ . . . . . . . . . . . . . . . . . . 37 6.2 The 1971 San Fernando Earthquake ........... _ .... _ . _ . . . . . . . . . . . . . . . . . . . . . . . 37 6.3 The 1987 Whittier Narrows Earthquake ... _ ..... _ ........ _ .... _ .. _ . . . . . . . . . . .. 38 6.4 The 1989 Lorna Prieta Earthquake ...... _ ...... _ ............... _ . _ . . . . . . . . . . . 39 6.5 Summary ......... _ ..... - ............. _ .... _ . - .............. _ . . . . . . . . . . . 40

ii

7. Conclusi<lns ......................................•........................... 41

References ..................................................................... 43

Tables ....................................................................... 49

Figures .......... - . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65

Appendix A - Summary of UBC Prov~ioos . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .. 212

Appendix B - Filtering of Relative Displacement Histories ........................... 218

III

LIST OF TABLES

Table Page

3.1 Allowable shear in lblft for horizontal plywood diaphragmswith framing of douglas fir-larch or southern pine [77] ............................. 49

3.2 Diaphragm shear stiffness [3J ........................................... _ . 50

3.3 Fastener slip equations [74J ........... . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 50

3.4 Strength of metal-<ieck diaphragms [48J . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 51

3.5 Strength and flexibility of connections in metal deck diaphragms [48] ............. 52

3.6 Strength of composite diaphragms [48J ..................................... _ 53

3.7 Maximum diaphragm dimension ratios [77J .................................. 53

4.1 Summary of experimental tests of diaphragms subjected to load reversals. . .. . . . . . . . 54

4.2 Summary of ABK diaphragm tests (1J . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .. . . . . . . 55

4.3 Summary of dynamic tests of tilt-up wall panels [6, 8, 7] ...................... 0 56

5.1 Characteristics ofinstrumented tilt-up buildings .............................. 57

5.2 Summary of strong-motion instruments in the Hollister warehouse. . . . . . . . . . . . . . . . 58

5.3 Summary of relative displacement data from the Hollister warehouse . . . . . . . . . . . . . . 59

5.4 Summary of the strong-motion instruments in the Redlands warehouse .......... _ . 60

5.5 Summary of relative displacement data from the Redlands warehouse 61

5.6 Summary of the strong-motion instruments in the Milpitas industrial building ............................................... 62

5.7 Summary of relative displacement data from the Mil pi tas i nd ustrial building ............................ _ . . . . . . . . . . . . . . . . . . 63

6.1 Summary of observed damage in tilt-up buildings following earthquakes in the U.S. .......................................... 64

B.l Filter limits used to process strong-motion records ............................ 220

iv

LIST OF FIGURES

Figure Page

2.1 Common arrangement of panel pick points [79] ............................... 65

2.2 Tilting procedure ....................................................... 66

2.3 Examples of improper placement of openings [78] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 67

2.4 Locations of openings within tilt-up panels [78] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 68

2.5 Roof framing member connected to the tilt-up panels at the panel-to-panel joint [78] ...................................... 69

2.6 Common types of diaphragms used in tilt-up construction throughout the U.S. [10] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 70

2.7 Typical plywood diaphragm [35] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 71

2.8 Blocked section of a plywood roof diaphragm [57]. ............................ 71

2.9 Prefabricated section of a plywood roof diaphragm [33] ........................ 72

2.10 Plan view of a steel diaphragm [64] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 73

2.11 Typical Zr- and C- type shear transfer connections [64] ......................... 74

3.1 Idealized force distribution in diaphragm [33] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 75

3.2 Idealized model of a tilt-up wall panel [16] .................................. 76

3.3 Test configuration for AO-SEASC lateral load tests [16]. . . . . . . . . . . . . . . . . . . . . . . . 76

3.4 Free body diagram of a tilt-up panel subjected to lateral loading [16] .............. 77

3.5 Distribution of lateral forces to supporting walls in structures with flexible and rigid diaphragms [65] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 78

3.6 Configuration of metal~eck diaphragm . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 79

3.7 Corrugations in metal decking [48] ......................................... 80

3.8 Strength of metal~eck diaphragm limited by connections along edge of diaphragm [48] ................................... 81

3.9 Strength of metal~eck diaphragm limited by connections in an interior panel [48] ........................................ 81

3.10 Strength of metal~eck diaphragm limited by connections in comer [48] ................................................ 82

3.11 Deformation of metal deck [48] ................ . ......... ..... . ... .... ..... 83

3.12 Pull-out strength of various roofing nails [20] ................................ 84

3.13 Displacement discontinuities in non-rectangular diaphragms [14] . . . . . . . . . . . . . . . . . 85

3.14 Displacement discontinuities in buildings with stairwells and elevator cores [14] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 86

v

Figure

3.15 Analysis of diaphragm modelled as a Vierendeel truss [74]

3.16 Typical panel-to-foundation connections in buildings

Page

87

constructed before 1971 [55,80] ........................................... 88

3.17 Typical panel-to-foundation connections in buildings constructed after 1971 [40] ............................................... 89

3.18 Typical panel-to-panel connections in buildings constructed before 1971 [55,80] ........................................... 90

3.19 Typical panel-to-panel connections in buildings constructed after 1971 [40,80] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 91

3.20 Typical purlin to wood ledger connection in buildings constructed before 1971 [35] .......................... . . . . . . . . . . . . . . . . . . . . 92

3.21 Typical plywood to wood ledger connection in buildings constructed before 1971 [25] .............................................. 92

3.22 Typical purlin to wood ledger connection in buildings constructed after 1971 [25] ............................................... 93

3.23 Plan view of roof framing members showing subdiaphragm details. . . . . . . . . . . . . . . . 94

3.24 Strengthening of existing panel-to-roof connections for walls wi th parapets [25] ............................................... 95

3.25 Strengthening of existing panel-to-roof connections at top of wall [25] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 96

3.26 Purlin-to-purlin connection across a glulam beam [22] ......................... 96

3.27 Use of metal ties through purlins to create a subdiaphragm (walls with parapets) [25] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97

3.28 Usc of metal ties through purlins to create a sub diaphragm (topofwall)[2S] ..... ......... ........ .... ... ........ ............ ...... 98

3.29 Usc of metal strap attached to plywood and blocking to create a subdiaphragm [25] ............................................. 99

3.30 Use of metal strap attached to blocking to create a sub diaphragm [25] ............. 100

3.31 Detail of a direct gIuIam-to-panel connection [55] . . . . . . . . . . . . . . . . . . . . . . . . . . . .. 101

4.1 Measured force-displacement response of nailed connections .................... 102

4.2 Measured force-displacement response of plywood wall panels and diaphragms ........................................................ 103

4.3 Measured free-vibration response of plywood diaphragm [39] ................... 105

4.4 Test configuration for ABK diaphragm tests [1] ............................... 106

4.5 Measured force-displacement response of metal-deck diaphragm [1] . . . . . . . . . . . . . . 107

vi

Figure Page

4.6 Representation of plywood diaphragm as a series of nonlinear springs [11] . . . . . . . . . . lOB

4.7 Initial hysteresis rules for plywood diaphragms [11] . . . . . . . . . . . . . . . . . . . . . . . . . . . . lOB

4.B Force-deflection envelope and hysteresis rules for plywood diaphragms developed after ABK tests [4] ................................... 109

4.9 Revised force-deflection envelope and hysteresis rules for plywood diaphragm [9] .................................................. 110

4.10 Comparison of measured and calculated displacement response of diaphragm N [9] .............................................. 111

4.11 Best-fit backbone curve through load-displacement data for nailed connections [39] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 112

4.12 Comparison of measured and calculated response of wall panels . . . . . . . . . . . . . . . . . . 113

4.13 Test configuration for dynamic testing of tilt-up wall panels [6,8,7] . . . . . . . . . . . . . . . 114

4.14 Comparison of maximum response of panels 2, 3, and 4 with a rigid roof diaphragm subjected to EI Centro base motion [6,8,7] 115

4.15 Displacement and acceleration distributions in panel 3 with a rigid roof diaphragm subjected to varying intensities of the EI Centro base motion [6,8,7] 115

4.16 Displacement, acceleration, and moment distributions in panel 4 with a rigid roof diaphragm subjected to varying intensities of the EI Centro base motion [6,8,7] ..... 116

4.17 One-story tilt-up building used to develop analytical model [4] .................. 117

4.18 Calculated acceleration response of center longitudinal wall panel for lightly-damped model [4] ............................................. 119

4.19 Calculated acceleration response of center longitudinal wall panel for moderately-damped model [4] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 120

4.20 Calculated distributions of acceleration and moment along the height of the center longitudinal wall panel [4] ..................................... 121

4.21 Analytical models of one-story tilt-up building [53,54] . . . . . . . . . . . . . . . . . . . . . . . . . 122

4.22 Calculated roof-to--panel connection forces [53,54] . . . . . . . . . . . . . . . . . . . . . . . . . . . . 123

4.23 Calculated distribution of shear forces in the diaphragm [53,54] .................. 123

4.24 Analytical model of Hollister warehouse for input motion

4.25

4.26

-4.27

in the transverse direction [9] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 124

Relative displacements at the center of the diaphragm in the Hollister warehouse during the 1986 Morgan Hill earthquake [9] 125

Plan views of tilt-up buildings in Downey and Whittier [9] .......... . . . . . . . . . . . . 126

Analytical model of Downey building for input motion in the transverse direction [9] . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 127

vii

Figure Page

4.28 Analytical model of Whittier building for input motion in the transverse direction [9] .. " . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 128

5.1 Hollister warehouse ..................................................... 129

5.2 Floor plan - Hollister warehouse . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 130

5.3 Cross sections - Hollister warehouse . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 131

5.4 Typical panel reinforcement details - Hollister warehouse. . . . . . . . . . . . . . . . . . . . . . . 132

5.5 Elevations - Hollister warehouse. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 133

5.6 Locations of strong-motion instruments - Hollister warehouse ............ . . . . . . . 134

5.7 Measured ground accelerations - Hollister warehouse .......................... 135

5.8 Linear response spectra - Hollister warehouse 136

5.9 Hollister warehouse - Acceleration Histories 1984 Morgan Hill earthquake ............................................. 139

5.10 Hollister warehouse - Acceleration Histories 1986 Hollister earthquake ................................................ 141

5.11 Hollister warehouse - Acceleration Histories 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 143

5.12 Hollister warehouse - Normalized Fourier amplitude spectra 1984 Morgan Hill earthquake ............................................. 145

5.13 Hollister warehouse - Normalized Fourier amplitude spectra 1986 Hollister earthquake ................................................ 147

5.14 Hollister warehouse - Normalized Fourier amplitude spectra 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 149

5.15 Hollister warehouse - Absolute displacement histories. . . . . . . . . . . . . . . . . . . . . . . . . . 151

5.16 Hollister warehouse - Relative displacement histories 1984 Morgan Hill earthquake ............................................. 154

5.17 Hollister warehouse - Relative displacement histories 1986 Hollister earthquake ................................................ 157

5.18 Hollister warehouse - Relative displacement histories 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 160

5.19 Redlands warehouse. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 163

5.20 Aoor plan - Redlands warehouse .......................................... 164

5.21 Transverse cross section - Redlands warehouse ............................... 165

5.22 Elevations - Redlands warehouse .......................................... 166

5.23 Locations of strong-motion instruments - Redlands warehouse. . . . . . . . . . . . . . . . . . . 167

viii

Figure Page

5.24 Measured ground accelerations - Redlands warehouse. . . . . . . . . . . . . . . . . . . . . . . . . . 168

5.25 Linear response spectra - Redlands warehouse. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 169

5.26 Redlands warehouse - Acceleration Histories 1986 Palm Springs earthquake . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 170

5.27 Redlands warehouse - Acceleration Histories 1992 Landers earthquake [66] ............................................. 172

5.28 Redlands warehouse - Acceleration Histories 1992 Big Bear earthquake [37] ............................................ 174

5.29 Redlands warehouse - Normalized Fourier amplitude spectra 1986 Palm Springs earthquake . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 175

5.30 Redlands warehouse - Absolute displacement histories ......................... 177

5.31 Redlands warehouse - Relative displacement histories 1986 Palm Springs earthquake . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 178

5.32 Milpitas industrial building ............................................... 181

5.33 Boor plans - Milpitas industrial building .................................... 182

5.34 Elevations - Milpitas industrial building. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 184

5.35 Locations of strong-motion instruments - Milpitas industrial building ............. 185

5.36 Measured ground accelerations - Milpitas industrial building .................... 186

5.37

5.38

5.39

5.40

5.41

5.42

5.43

5.44

6.1

--6.2

Linear response spectra - Milpitas industrial building 187

Milpitas industrial building - Acceleration Histories 1988 Alum Rock earthquake .............................................. 189

Milpitas industrial building - Acceleration Histories 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 191

Milpitas industrial building - Normalized Fourier amplitude spectra 1988 Alum Rock earthquake .............................................. 193

Milpitas industrial building - Normalized Fourier amplitude spectra 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 195

Milpitas industrial building - Absolute displacement histories. . . . . . . . . . . . . . . . . . . . 197

Milpitas industrial building - Relative displacement histories 1988 Alum Rock earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 199

Milpitas industrial building - Relative displacement histories 1989 Lorna Prieta earthquake. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 202

Plan view of Elmendorf Warehouse indicating locations of damage [32] . . . . . . . . . . . . 205

Typical roof framing plan and elevation of fire wall in Elmendorf Warehouse [32] ............................................... 206

ix

Figure Page

6.3 Failure of a wood ledger beam in cross-grain bending. . . . . . . . . . . . . . . . . . .. . . . . . . 207

6.4 Summary of damage in tilt-up buildings during the 1971 San Fernando earthquake [41,55] ...................................... 208

6.5 Use of steel kickers to increase the strength and stability of tilt-up panels with large openings [10] .................................... 210

6.6 Skewed wall joints [35] .................................................. 211

6.7 Typical connection between wood purlin and steel ledger [10] . . . . . . . . . . . . . . . . . . . . 211

B.l Estimate of processing noise present in corrected strong-motion records [68] ....... 221

B.2 Ormsby filter used to process CS:MIP records [38] ... . . . . . . . . . . . . . . . . . . . . . . . . . . 221

B.3 Absolute displacement response ........................................... 222

B.4 Unfiltered, relative displacement response at the roof. . . . . . . . . . . . . . . . . . . . . . . . . . . 223

B.5 High-pass filter used to calculate relative displacement response ................. 224

B.6 Filtered, relative displacement response at the roof. . . . . . . . . . . . . . . . . . . . . . . . . . . . . 224

x

1. INTRODUCTION

Tilt-up construction is an efficient and economic method for constructing low-rise structures which has

become popular throughout the United States. Wall panels in tilt-up structures are cast horizontally at the

construction site, rather than in a prefabrication plant or on-site in vertical forms. Tilt-up construction

derives its name from the process of "tilting" the wall panels up into their final vertical position. Once in

place, the panels are connected to one another using pilasters, steel plates, or splicing "chord" steel at the roof

level, so as to form a structurally continuous wall system. The panels are then connected to the foundation

using cast-in-place "dowel-type" connections [80]. Finally, the roof is attached to the walls through ledger

beams attached to the panels.

Til t-up construction offers certain advantages to contractors when compared with conventional cast-in­

place walls or precast wall sections shipped to the site [3]. Tilt-up walls are usually cast horizontally on the

floor slab, therefore, form costs are low only the edges of the wall need to be formed. Further, both compac­

tion of the concrete and preparation of special surface finishes are easier when panels cast horizontally rather

than vertically. Also, transportation costs and restrictions on panel size and configuration due to vehicle li­

mitations are virtually eliminated when the panels are cast on site [70]. Generally, tilt-up walls are only han­

dled once (when they are tilted into place) during the construction process [30]. Consequently there is less

chance of damage to a til t-up wall panel, as compared to the use of conventional precast elements, which must

be handled at least twice. Tilt-up panels can also function as shear walls [70] thereby eliminating the need

for perimeter bracing and reducing overall building costs.

There are, however, several distinct disadvantages to tilt-up construction. Some of its constructibility

advantages are lost if the structure is located on a relatively confined site. Operations are difficult if the area

of the floor slab that is free of utilities and can be used to cast panels is less than 6,000 ft2, or if the width of

the building is less than 50 ft. Generally it is not cost-effective to construct these small structures using tilt-up

panels [31]. U ncongested, non-urban sites are desirable for tilt-up construction because adequate room is

needed for casting the panels and to allow movement of the crane used to erect the panels.

Tilt-up construction is cost-effective if a proposed building is one- to two-stories in height and has a

relatively simple configuration (meaning the structure is built with perpendicular corners and has large-area

offsets, if offsets are desired). Examples of simple configurations are structures with rectangular, L-shaped,

-er H-shaped plan geometries. Structures with small-area offsets, although favored for aesthetic reasons, may

increase the cost significantly and can lead to less reliable seismic response calculations.

1

1.1 Past Seismic Performance

During the 1964 Alaska earthquake and the 1971 San Fernando earthquake, typical damage in tilt-up

structures included partial collapse of roof sections due to failure of the panel-ta-roof connections and col­

lapse of wall panels following failure of the panel-ta-roof and panel-ta-panel connections [32,41,55]. As

a consequence of the structural behavior during those earthquakes, building code provisions were revised in

an effort to improve the seismic performance of tilt-up construction [75,76,77]. The response of tilt-up

construction during the 1987 Whittier Narrows and 1989 Lorna Prieta earthquakes showed that some im­

provements had been achieved. However, the degree of damage to some tilt-up buildings in the 1987 and

1989 earthquakes was still unacceptable [10,35,69].

The seismic performance of tilt-up construction is closely linked to the connection details. The designer

of tilt-up structures is faced with a difficult task of detailing each connection to provide the stiffness required

to resist service loads within permissible deflection limits while also ensuring that each connection has suffi­

cient ductility, energy-<iissipation capacity, and stability to survive seismic loads. The connections must also

accommodate the expansion and contraction of structural elements due to temperature, creep, and shrinkage

[21].

1.2 Research Needs

In the two decades since the San Fernando earthquake, considerable efforts have been made to improve

the seismic performance of tilt-up construction. Building code provisions have been revised [63]; lateral­

load tests on slender walls have been performed [52] and the results led to the adoption of new design proce­

dures for til t-up wall panels [5]; in-plane bending tests have been conducted on a variety of diaphragms repre­

senting typical roof construction [1,23,43,44,45,72,74]; analytical models have been developed to calculate

the overall response of tilt-up structures to seismic loadings [1,4,6,7,8,12,53,54]; and isolated tilt-up panels

have been subjected to simulated earthquake loading [6,7,28]. The results of these investigations have led

to an improved understanding of the behavior of tilt-up structures during strong ground motion.

However, the performance of tilt-up construction in recent earthquakes demonstrates that additional re­

search is needed if seismic damage is to be reduced to acceptable levels. There have been no tests of complete

tilt-up systems and attempts to validate analytical models oftilt-up construction using the measured response

of buildings during recent earthquakes have been limited. Physical testing has consisted onI y of tests on single

panels and isolated diaphragms. There have been few tests to measure the capacity and ductility of typical

connections used in tilt-up buildings. Finally, there are a large number of existing tilt-up structures that have

details which do not satisfy current building code regulations. Repair and rehabilitation procedures must be

developed to reduce the seismic vulnerability of these structures.

2

1.3 Objective and Scope

This report is intended to summarize existing infonnation about the seismic performance of tilt-up

construction. The scope of the report is limited to traditional, tilt-up structures in which concrete wall panels

are cast horizontally. No attempt is made to interpret the response of tilt-up frame structures.

This report is divided into seven chapters. Chapter 2 describes the influence of construction techniques

on the design of tilt-up structures. Design considerations for the wall panels, the roof diaphragm, and the

critical connections used in tilt-up construction are discussed in Chapter 3. In Chapter 4, the results of pre­

vious experimental tests of tilt-up wall panels and roof diaphragms, and previous analytical studies are sum­

marized. Acceleration histories recorded during recent earthquakes measured in three tilt-up buildings in

California are evaluated in Chapter 5. Chapter 6 summarizes the damage observed in tilt-up structures fol­

lowing the 1964 Alaska, the 1971 San Fernando, the 1987 Whittier Narrows, and the 1989 Lorna Prieta earth­

quakes. Results are summarized in Chapter 7.

1.4 Acknowledgements

This repon was completed with the assistance of graduate students Gregory 1. Lakota and Fernando S.

Fonseca. Funding for this study was provided by the National Science Foundation, under Grant

BCS--9120281, for which the cognizant NSF program official was Dr. Henry 1. Lagorio and subsequently is

Dr. M.P. Singh. Opinions and findings expressed in this report do not necessarily reflect those of the sponsor.

3

2. CONSTRUCTION OF TILT-UP STRUCTURES

Most important developments related to the design and construction of tilt-up structures may be traced

to innovations in the field. Tilt-up was first used in the early 1900's as an efficient method for fabricating

durable concrete wall panels used in military structures [13]. Contractors found that the quality of concrete

panels cast horizontally and tilted into place exceeded that of traditional cast-in-place walls. Until the 1960's,

tilt-up construction was used almost exclusively for one and two-story warehouses and industrial structures

where economical, quick construction was emphasized [24]. During the past 30 years, increased attention

has been placed on aesthetics, and the uses for tilt-up structures now include office buildings, shopping cen­

ters, and other commercial buildings. Construction techniques have been continuall y refined as the market

for tilt-up structures has continued to expand.

Typical techniques for fabricating and erecting the tilt-up wall panels and roof diaphragms are reviewed

briefl y in the following sections. The influence of these construction techniques on the design of tilt-up struc­

tures is also discussed.

2.1 Wall Panel Construction

Knowledge of fabrication and erection techniques for tilt-up panels is required to proportion the panels

effectively. Although the panel height is detennined by the architect, panel weight, and therefore width and

thickness, is often limited by the capacity of the crane used during construction. Stresses induced in the panels

d uri ng Ii fting must also be considered during design [79]. Cables are attached to connections cast in the wall

panels at the pick points (Fig. 2.1) and used by the crane to lift the panels from a horizontal to a vertical posi­

tion. Improper placement of the pick point can result in extensive cracking of the panel during tilting.

Other factors considered in design include panel fabrication, positioning of the crane at the site, and the

lifting schedule. The proposed building floor plan, panel dimensions, and the architectural treatments to the

exterior panel surface must be considered to ensure that panel fabrication and erection are completed effi­

ciently and economically [46]. For example, the outside face of the wall panels is typically cast against the

floor slab. The crane is then attached to the inside face of the panel and the panels are positioned from inside

the building (except when erecting the last few panels) [46]. This procedure prevents excessive head swing

from the top of the panel and provides excellent traction for the crane when it operates on the floor slab

(Fig. 2.2). Hence, the panel erection process influences the proportioning and fabrication of the wall panels

and the design of the building foundation.

Special attention must be paid to the design of concrete panels. Variations in width should be minimized

and attention paid to large openings in order to ensure structural integrity and maintain serviceability after

the panels have been lifted into place [10,35,78]. The openings should be located so as not to interfere with

4

the load path within the panel (Fig. 23). It is desirable that openings be placed so as not to intercept panel

joints (Fig. 2.4), because differential movements between panels can cause doors to stick or windows to break

[78]. However, piers must be of sufficient size to resist shear forces, and an arrangement with panel joints

through the openings may be more desirable based on strength considerations.

Care must also be taken to ensure the serviceability of connections within the structure. Roof framing

members should not be connected to the walls at the panel-to-panel joints in order to accommodate thermal

expansion of the panels (Fig. 2.5) [78]. Thermal effects are an important consideration for panel-ta-panel

connections, because very stiff connections can cause cracking and eventual degradation of the panels [78J.

2.2 Roof Construction

Roof construction for tilt-up and other low-rise buildings consists of the assembly of three structural ele­

ments: the framing members, the roof skin, and the fasteners. In the interest of minimizing project costs, roofs

are usually constructed to serve both as an outer protective covering for the building and as a structural dia- f" phragm to resist lateral loads. Wood and steel are often used as the roofing elements in tilt-up buildings, with

plywood-sheathed roofs being the most common form of roof construction in the western U.S. (Fig. 2.6).

Metal deck roofs are often used in the eastern U.S.

Building performance is often directly related to the choice of fasteners in the roof system [57]. Because

vertical loads on the roof of a typical low-rise building are relatively small compared with lateral loads from

wind or earthquakes, the capacity of the roof is proportional to the amount, distribution, and shearing resis- f·

tance of the fasteners.

2.2.1 Wood Diaphragms

A typical p13n view of a plywood diaphragm is shown in Fig. 2.7. Glued-laminated (gluIam) beams run

in the transverse dHectlOn of the building and are connected to the tilt-up wall panels and interior columns.

Sawn purlinssPJn bcrwecn the gluIam beams, and are overlain byaskinofplywood. Nails are used to connect

the structural members

Wood rooL dcsl!~ned to resist large lateral loads should be constructed as blocked rather than unblocked

systems [33 J. In a blocked roof, framing members are located around the entire perimeter of each 4x8-ft ply­

wood panel in the roof diaphragm (Fig. 2.8) [57]. Blocking prevents buckling of the plywood under lateral

loads. The shear capaclly of a blocked roof is 1.5 to 2 times the strength of a similar unblocked diaphragm

[57J. However, if the design shears are low, which might occur if the proposed building is not designed to

resist earthquake loads, an unblocked diaphragm is probably the most cost efficient choice.

PaneIized roof systems are often used to minimize the cost of constructing a wood diaphragm. Panel sec­

tions are fabricated on the ground from purlins and blocking members overlain with sheets of plywood

5

(Fig. 2.9). The grids are then lifted into position and connected to gIulam beams and purlins already in place.

Speed of construction is the primary advantage of this technique: an experienced crew of five workers can

fabricate up to 20,000 ft2 per day [33}.

2.2_2 Metal Deck Diaphragms

Truss girders and steel joists typically serve as the main structural members in metal deck diaphragms

(Fig. 2.10). Where shear is transferred from the diaphragm to the walls, a perimeter steel angle ledger is typi­

cally used as a shear collector, as shown in Section A-A of Fig. 2.10. In situations where diaphragm-to-wall

connections are embedded steel plates orsteel framing members, typical connections are as shown in Sections

B-B, C-C, and D-D. The metal decking typically consists of ribbed members that are either puddle-welded,

screw fastened, or pin-attached to the framing members.

Similarly to blocking in a wood diaphragm, buckling of the roof skin is prevented by installing channel

or Z- or C-type metal deck members transverse to the ribs at each panel end (Fig. 2.11). Metal decking is

typically 20-ft long and spans 2 to 3 joists. Therefore, placing these "blocking" elements every second or

third joist provides a mechanism for transferring large shears within the diaphragm.

2.2.3 Composite Diaphragms

Composite diaphragms usually comprise a metal deck diaphragm, as discussed in Section 2.2.2, overlaid

with a layer of concrete. The concrete fill acts as a global buckling mechanism for the metal deck diaphragm

skeleton.

6

3. DESIGN OF TILT-UP STRUCTURES TO RESIST LATERAL LOADS

During design, the wall panels located around the perimeter of tilt-up buildings are typically assumed

to form a box which resists the horizontal and vertical loads. The use of load-carrying members around the

perimeter of the structure increases the available area in the building by eliminating the need for internal brac­

ing. When a uniformly distributed horizontal load is applied at the roof level, the roof diaphragm acts as a

deep beam (Fig. 3.1): the interior roofing members represent the web and the perimeter chords represent the

flanges (members BF and CG in Fig. 3.1). Similarly to a plate girder, the diaphragm web is designed to resist

the in-plane shear forces and the flanges are proportioned to resist the axial forces developed due to bending

[3].

Shear forces developed in the diaphragm are transferred to the end walls and are then carried as horizontal

shear into the foundation. Chord reinforcement, located in the panels at the elevation of the diaphragm or

in the edge of the diaphragm itself, restrains the out-of-plane deflections of the tilt-up panels which result

from the in-plane deformations of the diaphragm.

Tilt-up systems represent an economical alternative to metal-clad or masonry buildings in the competi­

tive environment of low-rise commercial and industrial structures. In order to reduce the total cost of a build­

ing, the effort spent on design of tilt-up systems is usually minimized [15]. Maximum advantage is taken

of standardized design procedures and minimum building code requirements. Although this approach pro­

vides a quick and inexpensive method for proportioning tilt-up wall systems, it is only reliable for regular,

rectangular buildings with few openings in the wall panels or offsets in the perimeter. More sophisticated

analytical methods may be required for the design of buildings with irregular geometries.

Design of a tilt-up system involves the proportioning of three components: the tilt-up wall panels, the

horizontal diaphragm, and the primary connections (those between the wall panels and the diaphragm, be­

tween adjacent wall panels, and between the wall panels and the foundation). Methods used to design these

structural elements and factors affecting component performance are discussed in the following sections.

3.1 Wall Panels

The provisions of the Uniform Building Code [77] govern the design of tilt-up wall panels in most re­

gions of high seismicity in the U.S. Panels must have sufficient strength to resist moments and axial forces

due to the factored vertical and lateral loads, and must have sufficient stiffness to control deflections under

service loads. Because slender walls may develop significant out-of-plane deflections, P-Ll moments must

be considered when evaluating both panel strength and stiffness.

7

Individual wall panels are typically modelled as uniformly-loaded, simply-supported beams. (Fig. 3.2).

The midspan deflections corresponding to the cracking moment, Llcr, and nominal flexural capacity, Lin, may

be approximated as:

(3.1)

(3.2)

where Mer is the cracking moment of the panel, Mn is the nominal flexural capacity of panel, h is the distance

between supports, Ee is Young's modulus for concrete, Ig is the moment of inertia corresponding to gross sec­

tions, and ler is the moment of inertia corresponding to fully cracked sections.

The UBC limits midspan deflections under service loads, LIs, to [77]:

A < h LJ S - 150 (3.3)

where Lls is calculated assuming a linear variation of displacement between the cracking moment and the

nominal capacity:

. (Ms - Mer) ( A A) Ll s = Ll cr "'T"' ( M n _ Mer) LJ n - LJ er (3.4)

where Ms is the maximum moment in the wall under service loads.

Typically, the provisions ofUBC Section 2336 are used to determine the design lateral forces for the wall

panels. The specified lateral force for design is [77]:

(3.5)

where Fp is the lateral force resisted by the panel, Z is the seismic zone factor, I is the importance factor, Cp

is defined as 0.75 for exterior walls, and Wp is the weight of the panel. For a building located in seismic zone

4 with an importance factor of 1, the design lateral force is equal to 30% of the panel weight.

The USC deSign procedure [77] is based on the results of a series of lateral load tests conducted by the

ACI-SEASC Task Committee on Slender Walls [16]. Twelve tilt-up wall panels, with slenderness ratios

ranging from 30 to 60. were tested during this investigation. The test configuration is shown in Fig. 3.3. The

Task Committee found that previous design procedures [81], which assumed that the entire wall panel was

fully cracked, overestimated mid-panel deflections. An iterative approach for estimating deflections was

proposed where the panel midspan deflection is calculated using Eq. 3.6 based on the magnitude of the mid­

span moment under service loads and the midspan moment is determined using Eq. 3.7 which includes the

influence of P-Ll effects (Fig. 3.4).

8

M<Mer

Mer < M < My (3.6)

h2 P L1 P M=~+-P-+p L1 +~ 8 2 0 2

(3.7)

where w is the lateral load, L1 is the midspan deflection, M is the midspan moment, My is the yield moment

for the panel, Pp is the weight of the panel, Po is the applied vertical load at the top of the panel, and e is the

eccentricity of the applied load, Po.

The design procedures in the UBC and Task Committee report are based on the following assumptions:

A wall panel behaves as a uniformly-loaded, simply-supported member: maximum moments

and deflections occur at midspan and the horizontal displacement of the top of the panel relative

to the base is ignored.

The panel cross-section is constant over the height of the panel.

Many common tilt-up structural configurations do not satisfy the conditions implied in the design proce­

dures. Under seismic loading, the roof of a tilt-up structure moves relative to the base violating the assumed

simply-supported boundary conditions, concentrated loads are transferred to the panel at intermediate points

along the panel height in buildings with multiple stories, and panels are frequently cast with large openings

causing varia tions in the moment of inertia over the height ofthe panel. Proportioning of panels with openings

for seismic loads appears to be the most important of these concerns. Damage was observed in panels with

openings following the 1987 Whittier Narrows [10,35] and 1989 Lorna Prieta [69] earthquakes.

3.2 Diaphragms

A diaphragm transfers lateral forces from one lateral-load resisting system to another. In the process of

transferring these forces, the energy dissipated by the flexible diaphragm can reduce the magnitude of the

forces that the other structural elements must resist. In tilt-up structures the roof is typically the primary dia­

phragm, however, vertical diaphragms, such as those used to subdivide the structure orto compensate for wall

offsets, may also be found in tilt-up construction. In the following sections, emphasis is placed on horizontal

diaphragms.

Horizontal diaphragms in tilt-up structures are typically designed to be flexible and may sustain sizeable

~-plane deformations when subjected to lateral loads. As shown in Fig. 3.1, the horizontal shear developed

in the diaphragm is resisted by the transverse walls which musttransfer that shear to the foundations. Continu-

9

ity within the diaphragm and between the diaphragm and the transverse wall is dependent upon the strength

and deformation capacity of various connections. Four general criteria must be satisfied [73]:

Connections between adjacent sections of the roof (e.g. BIKCandIJLKinFig. 3.1) must restrain

relative horizontal deflections.

Connections between the roof framing members and the diaphragm skin must prevent buckling

of the skin.

Connections between the diaphragm and the lateral-load resisting walls must be sufficient to

transfer the diaphragm shear (e.g. connection between roof panel BIKC and wall panel ABCD

in Fig. 3.1).

Connections between the sections of the diaphragm and the diaphragm chord (e.g. chords BF and

CG in Fig. 3.1) must be sufficient to transfer the shear resulting from out-of-plane bending of

the longitudinal wall panels.

Connection details vary depending upon the materials used to construct the diaphragm. Factors influencing

the design and behavior of wood and metal-deck diaphragms are discussed in the following sections. Metal­

deck diaphragms generally provide more stability, stiffness, and resistance to environmental effects than

wood diaphragms. Experience has shown that panel-to-roof connections in tilt-up structures with metal

deck diaphragms perfonn better under severe loading than panel-to-roof connections in tilt-up structures

with plywood diaphragms. However, connections between roof elements in a metal deck do not perfonn as

well as those in plywood diaphragms under the same conditions. Regardless of connector performance, the

materials used for diaphragm construction are usually chosen to minimize the initial cost of construction.

3.2.1 Diaphragm Strength and Stiffness

The distribution of forces from the diaphragm to the tilt-up wall panels depends on the stiffness of the

diaphragm [3]. As shown in Fig. 3.5(a), forces are distributed in proportion to the tributary area supported

by the wall panels in buildings with flexible diaphragms. In contrast, forces are distributed in proportion to

the relative stiffness of the wall panels (Fig. 3.5(b)) in buildings with rigid diaphragms. ACI Committee 551

[3] classifies diaphragms according to the shear stiffness (Table 3.2) and reports that most plywood and

metal-<ieck diaphragms may be considered to be semi-flexible (Fig. 3.5(a)). Composite and concrete dia­

phragms are typically semi-rigid or rigid (Fig. 3.5(b )).

According to one school of thought, a rigid diaphragm is beneficial for lateral-load resistance because

the out-of-plane deflections in the wall panels are reduced [42,50,73]. However, flexible diaphragms and

flexible roof-to-wall connections provide a mechanism for energy dissipation which reduces the magnitude

of the forces transmitted to the perimeter walls [71].

10

(a) Wood Diaphragms

Historically, plywood diaphragms have been the most common type of diaphragm used in tilt-up

construction on the West Coast. The allowable shear strength of various plywood diaphragm configurations

is summarized in Table 3.1 [77]. The results of monotonic experimental tests [23,43,44,45,72,74] sponsored

by the American Plywood Association in the 1950's and 60's form the basis for these design provisions. The

nominal shear strength of plywood diaphragms is typically 3 to 4 times the allowable shear stress for design

[57].

The UBC requires that the in-plane deformations of the diaphragm must not exceed the deflection limits

of the supporting elements [76]. The following equation was developed from tests by Countryman [23] and

is suggested by the American Plywood Association [57] for calculating deflections in single-layered, ply­

wood sheathed diaphragms under service loads:

d= 5vL3 + vL +0094Le +27 (Ltc X) 8EAb 4Gt' n 2b

(3.8)

where d is the maximum deflection of the diaphragm, in.; v is the diaphragm shear, lb/ft; L is the diaphragm

length, ft; b is the diaphragm width, ft;A is the cross-sectional area of the chord, in.2; E is the elastic modulus,

psi; G is the shear modulus, psi; t is the effective thickness of the plywood; en is the deformation of the nails,

in.; Ltc is the slip in the individual chords, in.; and X is the distance between the support and the splice, ft.

The four components of Eq. 3.7 correspond to deflection due to diaphragm bending, deflection due to

diaphragm shear, deflection due to slip of the individual nails, and deflection due to slip at the chord splices,

respectively. Representative values of fastener slip, en, are summarized in Table 3.3. The individual chord

splice slip, Ll c , has not been quantified in any oftbe building codes or design recommendations and is usually

assumed based on data from relevant tests or engineering judgement. When the flange chord is steel reinforc­

ing bar or steel angle ledgers as in concrete tilt-up construction, the splice slip component is reduced to the

minimal effect of web-flange shear transfer between the perimeter chord and the boundary diaphragm ele­

ments [34].

(b) Metal-Deck Diaphragms

Guidelines for the design of metal-deck diaphragms are published by the Steel Deck Institute [47,48].

The results of experimental tests conducted at West Virginia University [49] form the basis for these provi-

sions.

As shown in Fig. 3.6, the panel length for metal-deck sheets typically corresponds to 2 or 3 times the

~urlin spacing. Connections between the corrugated decking and the supporting members are shown sche­

matically in Fig. 3.7. The strength of the diaphragm is typically controlled by failure of the connections in

11

the metal deck or local buckling of the metal4ieck panels [48]. Nominal diaphragm strengths for each mode

of failure are summarized in Table 3.4. Three conditions must be evaluated to determine the shear strength

of a diaphragm that is limited by the connections: failure of the structural connections between the metal deck

and the supporting members along the edge of the diaphragm (Fig. 3.8), failure of the structural and sidelap

connections (connections between adjacent metal4ieck panels) in an interior panel (Fig. 3.9), and failure of

the corner fasteners (Fig. 3.10) [48].

The deflection of a metal-deck diaphragm is larger than the deflection of a comparable, continuous plate

of uniform thickness because the metal-deck diaphragm is made from individual sheets of finite width that

are joined at discrete points along the edges [49]. Stress fields are discontinuous within the metal4ieck dia­

phragm due to these gaps leading to larger displacements. The corrugations in the metal deck are susceptible

to warping at the ends of the panels, which also increases the deformations.

Studies of metal-deck diaphragms [49] have identified four phenomenon that must be considered when

calcula ring diaphragm deflections: shear displacement of the diaphragm, end warping of the deck panels, slip

at the interior sidelap connections, and slip of the supporting system of purlins and edge beams. An underly­

ing assumption in this approach is that the shear stiffness of the metal deck is small compared with the flexural _ ... ££~ ___ ~_~_£_~ ___ 1 ___ L ___ ~_£ ___ ... : __________ :...1 ___ ...1 rA"7' :Slllln~:s. 1 Ht::lt::J.Ult::, UllIY :SHt::a! Ut::1UllllCilJ.Ull:s Cilt:: \,;Ull:sIUt::lt::U l't I J.

The displacement due to pure shear (Fig. 3.11(a)) may be calculated as [48]:

L1 = (P a) 2 (1 + v) ~ S LEt d

(3.9)

where L1s is the pure shear displacement, in.; P is the applied diaphragm load, kip; a is the diaphragm width,

ft; L is the diaphragm length, ft; v is Poisson's Ratio; E is Young's Modulus, ksi; t is the thickness of the deck

element, in.; d is the corrugation pitch, in.; and s is the developed flute width, in. (Fig. 3.7(b)).

Unless the corrugated deck elements are restrained, an extra component of deflection results from warp­

ing (Fig. 3.11 (b)). This component of displacement is derived from treating the corrugation as a beam on an

elastic foundation and leads to rather cumbersome expressions for warping displacement. However, warping

constants, D n , are tabulated in the Steel Deck Institute manual [48] for common deck panels so detailed cal-

culations are unnecessary.

The influence of fastener and support slip are included in the coefficient C [48]:

(3.10)

where Sf is the structural connection flexibility, in./kip; Ss is the sidelap connection flexibility, in.!kip. The

terms aI, a2, 11s, and I1p are related to the number and arrangement of the fasteners, and are defined in

12

Table 3.4. Fastener strengths and flexibilities are defmed in Table 3.5 and discussed in Section 3.2.2. The

slip coefficient, C, decreases with an increase in the number and stiffness of the fasteners, or with an increase

in the thickness of the metal deck.

The shear displacement, warping constant, and slip coefficient are combined to give the total deflection

of a diaphragm subjected to a 10adP as follows [48]:

Lt t = Lts + (¢Dn + C) :t~ (3.11)

whereLt r is the diaphragm deflection, in. and the factor¢ reflects the influence of purl in spacing on warping.

Values of ¢ are tabulated in Ref. 48 and range from 1.0 for deck sheets that span over two or three purlins

to 0.58 for deck sheets that span over eight purlins.

The shear stiffness, expressed in kip/in., of a metal deck diaphragm may be calculated as [48]:

G' E t (3.12)

2.6 J + ¢Dn + C

(c) Composite Diaphragms

When additional stiffness is required in a metal-deck diaphragm, the decking is often over-lain with con­

crete. In concrete composite diaphragms, the shear strength is dependent upon the type of concrete used.

Nominal strengths are presented in Table 3.6 for composite diaphragms with structural and insulating con-

cretes.

The shear stiffness of concrete composite decks may be derived from Eq. 3.12. The concrete fill prevents

warping of the corrugated elements and the stiffness of the concrete fill must be considered [48]:

G' = E t + 3.5 d f.F,)O.7 2.6 2+ C e Ve

(3.13)

where de is the depth of the concrete cover above the top corrugations, in. and fe' is the specified compressive

strength of the concrete, psi.

3.2.2 Diaphragm Fasteners

(a) Wood Diaphragms

The fasteners used within the framing elements of a wood roof diaphragm can be broken down into three

categories [71]: nails, staples, and adhesives. Nails are by far the most common mechanical fastener in wood

~aphragm construction and are produced with either plain or mechanically deformed shanks. Nail pull-out

was a common cause of roof failures in low-rise buildings in the 1971 San Fernando earthquake [55] and at

13

that time most nails had plain shanks. By inducing deformation in the nail shank, an increase in the nail's

pull-out resistance occurs, along with a decrease in the required depth of penetration for the nail to achieve

resistance. Therefore deformed shank nails are now recommended for use in high seismic zones.

The pull-out strength of nails with lengths between % and 11/8 in. and various deformed shanks are

compared with 6d common nails in Fig. 3.12 [20]. In general, nails with helical threads provide more strength

and create a stiffer connection than nails with annular threads [71]. The size of the nail head is also important

[71]. A large nail head gives a larger bearing area and therefore more resistance against the nail pulling

through the diaphragm skin. Splitting of the plywood skin was also a common mode of roof failure in the

1971 San Fernando earthquake [55].

Staples are the second most common type of fasteners in plywood diaphragms. Staples are not as variable

in geometry as nails. They have such general classifications as "slender" or "thin" and "stout" or "fat" [71].

It is considered better practice to use many slender staples than a few stout staples because slender staples

cause less splitting of the plywood and can be driven with lighter tools. Staples can be used in place of nails

in order to control plywood splitting or when a small fastener spacing is required.

Two types of adhesives are used in diaphragm construction: rigid adhesives and mastic adhesives. Rigid

adhesives use staples or nails only to hold the wood in place until the adhesive has set [71]. Mastic adhesives,

however, resist service loads with the help of fasteners, and at large loadings the load is carried solely by the

fasteners while the mastic adhesive acts to reduce the amplitude of the deflection [71]. Although adhesives

provide strong and durable connections, their use is not widespread because of their relatively high cost.

(b) Metal- Deck Diaphragms

The fasteners used for connections within metal-deck diaphragms can be divided into three categories:

welds, screws, and power-driven pins [47]. Each type of fastener exhibits higher strength and stiffness when

the connection is between the metal deck and a structural member (structural connections) than when the con­

nection is between deck sheets (sidelap or stitch connections). The fastener strength and flexibility of struc­

tural connections will be denoted as Qf and Sf, while the fastener strength and flexibility of sidelap connections

will be denoted as Qs and Ss. The strength and flexibility of common connectors are presented in Table 3.4

[48].

Welded connections are the most common in metal decks due to the speed of construction [47]. The

strength of puddle welds without washers depends on the thickness of the metal deck, the diameter of the weld,

and the strength of the base material. Problems can occur if the amperage is too high during welding leading

to bum-through of the upper layer of deck, or if the amperage is too low there may be improper fusion into

the bottom layer [47]. When thin deck sheets (less than 0.028 in.) are used a the diaphragm, weld washers

14

are recommended because they act as a heat sink and control the size of the hole [48]. The strength of a welded

connection with weld washers is related to the thickness of the deck, the diameter of the hole in the washer,

and the electrode strength. The strength of welded sidelap connections is taken to be 75% of the comparable

strength of structural welded connections. Welded connection flexibility is usually small compared with the

flexibility of other types of fasteners because the slip around the welds is relatively small and limited primarily

to distortion of the deck element around the weld [48].

The equations for strength and flexibility of screwed connections are based on experimental data using

No. 12 and No. 14 screws and apply to both self-drilling and self-tapping types of screws [47]. For structural

connections, the strength is controlled by the thickness and yield stress of the decking. Strength depends on

deck thickness and screw diameter for sidelap connections.

Power-driven pins are shafts, which may be slightly tapered, that are driven through the deck elements.

Holes are not pre-drilled. The strength of structural connections depends on the type of pin and the thickness

of the deck, while the strength of sidelap connections depends only on the thickness of the deck.

3.2.3 Design of Non- Rectangular Diaphragms and Diaphragms with Openings

Due to the inherent flexibility of roof diaphragms in tilt-up buildings, large deflections are expected un­

der la teralloads. Conseq uentl y, tilt-up buildings with irregular plans may experience large incompatibilities

in displacements between adjoining sections of diaphragm near reentrant comers or near stairwells attached

to the roof (Fig. 3.13(a) and 3.14(a)). The concentration of displacements generates large shear forces and

has the potential to cause structural damage [14]. In order to resist these shear forces, the diaphragm must

be designed with structural members that ~'collect" the force and transfer it to the vertical wall panels. These

collector elements, called drag struts, receive the diaphragm force in shear and then "drag" the force back to

the vertical elements by anchorage [14]. Figures 3.13(b) and3.14(b) show that the addition of drag struts has

divided the diaphragm Into smaller rectangular diaphragms [14], and the displacements at reentrant corners

and staiIWells are compatIble with the surrounding structural elements.

Rather than proVide structural elements to resist the high shear forces developed at the reentrant corners

and stairwells, efforts can he made to eliminate these forces altogether by avoiding displacement incompati­

bilities at reentrant corner.; and stairwells as shown in Fig. 3.13(c) and 3.14(c). By not attaching the wall pan­

els to the roof in these areas, displacement incompatibilities at critical locations no longer exist. The unat­

tached walls and staiIWells will deflect as solitary units without affecting the global diaphragm response.

_ Smaller rectangular diaphragms within a global non-rectangular diaphragm are called "subdiaphragms,"

and are subject to the same code provisions and constraints as a typical diaphragm [34]. Specifically, all sub-

15

diaphragms must be sized such that they conform to the maximUm. diaphragm aspect ratios given in Table

No. 25-1 of the UBC (Table 3.7) [77].

High local shears that may be present in a diaphragm with openings must also be considered. Local shears

are typically considered by analyzing the diaphragm as a Vierendeel truss [74] as shown in Fig. 3.15. The

shear and bending forces along and across critical sections of the diaphragm must be calculated to determine

if the surrounding framing members have sufficient capacity to resist the amplified bending and shear forces

located in the vicinity of the opening. It is important to provide blocking members around the perimeter of

all openings and to provide a positive direct connection between the blocking and the surrounding framing

elements.

3.3 Connections in Tilt- Up Systems

Selecting appropriate connections is the most important aspect of designing tilt-up buildings to resist

earthquake loads. The capacity and ductility of the connections will determine whether or not a structure per­

fonus satisfactorily during an earthquake. The connections in a tilt-up structure can be divided into three

types: panel-ta-foundation connections; panel-ta-panel connections; and panel-ta-roof connections.

3.3.1 Panel-to-Foundation Connections

Typical panel-ta-foundation connections are shown in Figs. 3.16 and 3.17 [40,55,80]. The Uniform

Building Code [77] requires that connections between precast walls and the supporting member must resist

a tensile force in lb of at least 50*Ag whereAg is the cross-sectional area of the wall in in? Most designers

do not provide a physical connection between the tilt-up panel and the foundation as specified by the UBC

because in many instances the weight of the panel counteracts any uplift forces. Rather, a dowel connection

between the panel and floor slab is typically provided. If lateral loads are expected to produce uplift forces

in the til t-u p panels, then designers typically will provide one of the following types of connections: (1) physi­

cal connection between tilt-up panel and foundation (which in most instances is #4 bars at 4 ft on center),

or (2) sufficient panel-ta-panel connections to constrain the in-plane walls to behave as one monolithic shear

walL This monolithic behavior increases the resisting moment of the shear wall which counteracts the applied

moment from the lateral forces producing uplift. Currently, many engineers are trying to remove UBC provi­

sion 2615(i)3B which would allow designers to decide if panel-ta-foundation connections are needed.

3.3.2 Panel-to-Panel Connections

Panel-ta-panel connections have changed significantly during the past 30 years. In the 1960's continu­

ous, cast-in-place pilasters were often used to connect panels (Fig. 3.18(a». Another common detail was

to provide connections at six to eight foot intervals along the height of the wall (Fig. 3.18(b ». However, in

recent construction, a single continuous chord is typically provided at the roof level around the perimeter of

16

the building [10,80], with no other connections between panels, except at the comers of the building (Fig.

3.19). The perimeter chord provides a restraint that holds the tilt-up building together, so that it functions

as a unit under seismic loading. Designers recommend restricting panel-to-panel connections to the single

continuous chord in order to eliminate degradation of connections due to temperature and shrinkage effects.

Also, some designers believe that the increased amount of structural damping due to fewer panel-to-panel

connections more than compensates for the decreased lateral resistance that results from using less connec­

tions [80].

Pilaster connections are not common in new construction because the pilasters produce stress concentra­

tions at the connected panel edges, as a result of out-of-plane deflections, and they restrain movement due

to shrinkage and temperature effects [10].

3.3.3 Panel-to-Roof Connections

In tilt-up construction, the critical connection for seismic loading is usually the connection between the

roof diaphragm and the concrete tilt-up wall paneL Panel-to-roof connections must be designed to resist

forces normal and parallel to the plane of the paneL Inadequacies of these connection have been the cause

for many partial roof and panel collapses during the past three decades. The 1964 Alaska earthquake and the

1971 San Fernando earthquake gave clear evidence that the use of the popular wood ledger connection, as

shown in Figs. 3.20 and 3.21 [77], must be restricted to regions of low seismic risk and should be replaced

by some type of joist anchor in high seismic risk zones (Fig. 3.22) [25].

Wood ledger connections were found to be susceptible to three failure mechanisms: the ledger was placed

in cross grain bending by seismic lateral loads which resulted in the wood ledger splitting along the bolt line;

the bearing stresses of the nails in the plywood-to-ledger connection caused the nails to shear through the

plywood; and the force on the nails resulting from tension in the plywood overcame the pull-out resistance

of the ledger.

In 1976, the UBC [75] introduced four new code provisions to avoid these problems (Fig. 3.23). Those

provisions are reproduced in Appendix A. Section 2310 specifies a direct connection between the wall and

diaphragm capable of resisting at least 200 lb per lineal foot of wall. Section 2312G)2D requires continuous

ties between diaphragm chords to anchor these forces. Section 2312(j)3A prohibits the use of toe nails, nails

su bjected to withdrawal, or wood framing used in cross-grain bending or cross-grain tension in all seismic

zones except zone 1. Section 2312(j)3C draws attention to the need to have exterior panels able to accommo­

d~te structural movements resulting from both lateral forces and temperature changes. These provisions have

remained essentially unchanged through the 1991 edition of the UBC [77].

17

As specified in Section 2310 of the UBC, the panel-to-roof connection must resist a minimum. anchorage

force of 200 lb per lineal foot of wall. This provision rarely controls in tilt-up construction, however. Consid­

er, for example, a tilt-up warehouse with a plywood roof diaphragm constructed in California during the early

1970's. The panel height is likely to be greater than 17 ft. The design force for the panel, Fp , is calculated

using Eq. 3.5 where the zone factor is taken to be 0.4, the importance factor is taken to be 1.0, and Cp is taken

to be 0.75 [77]. This leads to a design lateral force for the panel ofO.3Wp where Wp is the weight of the panel.

However, the UBC states that when designing the connections in the middle half of the building, Cp must be

multiplied by 1.5 for flexible diaphragms. If the tilt-up panel is modelled as a simply-supported member,

then the connection force between the foundation and the panel is the same magnitude as the connection force

between the panel and the diaphragm, 0.225Wp. Assuming a minimum panel thickness of 51h in., the weight

of the panel is 1170 lb/ft, and the connection between the panel and the diaphragm must be designed to resist

265 lb/ft, which is greater than the specified minimum strength. Therefore, the minimum anchorage force

of 200 lb/ft should be considered to be a lower bound in tilt-up construction.

The unsatisfactory performance of many panel-to-roof connections indicated that continuous ties were

needed between diaphragm chords to distribute horizontal forces within the diaphragm and that direct, posi­

tive connections were needed for anchorage of the diaphragm to the panels. Because the use of continuous

ties from one end of a diaphragm to the other was highly inefficient, the concept of subdiaphragms was

introduced (Fig. 3.23). A series of small "diaphragms" within the total diaphragm were used to transfer an­

chorage forces to the wall from the diaphragm interior. For the 16x64-ft subdiaphragm EFGH in Fig. 3.23,

the longitudinal purlins serve as ties, if the purlins are connected directly to the wall (Fig. 3.22, 3.24, and 3.25)

and are made continuous over the interior glulam beams (Fig. 3.26). If, however, the purl ins do not frame

into the side walls. as is the case for some existing construction, then a retrofit can be made by introducing

ties into the 8x 16--ftsubdiaphragm HIJK(Fig. 3.23) by metal straps or rods, as shown in Figs. 3.28, 3.29, 3.30,

and 3.27, to create the continuous tie connection. In the transverse direction, the continuous tie can be pro­

vided by connecnng the glulam beams directly to the tilt-up wall as shown in Fig. 3.31. The subdiaphragm

concept, therefore. Simultaneously fulfills the provisions for continuous ties between diaphragm chords and

for closely-spaced tiCS for walls with negligible bending resistance between anchors.

Several varienes of "direct" connections of plywood sheathing and roof joists to the wall panel reinforce­

ment, as seen in Figs. 3.22, 3.28, 3.29, 3.30, 3.31, have been used. The advantages and disadvantages of each

of those connections are listed below each figure [25]. Connections used to retrofit the wood ledger in

Fig. 3.20 and 3.21 to provide better anchorage of the framing members to the wall panel by providing a "direct

connection" are shown in Figs. 3.24, 3.25, 3.27. Such schemes were used to repair and upgrade roof-to-wall

connections after the 1971 San Fernando and 1987 Whittier Narrows earthquakes.

18

After the 1971 San Fernando Earthquake, building codes also placed limits on plywood thickness [75].

For the details shown in Fig. 3.23, plywood was to be at least sh6-in. thick for sub-purlins (studs) placed

16 in. on center and at least 3/8-in. thick for studs placed 24 in. on center. These limits on plywood thickness

were implemented to reduce the likelihood of nail shearing through the plywood.

The influence of shrinkage in wood diaphragm elements must be considered when evaluating the durabil­

ity of panel-to-roof connections. Shrinkage in sawn lumber framing members may be approximated as

1132 in. shrinkage per 1 in. of width or depth as the member progresses from the green to the dry state [25].

Glulam beams can be expected to shrink Ih6 in. per foot of depth for every 3% moisture loss. This restraint

could lead to pull-out of the fasteners connecting the embedded strap to the plywood, or degradation of the

ledger due to cross-grain tension splitting along the bolt line.

3.4 Summary

Typical design procedures for tilt-up construction treat a building as a series of individual components,

rather than a structural system. The diaphragm is designed as a simply-supported shear beam to transfer later­

al forces into the end wails, and the wall panels are designed as slender columns, pinned at both ends, to resist

gravity and lateral loads. Code-specified forces are often used to design the critical connections.

19

4. SUMMARY OF PREVIOUS INVESTIGATIONS OF THE SEISMIC

BEHAVIOR OF TILT-UP CONSTRUCTION

Since the 1971 San Fernando earthquake, engineers throughout the U.S. have studied the seismic re­

sponse of many types of buildings, and developed design provisions to improve the performance of new

construction. In Southern California, emphasis was placed on reducing the seismic risk of unreinforced ma­

sonry and tilt-up buildings. These types of construction have sustained significant structural damage during

recent earthquakes and represent a large portion of the inventory of existing, low-rise, industrial buildings.

Much of the work related to tilt-up construction has been conducted by researchers at Agbabian

Associates [4,5,6,8,7,9,10,11,12,27,28], where analytical modelling procedures have been developed

based on the results of experimental tests. Analytical models of tilt-up systems have also been developed at

Dames and Moore [53,54].

Experimental tests of diaphragms subjected to cyclic loads have been conducted by Agbabian!Barnes/

Kariotis (ABK) [1] and researchers at the University of British Columbia [26], the University of California

[84,85], Stanford University [83], and Washington State University [39]. The ABK tests represent the most

extensive investigation with tests of full-scale plywood, wood-sheathed, and metal deck diaphragms. How­

ever, a detailed description of the results has not been published [2].

The results of these experimental and analytical studies are summarized in this chapter. Diaphragms are

discussed in Section 4.1, tilt-up wall panels are discussed in Section 4.2, and analytical models for complete

tilt-up systems are summarized in Section 4.3.

4.1 Cyclic Response of Diaphragms

During a design-level earthquake, the types of roof diaphragms used in most tilt-up structures are ex­

pected to experience nonlinear response. The nature of this response is extremely sensitive to the types of

connections used within the diaphragm and to the actual material properties of the diaphragm components,

which are highly variable. Most of the experimental research to date has focused on the behavior of wood

diaphragms, because wood diaphragms have been used almost exclusively in Southern California and the

Pacific Northwest during the past 20 years. The results of five experimental investigations of the cyclic re­

sponse of wood diaphragms and panels are summarized in Section 4.1.1. Data from individual connections

and complete diaphragms are presented. The iimited data from cyclic tests of metal~eck diaphragms are

described in Section 4.1.2. Methods for modelling diaphragms are discussed in Section 4.1.3. Analytical

representations of the diaphragms range from using several nonlinear spring elements to special-purpose fi­

nite~lement models with individual nail elements.

20

This section is not intended to summarize all experimental and analytical work related to diaphragms.

Only investigations that involve cyclic loading are discussed. The paper by Peterson [56] contains a compre­

hensive review of the literature related to wood diaphragms.

4.1.1 Experimental Tests of Wood Panels and Diaphragms

The first phase of many investigations of the behavior of plywood diaphragms and panels is devoted to

understanding the response of the individual nailed connections. The measured response of nails connecting

plywood and framing members is shown in Fig. 4.1. The data shown in Fig. 4.1(a) were obtained by cycling

the connection to a given force level [39], while the connection shown in Fig. 4.1(b) was cycled between given

displacement levels [26]. In both cases, the stiffness of the connection decreased as the amplitude of the dis­

placement increased, and the connection exhibited a region of extremely low stiffness as the applied load

passed through zero. Once the connection was pushed into the nonlinear region of response, specimens would

experience larger displacements when pushed to the same nominal force level (Fig. 4.1(a)) and specimens

pushed to a specified displacement would resist lower forces as the number of loading cycles increased (Fig.

4.I(b)).

Five experimental investigations in which complete diaphragms or panels were subjected to load rever­

sals are summarized in Table 4.1. Young and Medearis [83] ,Zacher and Gray [84,85], Itani and Falk [39],

and Dolan [26] evaluated the response of plywood, gypsum board, and waferboard panels, while ABK [1]

and ltani and Falk [39] investigated the behavior of plywood, lumber-sheathed, and gypsum board dia­

phragms. The general shape of the measured hysteretic response of complete diaphragms closely resembles

the behavior of the individual connections (Fig. 4.2). The force-displacement curves are pinched, diaphragm

stiffness decreases wi th increasing displacement, and diaphragm stiffness decreases as the number of inelastic

loading cycles at a constant displacement or force level increases.

Young and Medearis [83] found that 20 cycles at the nominal design level did not influence the capacity

of the wall panel nor the nonlinear force-displacement response. This observation was confirmed in smal1-

scale panel tests by Yasumura and Sugiyama [82] where panels were subjected to 50 cycles at ±60% of the

strength of nominally identical specimens tested monotonically. Accumulated damage was observed in tests

when the panels were subjected to loading cycles of ±80% of the capacity [82].

Young and Medearis [83] also estimated viscous damping factors from their test results. During load

cycles at the nominal design level, damping values of 0.07 and 0.10 were calculated for panels with one and

two layers of plywood, respectively. Polensek [58] identified damping factors between 0.07 and 0.11 from

low-amplitude, free-vibration tests of plywood floor systems. Itani and Falk [39] also estimated damping

coefficients from free-vibration tests (Fig. 4.3). At a displacement level of 0.1 in., damping factors in the

plywood diaphragm specimens were between 0.1 and 0.15. Equivalent damping factors increased to more

21

than 0.20 when the displacement level was increased to 0.6 in. As indicated in Fig. 4.3(b), the displacement

levels used in both free-vibration tests did not cause significant nonlinear response in the diaphragms.

The ABK diaphragm tests were designed to evaluate a number of prastical concerns such as the influence

of blocking, roofing materials, and retrofit nailing on the response of diaphragms [1]. Table 4.2 contains a

summary of the primary experimental variables in these tests. Each diaphragm was subjected to a series of

quasi-static load reversals and earthquake motions in real time. Schematic drawings of the diaphragm test

specimens are shown in Fig. 4.4. Comparisons of the quasi-static and dynamic response of diaphragm Dare

shown in Fig. 4.2( c) and (d). The average initial stiffness inferred from the low-amplitude quasi-static tests

of all diaphragms are reported in Table 4.2.

Data obtained during the dynamic, earthquake simulations indicate that diaphragm response remains

nearly linear up to accelerations of approximately O.lg [1]. Beyond O.lg, the nonlinear characteristics of the

diaphragm may be observed. Researchers noted that roofing material initially added stiffness to the dia­

phragm, however, the roofing material separated from the diaphragm when the accelerations reached approx­

imately 0.2g [1].

Zacher and Gray [84,85] compared the behavior of panels connected with nails and staples, and eva­

luated the influence of over-driving the fasteners. The results indicated that stapled panels do behave satis­

factorily, however the nailed panels were able to resist larger displacements before failure. Panels with nails

over-driven by 1/8" failed in a brittle manner at displacements that were less than 75% of the displacement

capacity of similar panels in which the nail heads did not break the plywood veneer. The displacement capac­

ity of panels with staples was also reduced when the staples were over-driven, however, the failure mode was

not as abrupt as observed for the nailed connections.

4.1.2 Experimental Tests of Metal-Deck Diaphragms

As indicated in Table 4.2, metal-deck diaphragms were also tested as part of the ABK investigation [1].

The measured response of diaphragm R is shown in Fig. 4.5. Response during the quasi-static tests

(Fig. 4.5(a)) is similar to that of plywood diaphragms. The metal-deck diaphragm displayed a pinched hys­

teresis curve and the effec tive stiffness decreased with increasing displacement. It is di ffi cui t to make concI u­

sions about the cyclic force-displacement response of metal-deck diaphragms from the dynamic data

(Fig. 4.5 (b)).

4.1.3 Analytical Models of Diaphragms

The measured data described in Section 4.1.1 form the basis for the analytical representations of dia­

phragms discussed in this section. In all cases, the nonlinear features of the analytical models were scaled

22

from available experimental data. No procedures are available to estimate the nonlinear response of a dia­

phragm given the nominal design properties discussed in Chapter 3.

In the late 1970's, Adham and Ewing [11] developed an analytical model where eight inelastic spring and

damper assemblies were used to model wood diaphragms (Fig. 4.6). Diaphragm properties scaled from the

monotonic tests performed by TIssel [74] were combined with the linear hysteresis rules shown in Fig. 4.7.

The calculated frequencies of plywood and lumber sheathed diaphragms ranged from 2.8 to 11.5 sec. which

is considerably larger than those inferred by Blume and Rea from full-scale, non-destructive tests of wood

dia phragms in school buidlings [17, 18,62].

Following the ABK tests [1], Adham [4] refined the hysteresis model for plywood diaphragms. A se­

cond-order curve was selected to model the force-deflection envelope of the diaphragm (Fig. 4.8(a)):

F(e) = Fu e

~: + lei (4.1)

where F(e) is the force in the spring, e is the deformation of the spring, Fu represents the strength of the dia­

phragm, and K 1 is the initial diaphragm stiffness. When the diaphragm is subj ected to cyclic loading, the hys­

teresis rules defined in Fig. 4.8(b) are used to control the response. Based on the observed response of the

ABK diaphragms, the unloading stiffness, K 2, was assumed to be equal to the initial diaphragm stiffness, K 1,

and the force level used to define slip at low applied loads, F 1, was taken to be ten percent of the strength of

the diaphragm, Fu' Values of the critical parameters, Kr, K2, Fu, andFr, for an arbitrary plywood diaphragm

are calculated from the experimental data using the scaling rules listed below [4]:

(4.2)

F· = QF~ L D' l

(4.3)

where L is the length and D is the width of the diaphragm section under consideration, and the following pa­

rameters were taken from the ABK test results [1]:

L'

D'

K' 1

K' 2

F' u

F' 1

= = =

= =

length of diaphragm section in test = 20 ft

width of diaphragm section in test = 20 ft

observed initial stiffness of diaphragm in test = 324 kip/ft

observed reloading stiffness of diaphragm in test = 324 kip/ft

observed strength of diaphragm in test = 32 kip

observed strength at which diaphragm stiffness increases during cycling = 3.2 kip

During the NSF-sponsored TCCMAR program, the ABK tests [1] were re-evaluated and the diaphragm

hysteresis model was revised to include strength degradation at large displacements and variation of the un-

23

loading stiffness with the level of deformation [9,36,29]. The force-{!eflection envelope and hysteresis rules

for the revised model are shown in Fig. 4.9. The unloading stiffness, Ku, was defined as:

(4.4)

where ey represents the yield deformation and is defined as 113 Fu/Kl' emax is the maximum deformation of

the diaphragm during previous loading cycles, and y is assumed to be 0.2 for wood diaphragms. Viscous

damping was ignored in the revised diaphragm model (Fig. 4.6), the hysteretic damping waS considered to

be sufficient. Calculated and measured displacement response of diaphragm N are compared in Fig. 4.10

during one of the later earthquake simulations [9].

Itani and Falk [39] and Dolan [26] developed special-purpose finite-element codes to analyze the re­

sponse of plywood diaphragms and panels. The researchers used similar modelling techniques: framing

members were represented using linear beam elements, the plywood sheathing was modelled with linear

plane-stress or shell elements, and nonlinear spring elements were used to model the nailed connections be­

tween the framing and sheathing. Special gap elements were used in both investigations to allow adjacent

sheets of plywood to separate, but not overlap. Dolan [26] used similar bi-linear elements to represent the

connections between framing members, while Itani and Falk [39] used hinged connections to attach all fram­

ing members.

The general nature of the calculated response in both investigations was governed by the choice of nonlin­

ear nail element. Data from connection tests (Fig. 4.1) were used to develop envelope curves for the nailed

connections (Fig. 4.11). Approximately 100 connections were tested in each investigation. Itani and Falk

[39] chose a power curve to represent the data,

Fcoll = alL:] Q2 (4.5)

while Dolan [26] used a three-parameter model,

(4.6)

where Fcon is the force resisted by the connection, L1 is the displacement of the connection, and al, a2, Po,

Ko, and K2 are constants whose val ues were determined from the experimental data using a curve fitting tech­

nique.

- Both finite--element models were used successfully by the researchers to reproduce their experimental

results (Fig. 4.12).

24

4.2 Cyclic Response of Tilt- Up Wall Panels

Agbabian Associates conducted a series of dynamic tests on tilt-up wall panels in the early 1980's

[6,8, 7,28]. The experimental setup for these tests is shown in Fig. 4.13. Three wall panels were tested. Key

parameters of the experimental program are summarized in Table 4.3. Two load cells, 1 displacement trans­

ducer, and 9 velocity transducers were used to measure the response of the panels. Displacements and accel­

erations along the height of the panel were later calculated from the velocity data.

These experiments were closely linked to the analytical modeled described in Section 4.3. Researchers

calculated the transverse response of a representative tilt-up building at the top of the longitudinal wall panels

using the 1940 El Centro and 1971 Castaic (San Fernando) earthquake records. This calculated response was

then used as the input motion at the top of the wall panel, and the ground motion was used to drive the base

of the panel (Fig. 4.13). Response was calculated for tilt-up buildings with rigid and flexible diaphragms.

Each wall panel was subjected to a series of 9 or 10 earthquake simulations. The effective peak ground

acceleration was increased from 0.2g to O.4g in the later tests. By the end of the testing sequence, all panels

had experienced inelastic response. The El Centro ground motion, combined with a rigid diaphragm, proved

to be the most severe test of the panels. The maximum acceleration and displacement response of the panels,

inferred from the measured velocity data, is shown in Fig. 4.14. The amplitude of the displacements increased

as the panels were subj ected to more loading cycles and sustained structural damage (Fig. 4.15).

The distributions of accelerations and displacements closely resembled the first mode shape of a panel

that is pinned at both ends (Fig. 4.14). The researchers, therefore, concluded that the response of the panel

a t mid-heigh t should govern the design, and that using fully cracked sections for panel design was a conserva­

tive assumption [6,8, 7]. Distributions of moments were also calculated along the panel height (Fig. 4.16).

Although the distributioil of moments did not correspond to the expected first mode shape, the researchers

concl uded that design procedures for walls were appropriate because the magnitude of the calculated mo­

ments was less than those calculated using the ACI-SEASC recommendations for slender walls [81].

4.3 Analytical Models of Tilt-Up Systems

In the early 1980's, Adham [4] developed an analytical model for tilt-up construction that was based on

an earlier representation of unreinforced masonry buildings [11]. Considering the representative tilt-up

building shown in Fig. 4.17, the following assumptions were made:

Earthquake motion in the transverse direction of the building was considered to be critical.

Only half of the building was analyzed due to symmetry (Fig. 4.17(c».

The roof diaphragm was modelled as a deep shear beam using four inelastic springs (Fig.

4.17( d». Hysteresis rules for the inelastic springs are defined in Fig. 4.8.

25

• The transverse wall panels were assumed to be rigid. Therefore, ground motion was assumed

to be transmitted to the roof without amplification at the end of the building (Fig. 4.17( d)).

• The longitudinal wall panels were assumed to deform primarily in out-of-plane bending. Linear

beam elements were used to represent these panels (Fig. 4.17(d)).

The response of the two longitudinal walls was assumed to be the same. Therefore, a single set

of beam elements could be used to model the longitudinal walls (Fig. 4.17( e)).

A total of 23 nodes, 4 inelastic springs, and 18 linear beam elements were used to model the 300' by 150'

warehouse shown in Fig. 4.17(a) [6]. The model did not include any type of connection between adjacent

longitudinal wall panels. A viscous damping factor of 5% was used for the beam elements, and two damping

factors (0.07% and 10%) were used for the nonlinear springs. The model was subjected to a scaled version

of the N69W component of the motion recorded at Castaic during the 1971 San Fernando earthquake.

Calculated acceleration response at the top and mid-height of the center longitudinal wall panel is shown

in Fig. 4.18 and 4.19 for the lightly-damped and moderately-damped models, respectively. In both case$,

the am pli tude of the response is greater at mid-height of the panel than at the top. The amplitude of the accel­

erations at the roof exceeded those at the ground by a factor of 1.4 for the moderately-damped model and 2.6

for the ligh tl y-dam ped model. This result implies that the connection forces between the wall panels and the

roof exceed those between the wall panels and the foundation. The calculated acceleration response of the

center of the roof was used as the driving function at the top of the panel for the experimental tests described

in Section 4.2 [6,8,7,28].

Distributions of the calculated accelerations and moments along the height of the center longitudinal wall

panel are shown in Fig. 4.20. Unlike the experimental data, the distribution of calculated moments resembled

the first mode shape of a pinned-pinned beam.

In the late 1980 's, researchers at Dames and Moore used a similar model to represent the seismic response

of tilt-up buildIngs 153.54]. The idealized building and analytical models for linear and nonlinear analyses

are shown in FIg 4.21. The transverse walls were assumed to be rigid and the longitudinal walls were mod­

elled using beam clements. The diaphragm was assumed to deform in shear. Initially, linear elements were

used to model the dlaphragm and longitudinal wall panels. Bi-linear models were later adopted to evaluate

the influence of member nonlinearity on structural response.

A 200 I by 200 I building was subj ected to the S69E component of the 1952 Taft ground motion. The initial

stiffness of the diaphragm was varied such that the natural period of the diaphragm ranged form 0.25 to 2.0

sec. Viscous damping factors of 5% and 10% were used. The calculated anchorage forces between the dia­

phragm and longitudinal walls exceeded 50% of the weight of the wall panels for the majority of the condi-

26

tions considered (Fig. 4.22). The magnitude of the forces was not reduced significantly in the nonlinear analy­

ses. Because panel-to-roof connections typically have limited ductility, the researchers concluded that linear

analyses were appropriate for tilt-up construction [53,54].

The distribution of shear forces in the diaphragm is shown in Fig. 4.23. The results indicate that shear

forces do not decrease linearly with distance from the end walls. Proposed shear distributions for design are

also indicated in Fig. 4.23.

Following the 1987 Whittier Narrows earthquake, researchers at Agbabian Associates revised their ana­

l ytical model to reflect the observed damage in tilt -up buildings and to take advantage of the improved model­

ling capabilities developed as part of the TCCMAR research program [29]. The modelling of three actual

buildings is described in Ref. 9. In all three cases, the nature of the analytical model is considerably different

from the earlier analyses [4,6]. For earthquake motion in the transverse direction, the following changes were

made:

• The longitudinal walls are not included in the analyses, because their contribution to the stiffness

of the building was considered to be negligible.

The transverse wall panels were modelled using linear beam elements.

• Nonlinear springs were used to represent the soil supporting the transverse wall panels. Panel

uplift could be evaluated with these elements.

A warehouse in Hollister, California (discussed in Chapter 5 of this report) was analyzed to demonstrate

the performance of the revised nonlinear model for diaphragms (Fig. 4.9). The analytical model of the 300'

by 1 DO' warehouse is shown in Fig. 4.24 for earthquake motion in the transverse direction. The concrete wall

panels were assumed to be uncracked in the analysis. The stiffness of the plywood diaphragm was inferred

from the results of the ABK tests [1] using the scaling procedure defmed in Eq. 4.2 and 4.3. The diaphragm

stiffness was subsequently increased by a factor of 3 to account for the roofmg material and insulation [9].

The response of this building during the 1986 Morgan Hill earthquake was recorded as part of the Califor­

nia Strong Motion Instrumentation Program [38]. Comparisons of the calculated and measured displacement

at the center of the diaphragm, relative to the top of the transverse walls, are shown in Fig. 4.25. The general

nature of the measured response is well-represented by the analytical model.

A 165' by S44 I warehouse in Downey, California and a 294' by 452' building in Whittier, California (Fig.

4.26) were analyzed as part of an investigation of tilt-up performance during the 1987 Whittier Narrows

earthquake [9]. Both buildings were located less than 10 miles from the epicenter. The Downey building

sustained minor structural damage during the earthquake, while no damage was observed in the Whittier

building [9].

27

Analytical models of the Downey and Whittier buildings for ground motion in the transverse direction

are shown in Fig. 4.27 and 4.28, respectively. The models included a more-detailed representation of the

transverse walls than was used to analyze the Hollister building and nonlinear springs were included beneath

the transverse walls in the Downey building to model the soil. The response of the longitudinal walls was

not modelled explicitly, however, the response of the longitudinal wall panels was evaluated by subjecting

an isolated panel to the calculated diaphragm accelerations and the input ground motion.

AI though the response of these buildings was not recorded during the Whittier earthquake, the ground

motion was recorded at six sites wi thin 15 miles of the epicenter, and both buildings were inspected thorough­

ly after the event. Therefore, the performance of the analytical model was evaluated by comparing the extent

of structural damage predicted using the analytical model and damage observed following the earthquake.

The results of the analyses of the Downey building agreed with the observed damage. For transverse

ground motion, panel uplift was calculated to occur in the west and interior walls. When the building was

subjected to longitudinal ground motion, the calculated forces between the diaphragm and the longitudinal

wall panels exceeded the strength of the connections. Evaluation of the longitudinal wall panels subjected

to transverse ground motion indicated that the dynamic moments were less than the cracking load for the pan­

els. The calculated damage in the panel-ta-foundation and panel-to-roofconnections was observed follow­

ing the earthquake, and cracking of the wall panels was not observed.

The correlation between calculated and observed damage in the Whittier building was not as good. The

analyses indicated damage in the panel-ta-roof connections, distress in the diaphragm along the south wall

of the building, and extensive cracking of the longitudinal wall panels, when the building was subjected to

transverse ground motion. None of this damage was observed in the structure. The researchers believed that

the skewed wall panels along the south end of the building may have led to problems modelling the dia­

phragm, and that the ground motion measured approximately 1.2 miles from the building was not representa­

tive of the motion at the site.

4.4 Summary

As indicated in this chapter, the seismic response tilt-up construction has been studied extensively in the

past 15 years. Analytical models for calculating the seismic response of plywood diaphragms and tilt-up

buildings have been summarized. However, little guidance is available for the engineer interested in perform­

ing independent calculations. The response of tilt-up buildings is closely linked to the nonlinear characteris­

tics of the diaphragms. All the researchers scaled experimental data to obtain the parameters used in their

analyses. The link between the design equations discussed in Chapter 3 and the analytical models is missing.

28

Therefore, the influence of variations in the diaphragm, such as the type or spacing of the fasteners or the

thickness of the plywood, on the structural performance can not be evaluated.

The seismic response of tilt -up construction is also related to the performance of the structural connec­

tions between the roof and wall panels, adjacent wall panels, and wall panels and the foundation. With the

exception of the study by Adham et al. [9] where the panel-to-foundation connections were modelled, con­

nections are not considered in the analytical models discussed. Although most damage observed after an

earthquake has been attributed to failure of the connections, the current analytical models can not be used to

evaluate the required strength and ductility of these critical elements.

29

5. MEASURED STRONG-MOTION RESPONSE OF TILT-UP BUILDINGS

Acceleration response histories have been recorded in tilt-up buildings during several recent earthquakes

as part of the California Strong Motion Instrumentation Program [37,38,59,60,61,66,67]. The physical char­

acteristics of three buildings for which data are available are summarized in Table 5.1. Two of the buildings,

the Hollister and Redlands warehouses, represent traditional tilt-up construction. The one-story structures

are rectangular in plan, have relatively few openings in the tilt-up panels, and are used primarily for storage.

The Milpitas industrial building, on the other hand, represents the recent trend of using tilt-up wall panels

in multi-story commercial buildings. The fIrst story in this building is used as a warehouse and the second

for offices. Every wall panel has openings for windows or doors.

The seismic response of each of the structures will be summarized in the following sections. Generaliza­

tions about the dynamic behavior of tilt-up buildings will also be presented.

5.1 Hollister Warehouse

A view of the north-eastcomerofthe Hollister warehouse is shown in Fig. 5.1 and the floor plan is shown

in Fig. 5.2. Six-in. thick tilt-up panels are used throughout the building, with the exception of four 7-in.

panels at the north and south ends of the longitudinal walls. Cast-in-place pilasters are used to connect adja­

cent wall panels. Cambered, glulam beams, ranging in depth from 22V2 in. to 28V2 in. with a width of 5 l/S

in., run in both the longitudinal and transverse directions of the building (Fig. 5.3). The beams are supported

by a single line of 8-in. standard pipe columns. The roof is formed from a grid of 4x14 and 4x10 purlins at

8 ft on center with 2x4 stiffeners at 2 ft on center, overlain by Ih-in. structural plywood. Blocking was pro­

vided throughout the diaphragm. The plywood is covered with 2-in. styrofoam insulation, I-in. fesco board,

and roofing rna terial.

Typical reinforcement in the wall panels consists of a single layer of #4 bars spaced at 12 in. on center

in each direction (Fig. 5.4) Two #9 bars form the chord in the longitudinal walls and a singl~ #5 bar is used

in the transverse walls. Chord reinforcement from adjacent panels was overlapped and welded.

The building was deSigned with nine openings in the tilt-up walls: four overhead doors for truck access

and five doors for personnel (Fig. 5.5). Two #5 bars were typically placed in the wall panels next to the open­

ings.

Records from thirteen strong-motion instruments were obtained during the 1984 Morgan Hill, 1986 Hol­

lister, and 1989 Lorna Pneta earthquakes. Five instruments recorded the ground motion, four monitored

transverse motion at the roof. three recorded longitudinal motion at the roof, and one instrument monitored

the out-of-plane response of a longitudinal wall panel at midheight. Instrument locations are indicated in

Fig. 5.6 and summarized in Table 5.2. Horizontal ground acceleration histories recorded at the base of the

warehouse are shown in Fig. 5.7 for the three earthquakes. Corresponding linear response spectra are pres­

ented in Fig. 5.8.

- Acceleration histories are shown in Fig. 5.9,5.10, and 5.11 for the Morgan Hill, Hollister, and Lorna Prie­

ta earthquakes, respectively. The following observations were made from the acceleration response:

30

• The amplitudes of the transverse accelerations at the center of the roof (channel 4) and the top

of the longitudinal wall (channel 5) were approximately 3 times greater than the corresponding

ground accelerations (channel 7). The out-of-plane motion at midheight of the center longitudi­

nal wall panel (channel 6) exceeded the ground accelerations by a factor of approximately 2.5.

The longitudinal accelerations at the center of the roof (channel 11) were observed to be ampli­

fied by a factor of 1.5 to 2 relative to the longitudinal ground acceleration (channel 13).

In-plane accelerations measured at the top of the walls (channels 2 and 3 for transverse motion

and channels 10 and 12 for longitudinal motion) were essentially the same as the accelerations

recorded at the base of the walls (channel 7 for transverse motion and channel 13 for longitudinal

motion). No appreciable amplification of the in-plane ground motion was observed at the roof.

Normalized Fourier amplitude spectra of the acceleration response histories are shown in Fig. 5.12, 5.13,

and 5.14. A summary of the predominant frequency for each channel is presented in Table 5.2.

Similarly to the acceleration histories, the Fourier amplitude spectra indicate that the frequency

content of the in-plane wall response is essentially the same as the corresponding ground motion.

Out-of -plane response at the center of the walls and response at the center of the diaphragm was

similar and may be used to identify the fundamental natural frequency of the structure. In the

transverse direction, the natural frequency was approximately the same during the Morgan Hill

and Hollister earthquakes, ranging from 1.6 to 1.7 Hz. The natural frequency decreased to 1.10

Hz during the Lorna Prieta earthquake. The decrease in structural stiffness observed during the

Lorna Prieta earthquake is consistent with the increased amplitude of the response and observed

damage following the earthquake [67].

The Fourier amplitude spectra from out-of-plane motion at midheight of the longitudinal wall

panels (channel 6) indicate amplification of response between 3 and 5 Hz. However, the out-of­

plane response of the panels is dominated by the transverse behavior of the building.

Longitudinal structural frequencies are not easily identified from the Fourier amplitude spectra.

The relative frequency content was essentially the same as the ground motion for frequencies

less than 2 Hz. Maximum amplification at the center of the roof occurred between 6 and 7 Hz.

The digitized data provided by the California Department of Conservation included displacement histo­

ries which were obtained by integrating the corrected acceleration response. Displacement response at the

base of the structure and the center of the roof is shown in Fig. 5.15. The longitudinal displacement of the

roof was essentially the same as the north-south ground displacement. Amplification of the transverse dis­

placements at the center of the roof may be observed.

The displacement of the structure relative to the ground may be interpreted as an indication of damage

during an earthquake. However, due to the nature of the numerical integration process, the magnitude of the

relative displacement response must be considered to be approximate. Differences between the integrated

structural displacement records and the integrated ground displacement are shown in Fig. 5.16, 5.17, and 5.18

as the "unfiltered" relative displacement records. It was observed that the predominant frequency of the

31

ground motion tended to dominate the calculated relative displacement response, especially for the transverse

response recorded at the top of the transverse end walls and the longitudinal response. The relative displace­

ment records were filtered in the frequency domain, in an attempt to remove the noise attributable to the

ground motion. Details of the filtering procedure are described in Appendix B. The filtered relative displace­

ment records for the Hollister warehouse are also shown in Fig. 5.16, 5.17, and 5.18. Calculated maximum

relative displacements are summarized in Table 5.3 for the unfiltered and filtered records. The following

trends may be observed:

The relative displacement records in the transverse direction at the center of the building (chan­

nels 4, 5, and 6) were not significantly affected by the filtering process. Maximum relative dis­

placement variations were typically within ± 15% for the unfiltered and filtered records, which

is consistent with the error expected for numerical integration of acceleration records [68]. The

primary difference between the unfiltered and filtered records, was that the filtered relative dis­

placement records tended to oscillate about zero displacement, while the unfiltered records oscil­

lated about the ground displacements.

The character of the relative displacement records in the transverse direction at the end of the

building (channels 2 and 3) were dramatically changed by the filtering process. In many cases,

the maximum relative displacement from the filtered records was less than one-half of the maxi­

mum relative displacement from the unfiltered records. Due to the significant change in the am­

plitude and frequency content of the relative displacement records at the top of the transverse

walls, the relative displacement records for channels 2 and 3 were considered to be unreliable.

Longitudinal relative displacement records (channels 10, 11, and 12) were also dominated by the

ground displacements. The amplitude of the longitudinal relative displacements was larger at the

center of the roof than along the longitudinal walls. However, the relative displacement records

for channels 10, 11, and 12 were considered to be unreliable.

The transverse relative displacements of the roof sustained by the Hollister warehouse during

the Lorna Prieta earthquake were an order of magnitude larger than the relative displacements

of the roof during the Morgan Hill and Hollister events. The maximum relative roof displace­

ment in the transverse direction during the Loma Prieta earthquake was on the order of 1 % of

the building height.

The out-of-plane displacements at the top of the longitudinal wall panel were consistently larger

than the response at midheight of the panel.

5.2 Redlands Warehouse

The east elevation of the Redlands warehouse is shown in Fig. 5.19 and the floor plan is shown in Fig.

5.20. The building is divided nearly in halfby a non-bearing stud partition wall. Panels south of the fire wall

are 22-ftwideand panels north of the firewall are2~ftwide. Panels along the transverse sides of the building

are 22V2-ft wide. All panels are 7-in. thick. Pilasters are used to connect adjacent panels.

32

Cambered glulam beams span between the longitudinal walls (Fig. 5.21). The Yz-in. plywood-sheathing

is supported by 4x14 purlins at 8-ft on center and 2x4 rafters at 2-ft on center. All four openings in the perime­

ter walls (two overhead doors and two personnel doors) were located in the east, longitudinal wall (Fig. 5.22).

Structural response during the 1986 Palm Springs, 1992 Landers, and 1992 Big Bear earthquakes was

recorded at 12 locations. Three instruments recorded ground motion, five recorded transverse response at the

roof, three recorded longitudinal response at the roof, and one recorded the out-of-plane response of a longi­

tudinal wall panel at midheight. Instrument locations are indicated in Fig. 5.20 and summarized in Table 5.4.

Horizontal ground acceleration histories recorded at the base of the warehouse are shown in Fig. 5.24 for the

Palm Springs earthquake. Corresponding linear response spectra are presented in Fig. 5.25. Digitized data

are not yet available from the Landers and Big Bear earthquakes.

Acceleration histories are shown in Fig. 5.26,5.27, and 5.28 for the Palm Springs, Landers, and Big Bear

earthquakes, respectively. Observations from the acceleration response are summarized below:

The maximum transverse acceleration response was measured at the quarter-point of the longi­

tudinal walls (channelS) during the Palm Springs and Landers earthquakes, indicating that the

non-bearing fire wall and overhead door openings in the longitudinal wall influenced the dynam­

ic response of the structure. During the Big Bear earthquake, maximum transverse accelerations

were recorded at the center of the longitudinal wall (channel 4). This change in behaviorindicates

that the stiffness of the fire wall decreased during the Big Bear event, however, no information

on observed damage is available.

The amplitude of the transverse accelerations at the roof (channels 3 and 5) were 3 to 5 times the

amplitude of the transverse ground acceleration (channel 12). The out-of-plane accelerations

at midheight of the center longitudinal wall panel (channel 2) were approximately 2.5 times the

magnitude of the transverse ground accelerations.

The rna gill tude 0 f longi tudinal accelerations at the center of the transverse walls at the roof level

(channe 1 9) were am plified by a factor of approximately 3 relative to the longitudinal ground ac­

celerations (channel 11).

The in-plane acceleration response at the top of the longitudinal (channels 8 and 10) and trans­

ve rse (c ha nn e I s 6 and 7) walls was a pproximatel y the same as the corresponding ground accelera­

tions (channel 11 in the longitudinal direction and channel 12 in the transverse direction).

Normalized Founer amplitude spectra for the Palm Springs earthquake acceleration records are shown

in Fig. 5.29.

The fundamental transverse natural frequency of the warehouse was observed to be 2.6 Hz. The

fundamental natural frequency in the longitudinal direction was 3.2 Hz.

The Fourier amplitude spectra for in-plane wall response were essentially the same as those for

the corresponding ground acceleration records.

33

Integrated displacement response at the base of the structure and the center of the roof is shown in Fig.

5.30. The longitudinal displacement of the roof was essentially the same as the north-south ground displace­

ment. The transverse response of the roof may be observed in the east-west absolute displacement record.

Unfiltered and filtered relative displacements for the Redlands warehouse are shown in Fig. 5.31. Calcu­

lated maximum relative displacements are summarized in Table 5.5 for the unfiltered and filtered records.

The following trends may be observed:

The relative displacement records in the transverse direction at the center of the building (chan­

nels 2, 3, 4, and 5) were not significantly affected by the filtering process. Roof-level relative

displacements at the quarter-point of the longitudinal wall (channel 5) exceeded those at the cen­

ter of the longitudinal wall (channel 3), indicating that the fire wall influenced structural re­

sponse.

The character of the relative displacement records in the transverse direction at the end of the

building (channels 6 and 7) were dramatically changed by the filtering process. The relative dis­

placement records for channels 6 and 7 were considered to be unreliable.

Longitudinal relative displacement records recorded on the top of the longitudinal walls (chan­

nels 8 and 10) were also dominated by the ground displacements. The relative displacement re­

cords for channels 8 and 10 were considered to be unreliable. Longitudinal relative displace­

ments recorded at the top of the south transverse wall (channel 9) were not significantly

influenced by filtering. The amplitude of the longitudinal relative displacements measured by

channel 9 were approximately one-fifth of the transverse relative displacements recorded by

channel 5.

The maximum relative roof displacement sustained by the Redlands warehouse during the Palm

Springs earthquake was less than 0.1 % of the building height.

The out-of-plane displacements atthe top of the longitudinal wall panel were consistently larger

than the response at midheight of the panel.

5.3 Milpitas Industrial Building

The north-west corner of the two-story Milpitas industrial building is shown in Fig. 5.32 and the floor

plans are shown in Fig. 5.33. The tilt-up panels are typically 24-ft wide with window openings at both the

first and second story levels (Fig. 5.34). Panel thickness varies between 16 in. along the panel edges to 8 in.

above and below the windows. Chord reinforcement is located at the second floor and roof levels and is

welded between adjacent panels.

Eighteen, structural steel tube columns are used to carry the vertical floor and roof loads. The columns

are arranged in a 24x30-ft grid. Deep, open web steel girders span in the longitudinal direction of the building

at the second floor level. Open web steel j oists span between the girders in the transverse direction and support

a metal deck and a 2Y2-in. concrete slab. Puddle welds were used to connect the metal deck to the joists and

girders. The pitched roof is supported by gIulam beams running in the transverse direction. The roof dia­

phragm consists of Y2-in. plywood sheathing with 2x4 joists and 6x16 purlins.

34

Thirteen instruments recorded the response of the building during the 1988 Alum Rock and 1989 Loma

Prieta earthquakes. Instrument locations are shown in Fig. 5.35 and summarized in Table 5.6. Five instru­

ments recorded the ground motion, the transverse building response was monitored by three instruments at

the roof and three at the second floor, and the longitudinal building response was recorded by one instrument

at the roof and one at the second floor. Horizontal ground acceleration histories recorded at the base of the

building are shown in Fig. 5.36 for the two earthquakes. Corresponding linear response spectra are presented

in Fig. 5.37.

Measured acceleration records are shown in Fig. 5.38 and 5.39 for the Alum Rock and Loma Prieta earth-

quakes, respectively. Observations are noted below:

The maximum transverse accelerations recorded at the center of the longitudinal walls at the roof

level (channel 4) were approximately 3 times greater than the maximum transverse groundaccel­

era ti ons (channel 9). Maxim urn transverse accelerations recorded at the second floor level (chan­

nel7) were approximately 25% greater than the transverse ground accelerations.

The in-plane response of the transverse walls measured at both the roof and second floor levels

(channels 3, 5, 6, and 8) was essentially the same as the transverse ground acceleration (chan­

nel7).

The magnitude of the longitudinal acceleration response, measured at the center of the transverse

walls (channel 11 at the roof and channel 12 at the second floor), exceeded the transverse accel­

eration response at both the roof (channel 4) and second floor level (channel 7) during both earth­

quakes. Amplification factors, relative to the base, exceeded 4 for longitudinal acceleration re­

sponse at the roof and were approximately 1.6 at the second floor level.

The corresponding Fourier amplitude spectra are shown in Fig. 5.40 and 5.41. The predominant natural

frequencies are between 3.5 and 5 Hz in the transverse direction and between 4.5 and 5.5 Hz in the longitudi­

nal direction. The frequency signature is less pronounced in the data from the 1988 Alum Rock earthquake.

Integrated displacement histories are shown in Fig. 5.42. The amplitude of the ground displacement is

an order of magnitude larger during the Lorna Prieta earthquake than during the Alum Rock earthquake. As

a result of the large ground displacements during the Lorna Prieta earthquake, the relative displacements of

the structure are not significant. Relative displacements at the roof may be observed in both the longitudinal

and transverse directions during the Alum Rock earthquake however.

Unfiltered and fIltered relative displacements for the Milpitas industrial building are shown in Fig. 5.43

and 5.44. Calculated maximum relative displacements are summarized in Table 5.7 for the unfiltered and

filtered records. The following trends may be observed:

• The signal-ta-noise ratios for the relative displacements in the Milpitas industrial building were

smaller than those observed for the Hollister and Redlands warehouses. Therefore, the reliability

of all the relative displacement data must be questioned.

The relative displacement records from all channels during the Lorna Prieta earthquake were sig­

nificantly affected by the filtering process.

35

Only two channels of relative displacement data during the Alum Rock earthquake appear to be

insensitive to filtering: channels 4 and 11, which represent the transverse and longitudinal re­

sponse at the center of the roof. The maximum relative displacement of the roof was less than

0.05% of the height of the building.

5.4 Summary

The measured response of three tilt-up buildings during seven recent earthquakes in California has been

presented. Although the structural systems used in the three buildings differ, the following generalizations

about the seismic response of tilt-up construction may be made:

Transverse accelerations were observed to be amplified by a factor of approximately 3 between

the base and the center of the roof. The measured out-of-plane response at midheight of the lon­

gitudinal wall panels was amplified relative to the ground accelerations. However, maximum

amplification was observed at the roof level.

The magnitude of the amplification of the longitudinal accelerations appeared to be dependent

upon the aspect ratio and structural characteristics of the building. Amplification factors ranged

from 1.5 in the Hollister warehouse to 4 in the Milpitas industrial building.

In-plane acceleration response of the transverse walls at the roof level was essentially the same

as the corresponding ground motion. The magnitudes of the acceleration histories were not am­

plified appreciably, and the frequency content of the signals was nearly identicaL

Displacement of the roof relative to the ground was more pronounced in the transverse than the

longitudinal building response. Maximum out-of-plane displacements at the roof level were

approximately twice those measured at the midheight of the panel.

Non-bearing partition walls and openings in wall panels may influence the behavior of tilt-up

construction. Maximum transverse acceleration response was observed at the quarter-point of

the longitudinal walls in the Redlands warehouse.

36

6. OBSERVED PERFORMANCE OF TILT-UP CONSTRUCTION

DURING EARTHQUAKES

Investigations of the performance of individual tilt-up structures during the 1964 Alaska, the 1971 San

Fernando, the 1987 Whittier Narrows, and the 1989 Lorna Prieta earthquakes are summarized in this chapter.

Typical types of damage are listed in Table 6.1, along with an indication of the frequency. The observed be­

havior of tilt-up construction during the four earthquakes is summarized in Sections 6.1 through 6.4. Section

6.5 contains some general observations.

6.1 The 1964 Alaska Earthquake

The 1964 Alaska earthquake damaged tilt-up structures at the Elmendorf Air Force Base and provided

the first evidence of the potential seismic vulnerability ofthis type of construction [32]. The Elmendorf Ware­

house suffered the worst structural damage in the area: three of five bays collapsed. The plan view of the build­

ing is shown in Fig. 6.1 and typical roof framing and concrete fire wall details are shown in Fig. 6.2. Adjacent

structures of different forms of construction sustained only minor damage, implying that the cause of the col­

la pse was not the rna gni tude of the earthquake (ML =8.3-8.6) but rather the structural system used in the ware­

house.

Following the earthquake, investigators identified the likely cause of failure to be pullout of the anchor

bolts from the tilt-up concrete walls. The anchor bolts connected the wall panels to the steel frame and ply­

wood roof diaphragm. This failure mechanism could have been prevented by installing ties in the concrete

column to confme the anchor. Brittle failure of the steel reinforcement in concrete columns and of the welds

connecting the cross bracing in the fire walls to the steel roof framing members was also observed [32]. These

connections were unable to develop the full strength of the structural members, suggesting problems related

to insufficient connection ductility, as well as connection strength.

6.2 The 1971 San Fernando Earthquake

Major structural damage was observed in tilt-up warehouses located in the Sylmar Industrial Tract and

San Fernando Industrial Tract following the 1971 San Femando earthquake. Studies of eight tilt-up buildings

are reported in Ref. 41 and 55. Collapse of the roof or wall panels was observed in four of these structures.

Most of the fail ures were attributed to inadequate connection details between the plywood diaphragm,

the ledger beams. and the ult-up panels (Fig. 3.20). Three modes of failure occurred at this interface:

plywood pulled through the nails;

nails pulled out of the ledger;

ledgers split in cross-grain bending (Fig. 6.3).

Once the panel-ta-roof connection failed, the roof framing system was susceptible to failure of the glulam­

ta-pilaster or purlin-to-Iedger connections. Loss of these connections allowed the framing members to slip

off their seats, leading to collapse of the roof. The out-of-plane resistance of the tilt-up panels is essentially

zero once the adjacent roof element has fallen or the panel-ta-roof connection has failed. The wall panel is

then also susceptible to collapse.

37

Evaluation of the roof and wall collapses that occurred during the 1971 San Fernando earthquake indi­

cates that most roof collapses originated in areas where the purlins framed into the wall panels (Fig. 6.4). High

in-plane shear forces develop along the shorter side of the diaphragm and therefore the plywood-ta-Iedger

connections along the end walls are more susceptible to damage than connections along the longitudinal

walls. Also, beam seats provide more stability and redundancy to connections between glulam beams and

concrete pilasters than the hangers used to form the connections between purlins and ledgers. After the ply­

wood-to-Iedger connections are lost, it is reasonable to assume that the next failure mechanism will occur

at the connection of purlins to other framing elements.

Significant damage also occurred in buildings that did not collapse. Cracking and spalling of the pilasters

or corbels was observed in three of the eight tilt-up buildings considered [41,55]. Cracking and permanent

out-{)f-plane deformations were also observed in the wall panels. Damage to wall panels was attributed to

excessive flexural deformation due to large in-plane roof deformations [41]. Damage to corbels and pilasters

spalling was also attributed to displacement of the roof diaphragm.

As a result of the poor performance of the plywood-ta-ledger connections during the 1971 San Fernando

earthquake (Fig. 3.20), provisions were added to Section 23120) of the Uniform Building Code to prohibit

the use of these connections in regions of high seismic risk [75]. Positive, direct connections between the roof

diaphragms and the supporting walls are currently required in Section 2310 of the UBC [77].

6.3 The 1987 Whittier Narrows Earthquake

The 1987 Whittier Narrows earthquake provided the first major test of the tilt-up design requirements

adopted following the San Fernando earthquake. The number of roof and wall panel collapses during the

Whittier Narrows earthquake was greatly reduced compared with those during the 1971 San Fernando earth­

quake, and all occurred in structures built before 1971. However, the magnitude of the 1987 earthquake

(ML=5.9) was considered moderate, as compared with the 1971 earthquake (ML=6.4). Therefore, the possi­

bility that greater structural damage will occur during a stronger earthquake can not be dismissed for tilt-up

buildings designed using the post-1978 UBC provisions.

The most severe damage in more modem tilt-up structures during the 1987 earthquake occurred in build­

ings that had wall panels with large openings. Observations of panels bowing out and cracking near the upper

corners of openings were common in buildings constructed after 1983. This type of damage has been attrib­

uted to two sources [35]:

insufficient panel reinforcement, or incorrect placement of reinforcement within the panel;

openings had been cut into existing panels without providing additional reinforcement.

Adham et al. [10] suggest that panels with large openings should be proportioned to resist flexural action as

if the panel were a frame: the vertical piers should be designed to resist their own inertial load plus that of

the portion of the wall above the opening. Design of these piers should conform to provisions in Section

2625(f)9B of the UBC entitled "Wall Piers" [77].

38

When modifications are made to an existing building, many owners overlook the need to replace the

strength and stability lost when an opening is cut in a tilt-up panel [10,35]. Steel columns or "kickers" are

usually recommended to increase the lateral-load resistance of a panel with large openings (Fig. 6.5).

The the response of tilt-up buildings during the 1987 earthquake also highlighted the importance of tying

the structure together and providing adequate collector elements to carry the force away from reentrant cor­

ners [35]. A number of failures of plywood in roof diaphragms could have been prevented if ties had been

provided to ensure that the purlins did not separate from the glulam beams (Fig. 3.26). Distress of roof ele­

ments was reported near reentrant comers and skewed joints where framing members were often not able to

transmit chord forces into the diaphragm (Fig. 6.6).

Cases of roof distress and plywood failure directly above vertical elements such as stair wells and interior

columns or walls were also reported. In many cases, collector elements were not provided to transmit dia­

phragm forces into the vertical members.

The Whittier Narrows earthquake also provided information on the seismic performance of a number of

common construction practices that were developed during the 1970's and 80's. In contrast to buildings

constructed before 1971 when cast-in-place pilasters provided continuous connection between adjacent wall

panels, the chord reinforcement at the elevation of the dia phragm is the often only panel-ta-panel connection

in modern tilt-up construction. Additional panel-to-panel connections are provided only in the comers of

the building. Changes in the connection design were adopted to improve the durability of tilt-up construction

under temperature and shrinkage induced loads. Two tilt-up buildings with minimal panel-ta-panel connec­

tions were studied following the Whittier Narrows earthquake [10]. Little or no structural damage was ob­

served. However, the individual panels were considered to be more susceptible to damage due to out-of­

plane bending than comparable structures with pilasters. The reduction in the number of panel-ta-panel con­

nections is also believed to lead to panel uplift [9] .

The structural implications of using steel ledgers and metal screws for the panel-ta-roof connections

were investigated by studying the behavior of the Downey and Whittier buildings (Fig. 4.26) [10]. Typical

panel-to-roof connection details are shown in Fig. 6.7 [9]. The metal screws failed along the transverse walls

in the Downey building. Structural damage was expected to be considerably greater if the building had been

subjected to the design-level earthquake [9]. Although replacing wooden ledgers with steel ledgers elimi­

nates the cross-grain bending failure mechanism in the panel-ta-roof connections shown in Fig. 6.3, it is not

sufficient to guarantee acceptable connection performance.

6.4 The 1989 Lorna Prieta Earthquake

The sa tisfactory performance of most engineered buildings during the 1989 Loma Prieta earthquake has

been attributed to the relatively short duration of the ground motion [69].

Although there were isolated cases of major damage to tilt-up concrete industrial buildings, collapses

were not as widespread as during the 1971 San Fernando earthquake. Several cases in which the contents of

the building influenced the structural performance were identified. A tomata-storage warehouse in Hollister

lost part of a wall when stacks of cans inside the building fell against the wall panels and broke the connection

39

between the wall panels and pilasters [69]. In an industrial park west of Watsonville, another tilt-up structure

sustained moderate structural damage to several pilasters and a wall panel separated from the roof diaphragm

when a free-standing steel-frame mezzanine struck the exterior walls [69].

6.5 Summary

The observed performance of tilt-up buildings during the 1964, 1971, 1987, and 1989 earthquakes indi­

cates that this form of construction is susceptible to structural damage (Table 6.1). Damage in buildings

constructed before the 1971 San Fernando earthquake may usually be attributed to:

• wood ledger members failing in cross-grain bending;

• nails pulling through the edges of the plywood at the ledger or at interior panel edges;

• edge nails pulling out of wood ledgers or interior framing members;

• brittle fracture of welded connections.

The damage statistics presented in Table 6.1 indicate that the seismic performance of tilt-up buildings

constructed after the 1971 San Fernando earthquake improved significantly. However, the following suscep­

tibilities were identified:

• cracking and permanent out-of-plane deformations in panels with large openings;

• excessive displacement or flexibility of the diaphragms, particularly for those with very large

spans;

• improper connection or anchorage details between adjacent wall panels and between the wall

panels and the foundation.

40

7" CONCLUSIONS

Tilt-up construction is a proven cost-effective method of erecting low-rise buildings. However, the need

to develop methods to mitigate the seismic hazards continues to grow with the increasing use of tilt-up for

commercial and industrial buildings that contain a large number of workers and expensive equipment. Past

seismic performance indicates that initial savings during construction can be offset by the cost of repairing

structural damage and of downtime following an earthquake.

Tilt-up buildings constructed before 1973 are susceptible to failure of the connections between the wall

panels and the roof diaphragm which often leads to the collapse of the roof and wall panels. The 1971 San

F emando earthquake highlighted the risk of vulnerable connections, and building code provisions were soon

modified [75] to avoid many problems, such as cross-grain bending in wood ledger beams. However, design

procedures developed for traditional warehouse construction may not be appropriate for buildings with geo­

metrically complex floor plans or a large number of openings in the panels. Damage to wall panels with open­

ings and roof connector elements during the 1987 Whittier earthquake indicates that modern tilt-up buildings

are also susceptible to seismic damage.

A review of current design procedures and detailed analytical models for tilt-up construction indicates

that buildings are often treated as a group of individual components, rather than a complete structural system.

The diaphragm and wall panels are considered independently and connections are typically not modelled in

the analyses. Although the seismic response is closely tied to connector performance, analytical models are

not currently available to evaluate the influence of connection details on the structural response.

The measured response of three tilt-up buildings in California was used to identify trends in the seismic

behavior. The buildings represented different eras in tilt-up construction: the structures in Hollister and Re­

dlands were one-story warehouses with plywood roof diaphragms and cast-in-place pilasters, while the

structure in ~1il?ltas was two-stories tall, included a metal-deck floor diaphragm and a wood roof, and had

window operungs ill every panel. However, the general nature of response was similar in all three buildings:

Transverse accelerations measured at the center of the roof were appro ximatelythree times larger

than the corresponding ground accelerations.

The aInputudes of the transverse accelerations and displacements at the center of the roof were

larger than the amplitudes of the response measured at mid-height of the center longitudinal wall

panel.

The in -p lane accelerations measured at the top of the transverse walls were essentially the same

as those measured at the base of the walls.

In addition. data from the Redlands warehouse demonstrated that non-bearing partition walls and openings

in wall panels may influence the response of tilt-up systems.

These observations are not consistent with the typical design assumptions. For example, the measured

response indicated that wall panels do not behave as columns pinned at both ends. The roof of the Hollister

warehouse sustained transverse displacements larger than 4 in. (1 % of the building height) relative to the base

41

during the 1989 Lorna Prieta earthquake. The middle of the panel experienced a maximum transverse dis­

placement of approximately 2.5 in. during the same event.

The differences between the expected and measured response indicate that the seismic response of tilt-up

construction is not completely understood. Additional work is required to develop analytical models that are

sensitive to the nature of the critical connections in the building and can be used to mitigate seismic hazards

in new and existing construction. With an improved understanding of system behavior, tilt-up construction

can be made as safe as it is economical.

42

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48

TABLE 3.1 ALLOWABLE SHEAR IN LB/FT FOR HORIZONTAL PLYWOOD DIAPHRAGMS WITH FRAMING OF DOUGLAS FIR-LARCH OR SOUTHERN PINE1 [77]

Blocked Diaphragms Unblocked Diaphragms

Minimam Minimam Nail spaciag at diaphDgDl boundaries (all

cases), at c:oatiaaos paad edges parallel to Nails spaced at 6" max. at sap ported cad Nominal Minimam Nominal load (Cases 3 :and 4) :and at all paael cdges

Common Petlctralion Nominal Width of (Cases 5 :and 6) Nail in Plywood Framing

Plywood Grade Szc Framing thick.ncss Member 6 4 21h2 22 (ia.) (in.) (in.) Load perpendicular to

Nail spaciag at other plywood pand edges unblocked edges :and Other c:onfigaratioos coalinaoas pand joiats (Cases 2, 3, 4, 5 :and 6)

6 6 4 3 (Case 1)

6d 11/4 51I6 2 185 250 375 420 165 125 3 210 280 420 475 185 140

STRUCTURAL I 8d P/2 3/S 2 270 360 530 600 240 180 3 300 400 600 675 265 200

10d 1 5/S 15

132 2 320 425 640 730 285 215 3 360 480 720 820 320 240

51I6 2 170 225 335 380 150 110 6d 1% 3 190 250 380 430 170 125

3/S 2 185 250 375 420 165 125 3 210 280 420 475 185 140

C-D,~, 3/s 2 240 320 480 545 215 160 STR UCIlJRAL II

and other grades 8d 1111 3 270 360 540 610 240 180

covered in U.B.C. 15132 2 270 360 530 600 240 180

Standard No. 25-9 3 300 400 600 675 265 200

15132 2 290 385 575 655 255 190 10d 1 5/s 3 325 430 650 735 290 215

1%2 2 320 425 640 730 285 215 3 360 480 720 820 320 240

1 Th ese val ues a re for sho rt-term loa ds due to wind or earthq uake and must be reduced 25% for normal loading. Space nails 12 in. on cen ter along intermediate framing members. Allowable shear val ues for nails in framing members of other species set forth in Table No. 25-17-1 of the U.B.C. Standards shall be calculated for all grades by multiplying the values for nails in Structural I by the following factors: Group ill, 0.82 and Group Iv, 0.65.

2 Framing at adjoining panel edges shall be 3-in. nominal or wider and nails shall be staggered where nails are spaced 2 in. or 21/2 in. on center.

3 Framing at adjoining panel edges shall be 3-in. nominal or wider and nails shall be staggered where 10d nails having penetration into framing of more than 1 5/8 in. are spaced 3 in. or less on center.

49

TABLE 3.2 DIAPHRAGM SHEAR STIFFNESS [3]

Range of Category Shear Stiffness Type of Diaphragm

(klin.)

Very Flexible < 6.7 straight and diagonally sheathed wood diaphragms

Flexible 6.7 -15 special diagonally sheathed wood diaphragms, plywood sheathing, lightly-fastened light-gauge steel decks

Semi-Flexible 15 -100 plywood sheathing, moderately-fastened medium-gauge steel decks

Semi-Rigid 100 -1000 heavily-fastened heavy-gauge steel decks, composite diaphragms

Rigid > 1000 cast-in-place concrete decks

TABLE 3.3 FASTENER SLIP EQUATIONS [74]

Minimum For Maximum Approximate Slip, en (in.) a,b

Fastener Penetration Loads up to (in.) (lb) GreenJDry DryiDry

6d common nail 11/4 180 (V n/434)2.314 (V n/456)3.144

8d common nail 1 7116 220 (V n/857) 1.869 (V n/616)3.018

10d common nail 1 5/8 260 (V n/977) 1.894 (V nn69)3.276

14-ga staple 1 to 2 140 (V n/902)1.464 (V n/596)1.999

14-ga staple 2 170 (Vn/674) 1.873 (V n/461 )2.776

a Fabricated green/tested dry (seasoned); fabricated dry/tested dry. Vn = fastener load.

b Values based on Structural I plywood fastened to Group II lumber. Increase slip by 20% when plywood is not Structural 1.

50

TABLE 3.4 STRENGTH OF METAL-DECK DIAPHRAGMS [48]

Mode of Failure Nominal Strength Allowable Shear for Design (kip/ft) (kip/ft)

Failure of Connections

Edge Connections _ ( )Qf Su - 2a 1 + np a2 + ne r

Interior Panel Qf S u = (2A (l - 1) + B) r

Corner Fasteners _}: (lf2B2)

Notation: Su Sc S L

Lv d w t

D Q[ G-al

a2

Su - (L2J{2 + B2) Qf

Stability Sc = (3~~O)(J3fdf25

= = = = =:

= = = = = = = =

nominal shear strength of diaphragm, kip/ft critical shear for stability of diaphragm, kip/ft allowable shear strength for design, kip/ft panel length, ft purlin spacing, ft corrugation pitch, in. width of the deck panel, in. thickness of the metal deck, in. panel depth, in. strength of a structural fastener, kip (defined in Table 3.5) strength of a sidelap fastener, kip (defined in Table 3.5) ~ Xe /w = end distribution factor ~ Xp /w = purlin distribution factor

Su S :5 2.75

S Su :5 2.35

Sc S :5 2.0

as Xe Xp fIe

= = = =

ratio of sidelap fastener strength to structural fastener strength, Qs/Q/ distance from the panel centerline to a fastener at the end support, in. distance from the panel centerline to a fastener at purlin support, in. number of intermediate sheet-ta-structure connections per panel length and between purlins at the diaphragm edge

rIp ~ A

N I s

= = =

= = =

number of purlins excluding those at ends or endlaps number of stitch connections within lengthL 1 for single-edge fasteners 2 for double-edge fasteners number of fasteners per foot along the ends moment of inertia of the sheet, in.4/ft developed flute width (2(e+w) + fin Fig. 3.7), in.

B = n s as + (2n p E x~ + 4 EX;)

1 DLv A = 1 - 240ft

51

for welded connections

for mechanical connections

TABLE 3.5 STRENGTH AND FLEXIBILITY OF CONNECTIONS IN METAL DECK DIAPHRAGMS [48]

(a) Structural Connections

Type of Strength

Qj Connector (kip)

Puddle Welds 2.2 t F u (d - t) without washers

wi th washers 99 t ( 1.33 do + 0.3 F xx t)

Screwed Connections 1.25 Fy t (1 - 0.005 Fy)

Power-Driven Pins Ramset26SD 62.5 t (1 - 5t)

Hilti ENP2-21-L15 61.1 t (1 - 4t) Hilti ENP3-21-L15

Hilti ENKK 52.0 t (1 - 3t)

(b) Sidelap Connections

Type of Strength ~ Connector (kip)

Puddle Welds 1.65 t Fu (d - t) without washers

wi th washers 74.25 t ( 1.33 do + 0.3 F xx t)

Screwed Connections 115 d t

Power-Driven PIns 240 r2

Notation: t = thickness of metal deck, in. d = average viSible diameter of weld, in. or major diameter of screw, in. Fu = specified mInImum strength of metal deck, ksi do = diameter of hole in washer, in. Fxx = electrode strength, ksi Fy = yield stress of metal deck, ksi.

Notes: Applicable for deck thicknesses between 0.0285 and 0.0635 in. Washers are recommended for deck thicknesses less than 0.028 in. Equations developed for No. 12 and No. 14 screws. Applicable for deck thicknesses between 0.024 and 0.60 in.

Flexibility Sf

(in./kip)

0.00115 {t

0.0013 {t

0.0025 {t

0.00125 {t

0.00156 {t

Flexibility Ss

(in./kip)

0.00125 {t

0.0030 {t

0.030

T

(1) (2) (3) (4) (5) (6)

Strength is independent of screw because deck typically fails before screws yield. Strength and flexibility do not depend on the type of pin.

52

Notes

(1)

(2)

(3)

(4)

Notes

(1)

(2)

(5)

(6)

TABLE 3.6 STRENGTH OF COMPOSITE DIAPHRAGMS [48]

Type of Concrete Nominal Strength (kip/ft)

Allowable Shear for Design (kip/ft)

Structural

Type I - insulating

Type II - insulating

Notation: Su S L w

0 ~ as Xe Xp fle

f1p I1s w fe' B

= ::

= ::

::

::

::

::

::

::

::

::

::

::

nominal shear strength of diaphragm, kip/ft allowable shear strength for design, kip/ft panel length, ft width of the deck panel, in. strength of a structural fastener, kip (defined in Table 3.4) strength of a sidelap fastener, kip (defined in Table 3.4) ratio of sidelap fastener strength to structural fastener strength, Qs/Q[ distance from the panel centerline to a fastener at the end support, in. distance from the panel centerline to a fastener at purlin support, in. number of intermediate sheet-ta-structure connections per panel length and between puriins at the diaphragm edge number of purlins excluding those at ends or endlaps number of stitch connections within lengthL unit weight of concrete, Ib/ft3

specified compressive strength of the concrete, psi. " ~ 2 . 4 ~ 2) ns as ~ \.:..np "- xp ~ L.. Xe

TABLE 3.7 MAXIMUM DIAPHRAGM DIMENSION RATIOS [77]

Horizontal Diaphragms Vertical Diaphragms

Ma:enal Maximum Maximum Span-Width Height-Width

Ratios Ratios

1. Diagonal sheathmg. convcouooal 3:1 2:1

2. Diagonal sbeatbmg. SpeClal 4:1 3Y2:1

3. Plywood and particleboard, nailed all edges 4:1 3Y2:1

4. Plywood and particle board, blocking omitted at 4:1 2:1 intermediate joints

53

VI ~

TABLE 4.1 SUMMARY OF EXPERIMENTAL TESTS OF DIAPHRAGMS SUBJECTED TO LOAD REVERSALS ------ --------

Number of Reference Diaphragms Diaphragm Diaphragm Type of Loading Experimental

Tested Material Size Variables

Young and Plywood thickness Medearis 8 plywood 8' x8' Static load reversals Nail size and spacing

(1962) [83] Influence of low-amplitude cycles

5 plywood Load reversals (28 mm/sec) Plywood thickness ABK (1981) 6 I" x6" sheathing 20' x 60' Forced vibration Roofing material and overlays

[1] 3 metal deck (earthquake motion) Blocking and chords

Zacher and Type of fasteners Gray (1985) 9 plywood 8' x8' Dynamic (2 Hz) Influence of over-driving fastener

[84,85] 4 gypsum board 3 cycles per displacement level Size of nail

Forced vibration I tani and Falk 5 plywood 8' x 24', (earthquake motion and sine sweep) Influence of openings

(1986) [39] 5 gypsum board 16' x 16', or Free vibration Sheathing orientation 16' x 28' Static load reversals

19 plywood Static load reversals Sheathing orientation Dolan (1989) 16 waferboard 8' x8' Sinusoidal loading Nail spacing

[26] Shaking table (earthquake motion) Vertical load - ----- -- ----- -~-----

Vl VI

TABLE 4.2 SUMMARY IOF ABK IJIAPHRAGM TESTS [1]

======F======:===============-' --

Diaphragm Description ID

B

C

D

E

El

H

II

K

N

P

Q

R

S

Y2/1 plywood, unblocked, chorded

lh/l' plywood, unblocked, unchorded, built-up roo

lh/l plywood, unblocked, chorded, built-up rnofing, retr

I" x 6 11 straight sheathing, unchorded, built-up ro

1" x 6/1 straight sheathing, unchorded, built-up roofing, re:

I" x 6" straight sheathing, 5/16/1 plywood overlay, C

1" x 6" diagonal sheathing, unchorded, built-up ro

I" x 6" diagonal sheathing, unchorded, built-up roofing, r

I" x 6" diagonal sheathing, 1/1 x 6/1 straight sheathing ove

Y2" plywood, blocked, chorded

3/4/1 plywood, 3/4" plywood overlay, blocked, chor

20-ga steel decking, unfilled, unchorded, button--punched se

20-ga steel decking, unfilled, chorded, button-punched sea

20-ga steel decking, 2Y2" concrete fill, chorded, button-punche

Measured stiffness during low-amplitude, quasi-static tests.

'ing

lfit nailing

lfing

:rofit nailing

lorded

)fing

trofit nailing

lay, chorded,

ied,

ams 18" O.C.

ms 16/1 O.C.

i seams 18" D.C.

-

Number of Static Tests

4

4

4

4

-

4

3

-

5

4

4

4

4

4 ---- ------

Number of Average Initial Dynamic Stiffness •

Tests (k/in.)

6 9.1

6 6.2

7 12.0

6 6.6

4 -7 14.7

6 18.1

4 -

7 45.5

10 20.4

6 52.9

8 16.6

6 27.9

7 ---------

TABLE 4.3 SUMMARY OF DYNAMIC TESTS OF TILT -UP WALL PANELS [6, 8, 7]

Effective Test No. Motion Earthquake * Peak Diaphragm

Seq. No. Acceleration Stiffness (g)

1 1 El Centro 0.2 Flexible 2 2 E1 Centro 0.2 Rigid 3 3 Castaic 0.2 Flexible 4 4 Castaic 0.2 Rigid 5 5 El Centro 0.4 Flexible 6 6 El Centro 0.4 Rigid 7 7 Castaic 0.4 Flexible 8 8 Castaic 0.4 Rigid 9 1 El Centro 0.2 Flexible 10 1 E1 Centro 0.2 Flexible

11 2 El Centro 0.2 Rigid 12 3 Castaic 0.2 Flexible 13 4 Castaic 0.2 Rigid 14 5 E1 Centro 0.4 Flexible 15 6 E1 Centro 0.4 Rigid 16 7 Castaic 0.4 Flexible 17 8 Castaic 0.4 Rigid 18 6 E1 Centro 0.4 Rigid 19 1 El Centro 0.2 Flexible 20 2 El Centro 0.2 Rigid

21 3 Castaic 0.2 Flexible 22 4 Castaic 0.2 Rigid 23 5 El Centro 0.4 Flexible 24 6 E1 Centro 0.4 Rigid 25 7 Castaic 0.4 Flexible 26 8 Castaic 0.4 Rigid 27 6 El Centro 0.4 Rigid 28 6 El Centro 0.4 Rigid 29 6 El Centro 0.4 Rigid

30·· 2 I EI Centro 0.2 Flexible 31-- 6 I EI Centro 0.4 Flexible ;

• El Centro - NS component from 1940 El Centro earthquake Cast3.1C - N69Vi rornponent from 1971 San Fernando earthquake.

Panel No. Reinforcement

2 5-#4

3 5-#3

4 5-#4

4** 5-#4

** Panel #4 was repalfed wIth epoxy after Test 29. Tests 30 and 31 were conducted on the repaired panel.

56

VI -l

TABLE 5.1 CHARACTERISTICS OF INSTRUMENTED TILT-UP BUILDINGS

~---

Typical Building Year Plan Panel Connections

Designed Dimensions Dimensions • Between Panels

(ft)

Hollister 18 ft (W) Warehouse 1979 300 x 100 30 ft (H) Pilasters

6 in. (T)

Redlands 22 ft (W) Warehouse 1971 242 x 90 27 ft (H) Pilasters

7 in. (T)

Milpitas 24 ft (W) Chord reinforcement Industrial Building 1984 168 x 120 38 ft (H) welded at roof and

(two-story) 8 in. (T) + 2nd floor levels.

W - width of panel, H - height of panel, T - panel thickness •• The thickness of most panels is increased to 16 in. along the vertical edges. ++. Digitized data are not yet available from CSMIP.

Type of Earthquake Diaphragm Records

1984 Morgan Hill Plywood Roof 1986 Hollister

1989 Lorna Prieta

Plywood Roof 1986 Palm Sprin¥s 1992 Landers·· 1992 Big Bear •••

Plywood Roof 1988 AI urn Rock Metal Deck 1989 Lorna Prieta (2nd:fil()or)

~

l.Il 00

Channel Number

1

2

3

4

5

6

7

8

9

10

11

12

13

TABLE 5.2 SUMMARY OF STRONG-MOTION INSTRUMENTS IN THE HOLLISTER WAREHOUSE ~-

Maximum Acceleration Response (g) Predominant Frequency (Hz)

Location Elevation Direction 1984 1986 1989 1984 1986 1989 Morgan Hollister Lorna Morgan Hollister Lorna

Hill Prieta Hill Prieta

Center North Wall Ground Vertical 0.306 0.308 0.177 4.54 3.56 0.29

1/4 Point South Wall Roof EW 0.066 0.139 0.268 0.66 0.83 0.59

Center North Wall Roof EW 0.088 0.139 0.257 0.66 0.81 0.59

Center Roof EW 0.251 0.296 0.818 1.64 1.71 1.10

Center West Wall Roof EW 0.227 0.281 0.718 1.64 1.71 1.10

Center West Wall Midpanel EW 0.208 0.347 0.553 1.64 1.71 1.10

Center West Wall Ground EW 0.058 0.112 0.252 0.66 0.83 0.59

Center North Wall Ground EW 0.079 0.117 0.254 0.66 0.81 0.59

1/4 Point South Wall Ground EW 0.062 0.110 0.237 0.66 0.83 0.59

Center Wes1t Wall Roof NS 0.065 0.147 0.388 0.76 1.22 1.94

Center Roof NS 0.115 0.248 0.445 6.84 6.40 1.90

Center East Wall Roof NS 0.069 0.126 0.349 0.81 1.22 1.90

Center North Wall Ground NS 0.065 0.140 0.361 0.81 1.22 1.90

VI \0

'TABLE 5.3 SUMMARY OF RELATIVE DISPLACEMENT DATA FROM THE HOLLISTER WAREHOUSE

--~----

Channel Number Location Elevation Direction

2

3

4

5

6

10

11

12

1/4 Point South Wall Roof

Center North Wall Roof

Center Roof

Center West Wall Roof

Center West Wall Midpanel

Center West Wall Roof

Center Roof

Center East Wall Roof

Location of Reference Channels: Channel 7: Ground, Center West Wall Channel 8: Ground, Center North Wall Channel 9: Ground, 1/4 Point South Wall Channel 13: Ground, Center North Wall

EW

EW

EW

EW

EW

NS

NS

NS

--

Ref. Channel·

9

8

7

7

7

13

13

13

-----------------_ .. _--------------

Maximum Relative Displacement (in.)

1984 Morgan Hill 1986 Hollister 1989 Lorna Prieta

Unfiltered Filtered Unfiltered Filtered Unfiltered Filtered

0.25 0.06 0.12 0.05 0.60 0.36

0.16 0.05 0.16 0.05 0.38 0.30

0.69 0.72 0.73 0.61 4.89 4.30

0.74 0.67 0.66 0.59 4.86 4.20

0.44 0.37 0.46 0.41 2.71 2.60

0.20 0.08 0.13 0.06 0.57 0.24

0.19 0.06 0.15 0.09 0.58 0.43

0.17 0.07 0.13 0.08 0.36 0.24 - ---------~ L _

0\ o

Channel Number

1

2

3

4

5

6

7

8

9

10

11

12

TABLE 5.4 SUMMARY OF THE STRONG-MOTION INSTRUMENTS IN THE REDLANDS WAREHOUSE

------- -~----- ~----------------- -- ---

Maximum Acceleration Response (g) Predominant Frequency (Hz)

Location Elevation Direction 1986 1992 1992 1986 1992 1992 Palm Landers Big Bear Palm Landers Big Bear

Springs Springs

Center West Wall Ground Vertical 0.027 0.04 0.06 5.62 - -

Center West Wall Midpanel EW 0.094 0.24 0.43 2.56 - -

Center West Wall Roof EW 0.129 0.41 0.75 2.56 - -

C..-enter Roof EW 0.121 0.34 0.69 2.56 - -

1/4 Point West Wall Roof EW 0.242 0.50 0.60 2.56 - -

Center South Wall Roof EW 0.039 0.13 0.16 2.05 - -

Center North Wall Roof EW 0.050 0.12 0.18 2.05 - -Center East Wall Roof NS 0.044 0.12 0.14 2.61 - -

Center South Wall Roof NS 0.113 0.48 0.44 3.15 - -

Center West Wall Roof NS 0.041 0.12 0.13 3.15 - -

Center West Wall Ground NS 0.041 0.12 0.13 2.71 - -

Center West Wall Ground EW 0.046 0.10 0.17 2.05 - ---~------------ --

0\ I-'

TABLE 5.5 SUMMARY OF RELATIVE DISPLACEMENT DATA FROM THE REDLANDS WAREHOUSE

Channel Number Location Elevation

2 Center West Wall Midpanel

3 Center West Wall Roof

4 Center Roof

5 1/4 Point West Wall Roof

6 Center South Wall Roof

7 Center North Wall Roof

8 Center East Wall Roof

9 Center South Wall Roof

10 Center West Wall Roof ~-.---.------ ------ ---

Location of Reference Channels: Channel 11: Ground, Center West Wall Channel 12: Ground, Center West Wall

Maximum Relative Displacement (in.)

1986 Palm Springs

Ref. Direction Channel· Unfiltered Filtered

EW 12 0.12 0.10

EW 12 0.20 0.19

EW 12 0.21 0.19

EW 12 0.34 0.36

EW 12 0.06 0.02

EW 12 0.05 0.01

NS 11 0.02 0.01

NS 11 0.07 0.07

NS 11 0.02 0.01

0\ N

TABLE 5.6 SUMMAI~V OF THE STRONG-MOTION INSTRUMENTS IN THE MILPITAS INDUSTRIAL BUILDING

Maximum Acceleration Response (g) Predominant Frequency (Hz) -

Channel Location Elevation Direction 1988 1989 1988 1989 Number Alum Rock Lorna Prieta Alum Rock Lorna Prieta

1 North End of East Wall Ground Vertical 0.034 0.081 1.64 0.44

2 South End of East Wall Ground Vertical 0.035 0.081 1.64 0.44

3 Center of East Viall Roof NS 0.076 0.114 3.49 0.18

4 Center of North 'NaIl Roof NS 0.174 0.329 4.83 3.69

5 Center of West ~Vall Roof NS 0.077 0.140 3.49 0.18

6 Center of East \Vall 2nd Floor NS 0.072 0.108 3.49 0.18

7 Center of North 'NaIl 2nd Floor NS 0.094 0.165 3.49 0.18

8 Center of West ~Vall 2nd Floor NS 0.074 0.125 1.39 0.18

9 Center of East Viall Ground NS 0.067 0.092 1.29 0.18

10 Center of West ~Vall Ground NS 0.063 0.101 1.29 0.18

11 Center of East Viall Roof EW 0.465 0.585 5.44 4.52

12 Center of East Viall 2nd Floor EW 0.127 0.255 5.18 4.52

13 Center of East Viall Ground EW 0.077 0.139 1.68 0.43

0\ w

TABLE 5.7 SUMMARY OF RELATIVE DISPLACEMENT DATA FROM THE MILPITAS INDUSTRIAL BUILDING

Channel Numhcr

3

4

5

6

7

8

11

12

l..tKation Elevation Direction

Center East Wall Roof

Center North Wall Roof

Center West Wall Roof

Center East Wall 2nd Floor

Center North Wall 2nd Floor

Center West Wall 2nd Floor

Center East Wall Roof

Center East Wall Roof

Location of Reference Channels: Channel 9: Ground, Center East Wall Channel 10: Ground, Center West Wall Channel 13: Ground, Center East Wall

EW

EW

EW

EW

EW

EW

NS

NS

-~---.-- -----~-

Maximum Relative Displacement (in.)

1988 Al urn Rock 1989 Lorna Prieta

Ref. Channel· Unfiltered Filtered Unfiltered Filtered

9 0.02 0.01 0.37 0.03

9 0.09 0.07 0.38 0.18

10 0.02 0.01 0.29 0.04

9 0.02 0.01 0.34 0.02

9 0.06 0.03 0.41 0.08

10 0.03 0.01 0.45 0.04

13 0.21 0.16 0.99 0.26

13 0.06 0.04 0.27 0.07

TABLE 6.1 SUMMARY OF OBSERVED DAMAGE IN TILT-UP BUILDINGS FOLLOWING EARTHQUAKES IN THE U.S. ----- ------------- ---------- ~-

Number of Observations of Damage

1964 1971 1987 1989 Observed Damage Alaska San Fernando Whittier Narrows Lorna Prieta

Buildings Buildings Buildings Buildings constructed constructed constructed constructed before 1973 after 1973 before 1973 after 1973

Damage to connection between plywood and interior 4 5 1 purlin or plywood and ledger

Damage to connection between roof framing member 1 4 12 1 and wall panel

~ Damage to panel to panel connection 2 1 1

Damage to panel to foundation connection 2

Damage observed in wall panels without openings 4 6 6 2

Damage observed in wall panels with openings 1 15

Damage to concrete wall panels or pilasters due to 5 2 3 1 bearing

Damage to roof collector elements 5

Collapse of roof or wall panel 1 4 4 1

No significant damage 13 19 1

Total number of buildings surveyed 2 7 35 46 3 2

f--;:~bt;~~~:;--: I I , ' : Ten,.;gn :

! 1:!Jl1'.<1 L I (

I ~' I ,

! --i ~ .:~ 1

1 °f-·~:'i : I ' , I , I

I r-------'--------, : I : 1...-------1 I

I I

OOUBlE I :/" ANGLE I I

LIFT : : PLATE I I I __ ., I

I I ~ _______________ J

, , , , I \ \ ,

,/' "

r-------------------------, : Singl. Spider ~. ; I .". I I T ens;on Sheor • • I

! d:=J' t.~ II I 0 .' I I 0 I t ". ,....... • L I

I . :Jl:', .:. : . I : I ,,)-_________________________ J

r--------------,

! J":! : . . : : Broce ." 1 I • I I ' ::-. I I 'I ----~ \ a .' I : • • I I I L ______________ J

r-------------, SWIVEL: jb : LIFT: D-"" : PlAn: I :~.~'.; : -----~ .~ , I

! ~~~: ! I I I I L ______________ ~

r---------,---------------, l SQ, HEAD I DIAMETER LENGTH I : COIL BOLT 1 ~H " .. & 6" :

I II I I I '.4" "" & 6" I : : I" AH & 6" t : : 1'4" ""&6" I I I Ilh" 6" I I I I I I I L _________ J _______________ ~

Fig. 2.1 Common arrangement of panel pick points [79].

65

z, , " " ,

" " " " " ....... , " , ....... , ....... ,

....... , ....... ,

,..)

66

.

VariabJe Widths Affect • Stack Casting

• Lift Rigging

• Weight/Crane Size

Avoid • Laterally Unbalanced

• Too Narrow Elements

• Interrupting Support Elements

Fig. 2.3 Examples of improper placement of openings [78].

67

D D Avoid Joints Thru Openings

~D DODD Better

Fig 2.4 Locations of openings within tilt-up panels [78].

68

/'

Fig. 2.5

./' ",

/' /'

.,/ ,/

Roof framing member connected to the tilt-up panels at the panel-tcr-panel joint [78].

69

--l o

Type of Diaphragm

• Wood

E8J Light-gage Steel

I~~~~ Prestressed Concrete

Fig. 2.6 Common types of diaphragms used in tilt-up construction throughout the U.S. f1 0].

Plywood Sheathing

-

Glulam r Beams

...... -

I 111~~~ .r-. 1

-- 11111 I 11111111 -

1111I1I1 I 11111111 .....

filii I 11111111 ~.~ IIH± ~ - ~

'"

I .~

/

V .......

V '-'

'-- Interior I

Fig. 2.7

Columns Conaete WaD Panels

Typical plywood diaphragm [35].

Full depth bridging

Blocking (acts as blocking) (may also be

positioned flatwise)

V 2X Joists

I-- Sawn 4X Purlins

Fig. 2.8 Blocked section of a plywood roof diaphragm [57].

71

Cj

a

72

-..l w

r-

A U

Deck Span

SECTION A·A

o

b+

~ ~ L-eo (l.)

..c: (/)

0 ·c (l.)

.s

l J

r Roof Framing Plan

SECTION B·B SECTION C·C

Fig. 2.10 Plan view of a steel diaphragm [64].

U

~I -

CJ

SECTION 0-0

c L

~ 3: L-eo (l.)

..c: (/)

Channel Type Sheartranz ®

Z-Type Sheartranz ®

1/16" Weld 1" Long Every Flute

Fig. 2.11 Typical Z- and C- type shear transfer connections [64].

74

1/16" Weld 1" Long Every Flute

-....) VI

Neg/fgibl6 End Wall

Defiectlon

0

/ 8-- / -- ---/. -/

~-I /

>-- ___ ...L ___

/ C

w

I 6 T / - __ L-~-

/ / ____ ...L ___

OfOPhrOgmJ ~ W~b ~

K

l--;>Z E / r----y

t' ilo /aphragm lang~s

Transverse (End) Wall

v~ -- -/- _Llr/- --1-;/ / SF/ V

LongItudInal (Slds) Wall

V WL8[ -2- I

MI

H

~ I X

_=r~8F ~

WLBF -8

""'" X

WLBF v max= Maximum Shear 1ft " Transverse Walls = 20

BC 2

T CG or C BF = MaxImum Tt1nslle ar Campresslve Force In Flange W LSF max max BOSC

For derivatfan of 6 T' S6S Sse. J.2.1

Fig. 3.1 Idealized force distribution in diaphragm [33].

h

Fig. 3.2

t

vERTICAL LOA~ LEVER SY STE M _ "

RO LL E R B EA RING '---•• ~3t

PLYWOOD AND TRUSS JOIST BACKING FOR AIR BAG --__ '---~_

TUBULAR STEEL FRAME

LEDGER ANGLE a TIE

TEST SPECIM

TIE

ADJUSTABLE SUPPORTS

/ ..-/

LOAD DRUM (WATER) ../

SIDE ELEVATION OF TEST SETUP

Idealized model of a tilt-up wall panel [16].

Fig. 3.3 Test configuration for ACI-SEASC lateral load tests [16].

76

w

~16---' 3

6--

Fig. 3.4 Free body diagram of a tilt-up panel subjected to lateralloading [16].

77

100'

R = 2Sk ext

~, 2~ , "

-i r- t

--\~-207

;;------------1= n--------------I I Krel = 0.303

I I 18'

II t I I I I , I I 24

: : Krel = 2 +-I I 18'

I I 1-I I I I I I I I I I

I

1------ 100'------1

I

Krel = 0.394

Krej = 0.303

1------------200'-----------i

k R int = 50 R =25 k 11::.

ext I ~ ext . well

-------1 I::.int. well L--------------- -------

~~-

I::. J diaph.

(a) Flexible Diaphragm

R = 4C k , R. = 20 k I R = 40 k 1 ~ I::. _ I::. ext: 10 t I ext I wall- dJeph

------------__ ----~I----------------------~I .

Fig. 3.5

(b) Rigid Diaphragm

Distribution of lateral forces to supporting walls in structures with flexible and rigid diaphragms [65].

78

-...) \0 a

'4 L .1

1 )(

--"--- -)( ..

'C-;d~:~:am Pur/ins

)(

)(

)(

)(

)(

)(

.2.. ____ -

.2.. ____

End Member

Sheet Side/ap

0 ..9.._ 0 --- -

)( )( )( )( )(

Corrugation Lmes )( Structural ConnectIon

o Sidelap Connection

Fig. 3.6 Configuration of metal-deck diaphragm.

w

-+ w

-+ w

~

Fig. 3.7 Corrugations in metal decking [48].

80

Fig. 3.8

Fig. 3.9

q; q;

___ Ll PANEL CENTER LINE t

Strength of metal-deck diaphragm limited by connections along edge of diaphragm [48].

I , , I , ,-,ev :x ! I I -­, I I IN I C» :x

Q a ----..

c.. x

- -- - --- -1----

Q 8 -

Fel

fpJ I I I I_ I , I I I I ,

-r-----1 - -- - --- - ---t­, I

N c..

x CENTER LINE I I I I

---a 8

....--a s

Fp 2 ~---t:11

---a s

I I I I I

t

Strength of metal-deck diaphragm limited by connections in an interior panel [48].

-Sl

~t

I I I I I I -rr----I I I I Q) :x

a 8 ~

I I I I I I I I I I I --7---4-----1

CENTER LINE : I I I I I

n I' v t ' }----~~----( ~----"*"",------1 >-------if-------{

C4 ......--

a 8

~

a 8

Fig. 3.10 Strength of metal-deck diaphragm limited by connections in corner [48].

82

, ENO CLOSURE , , ,

~</Q p/ ',/ lie

EDGE RESTRAINED BY NEXT UNIT

(e) UNIT RESTRAINED AGAINST END WARPING

(b) OPEN-ENDED UNIT

Fig. 3.11 Deformation of metal deck [48].

83

-..C::l -Q) (j ~

~ C5 s I;)

-t ..!:: ........ ~

s::::: I;) (l)

~

Nail Id

A B C 0 £ F G I J K L M N 0

200

150

~

100

50

0

~ 7' V r ~

7'

V ~ ~

~ ~ ~ ~

V ~ ,...

~ V 1/ ~ ;:;; ~

~ v ~ ~

V I ~ I 1/ I

,I I I 1/ I

~ I I V I ,I

,I V ,I

I ~ ~ I I I

~ ,I 1/ ,I

,I I 1/ ,I ~ v ,I ,I

,I I ,I V ~

~ ,I I ~ V I 1/ ~ I ~ 1/ V II

,I V I V 1/ V / 1/

ABCDEFGI

I

I I 7-v ~

7' I /. ~

I I ~

V I I ,I

V ~ ~ I I V ~

V V ~ ~

~ ~ ,I ~ v __ l..I 1.1

rz2l -I I I I ,I

,I ".

I ,I

,I ~

3

1

VB» Plywood

/2» Plywood

J K L M N 0

Nail Iden tification

Type Length Diameter in. in.

Annular Ring Shank 3/4 0.130 24 rings/in., brigh t steel Annular Ring Shank 1 0.105 22 rings/in. brigh t steel

Screw Shank 1 0.150 aluminum Annular Ring Shank 7/B 0.700 22 rings/in., brigh t steel Annular Ring Shank 7 0.740 24 rings/in. brigh t steel

Screw Shank 1 0.720 24 threads/in., aluminum Screw Shank 7/8 0.102 32 rings/in.~ aluminum

Annular Ring Shank 1 0.120 20 rings/in., brigh t steel Staple 1 1/8 16 go brigh t steel

File Shank 1 0.130 brigh t steel Annular Ring Shank 1 1/8 0.115 24 rings/in., aluminum

Steep Spirol Screw Shank 7 0.140 brigh t steel Barbed 3/4 0.145 galvanized

6d Common 2 0.113 brigh t steel

Fig. 3.12 PuIl~ut strength of various roofing nails [20].

84

00 VI

2

J

...... -?'''''

..-'"

............ ././ .............. ~/ ._----

(a) Displacements in diaphragms 1 and J are not compaUble with displacements in diaphragm 2.

[' .......... -_/~ 1"... " ........ _-/

4 6

" .;J

~: ...... --"': drag J I struts 5 I I I

! I I

"- /"- /', ........ _-/ ........ _-/ ...... _-""

(b) With the' addition of drag struts, displacements in diaphragms 4, 5, and 6 are now compatible.

...... ............

no conn(Jctlon b(JtW(JM walls and roof

B

./ ",'"

...... ./ ................. ,/

........ _------,..,.,..

(c) Offset walls should not be connected to the roof if high shears are expected at the re-entrant corners.

Fig. 3.13 Displacement discontinuities in non-rectangular di~phragms [14].

00 0'1

, /

" ~/ , / ......... ~ ". ......

----..._---------,-

staIrwell £/lJvator CarlJ StalrwtJIl

~'i

....... ""--- ........ _-'"

- ",:;"p; " //

" ......... -- ~~// -....-----------,-

(a) Displactlments In dIaphragm are not compatible wUh displacements In stairwells and elevator core.

I I I I , I I I StalrwtJIl £/lJvator CorlJ Stalrw6/1

DD D Dca. D St':fs Struts ____

I .......... I I '-_/ I I I I I I I I I I I I I I I I I I I I I I I I I I I

(b) With the addition of drag struts, displacemen ts are now compatible throughout the dIaphragm.

" '­---StalrwtJlI

" '--"'-....",,---

" "

---~..;'-~

// ..-/

£ItJvotor CorlJ Sta/rwtJlI

--------,-.,..,.

//

..-/

-... ................ ------//

..-/ ,.......-

(c) Without pane/-to-roof connections at the stairwells and elevator cars, the roof diaphragm does not develop the displacement IncompatibilitIes shown in (a) because the diaphragm Is not restrained at these poInts.

Fig. 3.14 Displacement discontinuities in buildings with stairwells and elevator cores [14].

<D I

o 00@ I I I I w = 290 pit

0-~~~~~~T 1/ I I nlulam

@- IT opening B I 16 '

I

o 1 IIV 111/ I 4' 0'- ~-----.-_______ -------J

R 1=6' ! x 8' .}.4- 2 ' 48' ___________ ~ .... t R

(a) Plan view

CD 0 0 0 CD I I 290 pit I 290plf I I

3444C 3444C

0- ~~ ~ - --- -1

Segment ! 1 Segment j /I I

@- J 5m ~ 0 0 - 2~1 - -(326 plf) -4- (181 pll)

0-0 0 - -

1 Segment j 1

Segment

l IV III

3444 T 3444T

0- ~ - --- -- ~ -<--

o plf o pit

(b) Free-body diagram

Fig. 3.15 Analysis of diaphragm modelled as a Vierendeel truss [74].

87

Precs'st panel

g r-t:

fr i;~ ad!?

~I :-1

~r-~t~~~~ I

~i L __ _

4-~6x2'-6'"

dowels to match col umn rein forcing

2'-0"

Bottom of footing level ThicJcness of pad increased to accommodate precast panels

Cold joint, r 1'-3"--'-rg" 'I· '-~l

I' - 3" I 1 l' - 3" i r----------- -~ __ .... I'II--I_~..,. Panel reinforced

6" x 6" /11 I Ox II I 0 at centerline of Welded wall

.t ~--:-O",,;---:~b--;~~::;;;;;;;;~:;$=tl::::..... 114 ~.onlinuoo~

I' ;.<!QUI

It .. dowels €I 24"

at 5 continuous and typical all wall panels

S,ll! il" &RI.!l! 'll:[O l'l[P p.u.EL [~ECT[[

Centerline 1 I ,

~

f1".I'~ Cra".

.' .I

~ I

~ .~i ~,

.:: <'3'

E >\ I

~

Contlnu"",. Strl~

FDDt'''_

Shallow Footing

c

2"

L .. J::((_ J

c

6"x6"/J:\Ox ft lO welded ,/ wire mesh

Slab Inflll

C.fttlftUDUe •• rl~

F_Unw

Deep Footing

Fig. 3.16 Typical panel-to-foundation connections in buildings constructed before 1971 [55,80]

88

P.C. PANEL PER ELEVATIONS

i" CONT. SOLID NON-SHRINK GROUT

FINISHED GRADE PER ARCH.

~OlL I . :... 2'-

;---- 1 #5 CONT.

1/4 DOWELS It 24-0/~

SLAB ON GRADE ---,

0'-0"

$ - ~ - - ... -~--t+.--~~~--' •• ~~ ,

(PER PLAN) ~ (,) __ orre::~=~~ ~--::; ,...

0'-0"

(0

I

, #5 CONT.

P.C. PANEL PER ELEVATIONS ---__

64 {L @ 24·o/c --------­

FINISHED GRADE

PER ARCH. --~

p.

I' ,.

EO. EO.

CONT.

2 #5 CONT.

TOP & 80T.

.....------#4 @ 24"o/c

/'

EO. EO.

PER PLAN

1 #4

COVER w / S MIN. CONC. OVER STL.

i .. CONT SOLID NON-SHRINK GROUT

Fig. 3.17 Typical panel-to-foundation connections in buildings constructed after 1971 [40].

89

Fig. 3.18

Note: Alternate honzoDtal bars to project 4" for column anchor;&ge e 2b" 0. ~. alaximum __ _

J/~'. chamfer (typ. )

l' - 0" L"-----:;1 .... Typical

,~tf-;.J-~i'n , ! I Prec.a:Jt pane:l l

.. . .. =f I

i>j I

~~ -r TYplC<l1 column bars

Horizont.' ReinfOf'cing

.. '. •.

ouaIl0-...----"" DETAIL <!>

ELEVATION

Typical panel-to-panel connections in buildings constructed before 1971 [55,80]

90

C,.,. .... rtc.1 C ..... bo ..... S ..... Ene...... T.p Chor" It.b.r .t .0110 E" .... f ...... 1

4 x i x 0"-6'" WElD It BOn THUS:

~~ PLAN

It 6 x iI x 0'-9· wi .3 1'5 0 Sole

.r P .... I

(HOClf: TOP &. BOT. BARS) ~

~ PLAN

L

L,/:I L/2

EDGE BARS

it 1 X t X 0'-6" wi 2- -" X 4- W.H.S. @ .. o/c-----

Elevation

s.~tlon

L.!J X .3 X ~. X 0'-6-wi 2 - r~ X f) W.H.S. 0 4-o/c

L2~ X 21 X t X

0'-5' ;(LD ANGLE

Fig. 3.19 Typical panel-to-panel connections in buildings constructed after 1971 [40,80].

91

Anchor Bolt

Wall Panel C7

Vv

Boundary Nails

Joist

Ledger

Fig. 3.20 Typical purlin to wood ledger connection in buildings constructed before 1971 [35].

Nails for diaphragm and chord ~ shear to ledger

--r'~~'t- ,I Panelized roof ~ , Pur I ins 8 ! -0 o. c . I ,r ,', ! f

.' tl, " ~~~.-~.--~--= ... ~-.~-.~==.-~-.~.-~--~-~.-~ ... -~-~---~-

Chord Steel---~I-+-'" \~ \

2 x 6 @ 24

\ Bol ts for chord and diaphragm shear ledger to wall

~ 3 x 6 ledger

Fig. 3.21 Typical plywood to wood ledger connection in buildings constructed before 1971 [25].

92

6 ~ ,

'0 \ () "

Chord Steel

Advantages:

Nails for shear transfer to ledger .I

Embedded strap with or without swivel for local forces normal to wall

___ Joist hanger

L __ - Hood ledger

@ 16" or 24"

Bolts for both vertical support and chord and diaphragm shear

a. Few pieces with simple, standard haraware. b. Easy to install with standard carpentry techniques. c. Good, direct shear transfer to wall. d. Accommodates ceiling on underside of joists.

Disadvantages:

a. Embedded straps have to cycle and align with joists and follow slope of ledger which requires setting joist locations at an early stage so joists and straps coincide. Swivel is of dubious merit if place­ment is significantly off as strap will run diagnonally across joists and nailing is partially lost.

b. Hay be adversely affected by shrinkage as strap is permanent in elevation into wall while roof and roofing settle around it. Strap also places bump in roof membrane subject to different thermal behavior than rest of roof.

Comment: This is a good current detail essentially developed after the San Ferando earthquake to provide a positive tie into diaphragm for local forces. There are some field problems setting the straps at the proper elevation and proper lateral location and some concerns over the effect of the straps on the roof membrane.

Fig. 3.22 Typical purl in to wood ledger connection in buildings constructed after 1971 [25].

93

\0 +:a

Wall Anchorage per uec .

Transverse Wall

Positive Dire Connection per UB Sec. 2310 (typica

-

~

t

,

\

-

I

1\

\ ~I/

VI\ --1 1-1-

J

V ,

F

II: I

II l n-

T I 1

: I

rf't.

r 1

\~ 1 ,l

1\ T l

\ ~ 1\

..2..,~

, K J\G / . ______ .4

Sub-diaphra9m per uae Sec. 2513 (typical)

I

I I

I I <J n-

T c5 I:

1 a

: .. I ()J

I

'" (~

r (/) 1 !§ I 1

l..: ~

I Cl... ,l x: 't' '~ 1

l/ v/

V

.. __ J "..

Longitudinal Wall

/ ,'/,'--'/

/ I I

I I / I

1 1 1 1 I 1 I I

I I I

11 I I 1 I 1 I I

n- n-. n-

T y T y T I 1

I I I I I

: I I I : I 2x Sub-pur/ins @ 2'-0" O. C. I

'" n- n- ,... '" i r r r r

I I I I l I 1 1 1 1 : : : 1 I I ,l 1 I 1 ,l

'" '" '" 't' 'I T ..., 't'

I I I I Al :B I l I I I 1><1

I ). I t J I I

D c "/ \ ~ .. - / .

Fig. 3.23 Plan vIew of roof framing members showing subdiaphragm details.

I

a ,I ~ (Q

1 ' .,/.,,~~'-~

::~ L t:.

=<F .;.\ r

"-l""-

I .. " .....

!.,,'>

Ij , "-... , . '"

. I ,.

I .... " - ;; . .-

Nails for diaphragm and chord shear -----;~~~+-~

to ledger

Bolts for vertical support, diaphragm and chord shear and __ _ local forces normal

A

'0

\HUC4 hanger

\ \ '., \--1 I

..... --~ I' : :.---- 4 x bloc k

!d :

I ~\ I

J I

"' \ "7 -~-

~ I

\ HUC4 han ger

PLAN VIEW

/ HUC4 hanger

/ i /~1 II/asher

l ; ~ !lII!!I'I'-=~. /'

- .. ~" f +--- Joists I t1 _ . . t.'l.. ~, or 24"

to wall , \ -Rod

@ 16"

n, \ L Coupling nut -'- 4 x block

l'~; ~I \ L_ Washer

--,T1/~- 3 x ledger

ELEVATION

Advantages:

a. Standard nardware. b. Installs with normal carpentry tools and technique. I.. Independent of joist positioning provided wall bolts are not within-

4-inches of joist centerline_ Tight tolerances not required; d. Good, direct shear transfer to wall~ e. ~ot adversely affected by wood shrinkage.

Fig. 3.24 Strengthening of existing panel-to-roof connections for walls with parapets [25].

95

Bolts chord wall

/Nail diaphragm and chord shear to sill

/ ;;' Joists @ 16" or 24"

~ ;-__ -h~~~--------~=~~~ .• ~~-r/'IRods for outward

local forces normal

tl· ' . • /J..

to wall

Wood ledger and bolts for vertical support

See Figure 9 for details of rod connection

Advantages:

a) Simple connection installed with standard carpentry techniques. b) Can be installed on existing construction almost as easily as on

new.

Disadvantages: Number of small parts resulting in an increased cost.

Fig.3.25 Strengthening of existing panel-to-roof connections at top of wall [25].

Fig. 3.26 Purlin-to-purlin connection across a glulam beam [22].

96

Chord steel for combination of diaphragm and sub-diaphragm tension

Q' .

Nails for combination of diaphragm and sub-diaphragm shear to ledger

Endnail each block for overturning force

plywood to blocking for local\ normal to wall \. ,

~.

LBlocking staggered each side of rod __ snug fit required between joists

'; _' _ Rods for local force normal to wall i

~~-,-- --- SUB-DIAPHRAGM ,,~

.\\-- Coupling nut o

1>. _~_,~L

Advantages:

\ \·--:301 ts for shear transfer to chord/wall \

'-.---- 3 x ledger to match joist de.pth

a. Relatively simple carpentry b. Tolerances are relatively loose. c. Not adversely affected by joist shrinkage. d. Can be applied to existing as well as new construction.

Disadvan tages:

a. IncreasE~c.1 hardware, lumber and nailing relative to (13).

16" r 24"

b. Snug fit on blocking hard to obtain unless rods are tightened before sheathing is in place.

c. Problem in blind nailing plywood into staggered blocks.

Fig. 3.27 Use of metal ties through purlins to create a subdiaphragm (walls with parapets) [25].

97

Nails to blocking for local forces normal to wall

Nailing for chord shear from diaphragm / and sub-diaphr~~_~~~ diaphragm shear. __ . __ -A.

r --------.-....

shears and diaphragm

Chord steel

// /

,/

./ . /'

.... //.-"

----"-' - --

-+--4 __ ----------- ~v---·--· .~---) Sub-diaphragm , Rod for local

I I

'COupling nut

forces normal to wall

2 x blocking snug fit to joists. Stagger opposite sides ,of rod

Fig. 3_28 Use of metal ties through purlins to create a subdiaphragm (top of wall) [25].

98

Chord Steel for combination of diaphragm and sub-diaphragm tension

Advantages:

,<) 4 '-,6

~.

,0

Nails for combination of diaphragm and sub-diaphragm shear to ledger

Embedded strap with or without swivel for r local forces normal to wall

I I ~trap to strap for full local force normal to wall

! I Nails-strap to plywood for proportion of local;

i I I force nonnal to wall /

[ i I Boundary nailing for sub-diaphragm shear ~

I I Sheetmetc ,1 I !~l strap i ------ !

~-~-+-------- SUB-DIAPHRAGM --- -+ flat blocking

~Bolts for shear transfer to chord/wall

~ 3 x ledger to match joist depth

a. Relatively simple carpentry. b. ~ork done from above.

Disadv2 n tages:

a. ~on-stand~rd sheet metal strap. b. Strap has to be set accurately to elevation following slope of roof. c. ~~y be adversely affected by shrinkage as strap is permanent in

elevation into wall while roof and roofing settle around it. Strap also places bump in roof membrane subject to different thermal behavlor than rest of roof.

Fig. 3.29 Use of metal strap attached to plywood and blocking to create a subdiaphragm [25].

99

.Boundary nailing for sub-diaphragm shear----r

, /lj--

.Chord steel for ~,

comb ina tion of............ " diaphragm and subdiaphragm tension

3 x ledger to __ match joist depth

'0

AI

Nails for combination of diaphragm and

I sub-diaphragm chord shears to ledger

/ Nails--To blocking for normal forces---;

/1 /

and blocks \ ---.:; normal forces

strap

I !

I !

SUB-DIAPHRAGM ---------i~ 1 Angle connection for local

normal forces

'L Bol ts for shear transfer to chord/wall

I r

Nails to strap for applicable portion of ~ local normal forces Some portion goes

off top of blocking to diaphragm plywood

Advantages:

a. Relatively simple carpentry wit.hout close tolerances. b. Adjusts to shrinkage.

Disadvantages:

a. {~on-standard angle and strap. b. Nailing to strap overhead - probably requires scaffolding~ c. Requires duplicate nailing to blocks. d. Blocks must be end connected for overturning.

Fig. 3.30 Use of metal strap attached to blocking to create a sub diaphragm [25].

100

-l---'-'---t--- ---

1/4

I" dia_ anch. bolts

-----2-#2 ties (Ii top __ "--_-H--

of a 1\ concrete columns typical

1/-1 Ba lance 112 tiCS ra 12" -Concr,ete column

olumn reinforcing

SECTION A-A SECTION [3- B

Glul:ll11 B(,:l111 Scat and Anchor Delail

Fig. 3.31 Detail of a direct glulam-to-panel connection [55].

101

400

300

200 -= 'C 100 c = 0 Q.

0 Q

~ -100 0

..J

-200

-300

-400

-·0.30 -0.10 0 0.10 0.30

SLIP (inches)

(a) Cycles to a given force level [39].

,.~

1.2

1.0

0.8

0.6

Z o~

:.. 0.2

"C 00 m 0 .0.2

....J .0 •

.01

.0 •

• 1:)

,'2 ·10 -8 ~ -4 -2 0 2 " 6 8 10

Displacement (mm)

(b) Cycles to a given displacement level [26].

Fig. 4.1 Measured force-displacement response of nailed connections.

102

40 l I

30 I 20 ~

i - 10 l z ,

(a) Static load reversals [83].

Envelope ~-

.:JI:. I

~ O+;----~~~~~~~~~~----co I o I ..J ·'0 i

I I

-20 ~ I I

-30 ~_

Cyclic Load-Deflection

Curves

40--~--~--~~--~--+-~--~--~~--~~

-60 -40 -20 0 20 40 60

Displacement (mm)

(b) Static load reversals [26].

Fig. 4.2 Measured force-displacement response of plywood wall panels and diaphragms.

103

a.. ~

0

-I..L. 0

I"")

I

mL-____ ~L_ ____ ~ ______ ~ ______ ~ ______ ~ ______ ~~----~~----~

'-1.6 -1.2 -0.8 -0.4 0.0 0.4 0.8 1.2 1.6 -(0018-003), 1 N.

(C) Quasi-static load reversals [1].

~r------'-------r------Ir------~------r-----~----~~------I

~~----~-------4-------+-------+-------r-------r~~---r--~~

a..

~O~-+---b~~~~~~~==4=~ r-­l.J.... o

w~----~~-----4~~--~~-----t-------t-------t-------t------1 ,

~~----~--~~~-------+-------t-------t-------r-------t------1 ,

~L-____ ~L-____ ~ ______ ~ ______ ~ ______ ~ ______ ~ ______ ~ ______ ~ -4 -3 -2 -1 0 2 3 4

-(0078-005), IN.

(d) Dynamic loading [1].

Fig.4.2 (cont.) Measured force-displacement response of plywood wall panels and diaphragms.

104

1"'\ CI) t ~ F2 al 0.60"

.~----------------------------~ CD

..r:. o c

v

f­Z

0.8 0.6 0.4 0.2 0.0

~ -0.2 w -0.4 ~ -0.6 --1' a.. -0.8

..-.------~

~ - 1 . 0 E;..,a-&, ........... "'-'-I~..&...I....IL...L..L.~r-.a...a...L..a-."'-'-I~..I...I-......... ___ .i-I...II---.....-..

Q 0.0 0.5 1.0 1.5 2.0 TIME (seconds)

(a) Initial displacement of 0.6 in.

20 F3

F2

0~~~~~~~~~~~~~~~~~ o 1 2 3 4

DISPLACEMENT (inches) (b) Force-displacement curve for monotonic loading to failure.

Fig. 4.3 Measured free-vibration response of plywood diaphragm [39].

105

PROGRAMMABLE ACTUATOR (TVP) 4" x 18" (102 I1I1l x 457 11I1l)

SERVO VALVES (2) 25 GPM (95 LPM)

LOAD CELL 30,000 LB (133 KN)

REMOVABLE REACTION PILLARS FOR QUASI­STATIC TESTS (TVP 4 PLACES)

DIAPHRAGM UNDER TEST

NOTE: THE I-TON (907 kg) LEAD WEIGHTS SHOWN ATTACHED FOR THE DYNAMIC TESTS REMAIN ATTACHED FOR THE STATIC TESTS, BUT INDUCE NO INERTIAL FORCES DUE TO THE SLOW TESTING SPEED.

LOW FRICTION ROLLER ASSEHBLIES (TYP 8 PLACES)

PROGRAHHABLE ACTUATOR

(a) Quasi-static tests.

PROGP~W~~LE AC~JATOR (TYP) 4" x 18" (102 .. x 457 Iml) SERVO VALVES (2) 25.GPM (95 LP~)

I-TOil (907 kg) LlAD WEIGHTS SUPPORTED ON LOW FRICTION ROLLERS (TYP I t:.A4 30 PLACES)

DIAPHRAGM UNDER TEST

PROGRAHHABLE ACTUATOR ~

(b) Dynamic tests.

Fig. 4.4 Test configuration for ABK diaphragm tests [1].

106

LOW FRICTION ROLLER ASSEMBLI ES (TV? 8 PLACES)

~r-------r-------.-------r-------r-------r-------r-------r------.

~r-------r-----~-------t-------t-------r----~~~~Jtt-------l

OOr_------r_----~~~_+~~------r_------r_------+_------+_------, I

~r-------r-~~~~------+-------r-------r-------+-------+-----~ I

."..

~~------~------~------~------~------~------~------~----~

CD

-.r N

-2.0

I

-8

-1.5

-6

-1.0 -0.5 0.0 0.5 1.0 1.5 2.0 -(0018-003). IN.

(a) Quasi-static load reversals.

/I I "'-<~ V / ~

~'

" 'v--

-4 -2 0 2 4 6 8 -(OOlS-OOS)' IN.

(b) Dynamic loading.

Fig.4.5 Measured force-displacement response of metal-deck diaphragm [1].

107

0 EXT£IIICAl. DOF

0 I NTEP:HAl SPRINCi t:J+ COMP. o -TENSION

~ I NTE 1IMAl. DNf'E R I + COMPo 0 - TENSION

DIS I' L.AC£I1ENT IOAIiE

Fig.4.6 Representation of plywood diaphragm as a series of nonlinear springs [11].

F COMP., +

64----------r~~~--~~---------6

TENS •• - ~ COtIP ....

PERMANENT SET 0.5 6

l TENS •• F

Fig.4.7 Initial hysteresis rules for plywood diaphragms [11].

108

0

x UJ u ex: 0 u..

0

x UJ U cr:: 0 "'-

12

8

4

0

-4

-8

-12 -16 -12 -8

F=----

-4 o 4 DEFLECTION

F u

--+e Kl

8 12 16

Force-deflection envelope of model

12

8

4

0

-4

-8

-12 -16 -12 -8 -4 0 4 8 12 16

DEFLECTION

Typical cyclic load-deflection diagram for model

Fig. 4.8 Force-deflection envelope and hysteresis rules for plywood diaphragms developed

after ABK tests [4].

109

It a. i .; ~ 0 u.

., a.

'S2 to-

~ 0 u.

zo

20

10

0

-10

-20

-6 -301---~--'---'---1r--'---~--'---'---'---'---'-~6

o -4 -2

20

10

O

-10

-20

Doformatlon. In

Xi' kips/in.

'Ux/~

~; inches

Fpinch' kips

Fu' kips

.,

Deformation. in

Fig.4.9 Revised force-deflection envelope and hysteresis rules for plywood diaphragm [9].

lIO

-2

. z-

o t-... 0-o·

Mid-Pont. Olaph. N

Tlme.s.o

(a) Calculated response .

nl~ __ ~!~~t __ ~, __ ~! __ ~f __ ~I t ~ 5 10 15 20 25 30

T1HE. SEC.

(b) Measured response.

Fig. 4.10 Comparison of measured and calculated displacement response of diaphragm N [9].

III

400~ ________________________ ~

'" ~ 300 c :J o a..

v

o « o ..J

200 .... -.-.

100

0.0 0.1 0.2 0.3 0.4 DISPLACEMENT Cinches)

Fig. 4.11 Best-fit backbone curve through load-displacement data for nailed connections [39].

112

25 W3 · • I-

20 ~ A I-

0 It Q) .. 0

.~ 15 l-I- nA

..:l v

0 « 0 ....J

· 0 ·

10 · D - • 0 .. 0 It D ~ Model

5 It ~~ .. 0 Experimental IoD ..

(21 L"~~ __ ~~.~·~·~ __ ·~·~· __ L-I~~. __ ~ •• ~I ____ .~ __ ~~

0.0 0.5 1.0 DISPLACEMENT

1 .5 Cinches)

(a) Finite-element model developed by !tani and Falk [83].

30 ..... \ - --"\ \"

\. -- \ ..... Z ~ \ --- 20 \ 0 '-

< 0 .....J

-- Pl,.,.-ood Sheathed Wall Test

10 Wafer-board Sheathed WaIl Test.

** SBlrJ.ll for Plywood Sheathed Wall

• • SHWALL tor W'aferboard Sheathed WaIl

DISPLACEMENT (mm)

(b) Finite-element model developed by Dolan [26].

2.0

Fig. 4.12 Comparison of measured and calculated response of wall panels.

113

...... ...... ~

W33 x 21,1 BACK STOP STEEL BEAH

CONCRETE FOUNDATION

ILEDGER WEIGHT

~r-----'----'i

.. ',-:'0' ! •. '

SERVO VALVE I>·.':··::J /'" CONCRETE .'j:.' TILT-UP WALL ACTUATOR :,':1: 3'-6" x 20' )( It"

''':.

.....

.. :~i::/ \.;.

\it,: .,'\4

\'~.~ ~~i;:::\

if,.

\\ BASE PLATE

TOTAL NUMBER

2 1 9

F: LOAD D: DISPLACEMENT V: VELOCITY

TOTAL 12 CHANNELS

Fig, 4.13 Test configuration for dynamic testing of tilt-up wall panels [6,8,7].

---l 8

GAGE LOCATION 8 a 1/8 HEIGHT

9 9

8 ------ 0.20g PANEL 2

7. 7 --0.20g PANEL'

(, 6 --- o.AtOg PANEL 2 (SEQ. 6)

'" z !:!

5 o .AtOg PANEl 4 (SEQ. 6)

l- S ~ ••••••••••••• 0.20g PANEL 3

4 ---- 0.409 PANEL 3 (SEQ. 6)

000

(a) Displacements, in.(mm) (b) Accelerations, g

Fig. 4.14 Comparison of maximum response of panels 2, 3, and 4 with a rigid roof diaphragm

subjected to El Centro base motion [6,8,7].

~ ~ ..:; .c

9 I

8 \ , ,

7~ , ,

:\

\ \ \ ... \ z

0 .. ~ J 12.20

~ I • "0 I

I I

) I I

I .0

~ '0 :-

i- ~ ..; .,..

(a) Displacements, in. (mm)

00

.••••••••••••• 0.209 (SEQ. 2)

--- 0."<>9 (SEQ. 6)

----- 0."<>9 (~Q. 9)

(b) Accelerations, g

Fig. 4.15 Displacement and acceleration distributions in panel 3 with a rigid roof diaphragm

subjected to varying intensities of the EI Centro base motion [6,8,7].

115

(a) Displacements, in. (mm)

(b) Accelerations, g

--c-13.2sec

--- t - 1. Me

----- t • 16.' sec

-- t • 21.2 sec

(c) Moments, in.-lb (N. m)

Fig. 4.16 Displacement, acceleration, and moment distributions in panel 4 with a rigid roof

diaphragm subjected to varying intensities of the EI Centro base motion [6,8,7].

TILT-UP WALL PANEL (rYP I CAL)

(a) Plan of tilt-up building.

WOOD ROOF DIAPHRAGM

DIRECTION OF EARTHQUAKE

HOTION /

Fig. 4.17 One-story tilt-up building used to develop analytical model [4].

116

~ROOf DIAPHRAGM

t

H • 20·

(b) Cross section.

SIDE EARTHQUAKE ~PUT./ "'.<lLL HOTION /"

o

(c) Conceptual model of building.

Fig. 4.17 (cont.) One-story tilt-up building used to develop analytical model [4].

117

INPUT

EARTHQ.UAKE INPUT MOrlON

(d) Representation of building with linear beam elements and nonlinear diaphragm elements.

o o

I~

'AJeC. l

(e) Model of building used in analyses.

Fig. 4.17 (cant.) One-story tilt-up building used to develop analytical model [4].

118

N U .. ... :? i3 I-

~ <C

.... l-

i <C

.... .., .. ......

% 0

~ ..J .., u ..... <C

~ V> ." <C

60

40

20

-20

-40 .

-60 o

60

Ato

20

o I

-20

-ltO .

n

fI f\ 1\ f\ 1\ fI '/\ r v llVVVVV !V

.~

. 0.8 1.6

.1 /.

3.2 TIHE. sec

A

\

4.0

A

V

V

4.8

(a) Acceleration at mid-height of panel.

D-Q-F g L I GHTL Y DAMPED

~ ~

5.6 6.4

D-Q-F 11 LI Gt:ITL Y DAMPED

-60+----------.------------~--------~------------r---------o----------.------------~--------~ o 0.8 1.6 2.1; 3.2 4.0 4.8 5.6 6.4

TIME. sec

(b) Acceleration at top of paneL

Fig. 4.18 Calculated acceleration response of center longitudinal wall panel

for lightly-damped model [4].

119

N u II ... ~ ... % c ;::

~ ~ < 101 I-

~

~ <

N U II

24

O-O-F 8 10: DAMPING

16

8

0

-8

-16

-24+------.-------r------~------~-----r------~----~------~ 0.8 1.6 2 • .!t 3.2

TIME, sec .!t.0 4.8

(a) Acceleration at mid-height of panel.

32 -,.----,,..---.--.------ ..... ---.....

24

5.6

D-O-F 11 lot DAMPING

6 • .!t

:? 16

:z: c

TIHE, sec

(b) Acceleration at top of panel.

Fig. 4.19 Calculated acceleration response of center longitudinal wall panel

for moderately-damped model [4].

120

---- CASE 1 LIGHTLY

1\ DAMPED ' ........

" -- CASE 2 '" 10~ DAMPED

\ kip-ft/ft

klp-ft/ft

0.52 9

~O.97 9 r---r----:'\

/I.

\ \ \

YO.59 9

/y1.39 9

/ /

/

I . I

(

I /

I

o 2.5 5.0 0 2

MAXIMUM MOMEJ.4T, klp-ft/ft

Absolute maximum moment at midspan TUW

MAXIMUM ACCELERATION, 9

Abf!olute maximum acceleration at midspan TU\-l

Fig. 4.20 Calculated distributions of acceleration and moment along the height

of the center longitudinal wall panel [4].

121

SIDEWALl

(a) Idealized tilt-up building.

I! w

(b) Model used for linear anal yses.

(c) Model used for nonlinear analyses.

Fig. 4.21 Analytical models of one-story tilt-up building [53,54].

122

!­Z UJ (j ii: 11. ill o o o <{ o .-J .....J « rc Ul I-

:5

--, r----...... 1.0

~- T.o.=D.l 52 5% DAMPED

". Tw:tn=OA s5 -------------- -~ :::::::-~--.- ----

0.9'

0.8' .

0.7' . O.G··

0.5 ..•.•... ~ .. ~ ... T w:\R=O.2 s

". Tw:\n=O.4 s OA" 10% DAMPED

Tw:tn=O.4 s 0.3·-

Tw:\u=0.2 s 0.2' .

SIMPLY-SUPPORTED WALL 0.1 . - FIXED-BASE WALL

0.25 0.5 1.0

ROOF PEnlOD [SECONDS]

Fig. 4.22 Calculated roof-to-panel connection forces [53,54].

"- ........ .

- "- ..... ,

100[ ...... -. .. .......... ~

~ 75 . "- ANAL YS/~ ~ •.. '" ".

~ SIMI'L Y-SUpponTEfj-W;",="L ". '. tii FIXED-BASE WALl. '. ~ "- ~

~ 50 . PROPOSED .....

o "-L!... cr: US :r: CJ)

lJ... o <.) tr:

2S .

o (l

Troof = 1.0 SEC

"f"­LlNEIIR /

VllnJlITION

U4

DIS7iiNCE FROM SUPPORT

U2

Fig. 4.23 Calculated distribution of shear forces in the diaphragm [53,54].

123

Fig. 4.24 Analytical model of Hollister warehouse for input motion in the transverse direction [9].

124

(a) Calculated response.

E 1.1

o '0 20 .30 "0

Time. Sec

(b) Measured response.

Fig. 4.2') Relative displacements at the center of the diaphragm in the Hollister warehouse

during the 1986 Morgan Hill earthquake [9].

125

1/,.021'

27.46'

27.083'

27.875'

27.083'

27.292'

14.1875'

I

~L 1122:25

1 :.

~ - -~ . _. -- -..... -. . -. ··II!"! ...

I

l-- - _. -- --- ._-

7 . _- .- .. - -_ . -IIFt- -

~ 10'

I " 122: 25'

,"pi, ~_5".·~. 5.5" THICK

THICK ..

GIRDERS " -6 PIPE COLUMNS (TYP)

DOWNEY 13UILDING

.24.25' 28 1 I < 36 I Of 56 I .c 56 I 56' < 56' " 56 I 56' 28 '

~ . A., ;;( ::if • '" %

T ! j I -1

:--,....;..

- -- ._- - -

~- - - - ~

/ 5.5" T1I1CK

-r~

12E ' .,#-

i24'

j24'

70.83 'i I 1 : .... V T24 I -GLU-Lk~. 1---+-__ ;...1 _-..;. __ ...l.L __ -:--_-+-_...::;..L-., * G I r.D:::::.S

I 1 i I ;24 ' ~

I ! i i 24'

~ ill t24' ..,p;.~.\ ill i24'

t\ .. l--4l---... l-____ 1-l.--___ -... 124 I ';J, \ 125 '

:,..: / .,' .. · I '\- 146 '

L6.5" THIC~K-P-A-1\-::L-S-l-:~.,...~-P-) ----.t,:-\\ \6" SQ

I

TUSE CGL""';

. (TY?)

OFF I CE ANNEX

WBITTIER BUILDING

Fig. 4.26 Plan views of tilt-up buildings in Downey and Whittier [9].

126

I--' N -..l

tll'/ltllT 89 j'i-~ . nr.r1ttlT GC \ ,//" "-

-,-/

Dll\rHRI\GH ELEHENT

VII ACCELERATION INPUT

~~ NOD[S (25 NODES) I~~.

~~.

INPLI\NE Wl\lL ELEMENT ?

VSOll [LEMEN! ~ ~ ~~~~CTION OF ~T HOTION

Fig. 4.27 Analytical model, of Downey building for input motion in the transverse direction [9].

48' 11

70.83 30 123 1 CD ~

,; t t • N

CD

Fig. 4.28 Analytical model of Whittier building for input motion in the transverse direction [9].

128

129

....... w o

• C)

J

~

L • to')

J OJ -• C)

J Ol -• C)

.. I Ol -• to

Ref. N A (N 25 W)~

7" Concrete Wall Panel 7" co:crele W:" Ponel ~ ~ [6" concrOI: WOI: :nOIS (1:/COi) r r r r r

.~ Glu;am 8eams :. : ! •

~ ~ ~ 8" ~ STD P/~e Column ~ l l : ::::: : : : . ..... . . . + .................•. H .. : ..... [.:..+H. . ........ H+........ ....+... ..... + ................... .

· . · . · . · . · . · . · . · . · . { 112' x 14' D.H. Door C\1

· . · . · . · . · . · . 12' x 14' D.H. Door

• Ol J ~

· . · . · . 7" Concrete Wall Panel ~ ~ 7" Con~rete Wall Panel · . · . · .

12' x 14' D.H. Door 12' x 14' D.H. Door " L,c :r ~ 6 6 U \~- ~ 6 C J~

23'-3" 14 Panels «I 18'-0" = 252'-0" I ~-----------------------------------------300~O .. ------------------------------------------~

Fig. 5.2 Floor plan - Hollister warehouse.

,....... w ,.......

J 1.0 .I ~

1 .~

t- 2J'-J.

\4"XI4"

I

7 Bays 0 J6'-O· .... 252'-0"

I • • • • ~GIUlam 8~ams

(8" If STD J Cl-;'

Pip" Columns I.£: ~~ r 6· CancrtJttJ Slab 1

~ ..... '---' ~ L...--I L...--I L...--I

FactJ of Pllasier

Facti of Concrtlitl Wall Pontll

JOO'-O'"

Longitudinal Section

6" Concrete Slob

~1/2" Plywood SheathIng 2'"x4" Rafters @ 2'-0" DC 4"x10" Pur/ins @ 8'-0" DC

Glulam Beams

8· If STD PlptJ Column

t-------100'-0'"-------I

Transverse Section

Fig. 5.3 Cross sections - Hollister warehouse.

2J'-J'",

• • J Cl ,I g

1 L.--' "----'

..­w N

CHORD BAR SCHEDULE

PANEL LOCA nON 1 BAR SIZE

TransverstJ Wall Panels - j5 Bar

End LOnl;ltudfnal Wall Panels I 1 - 19 Bar (8 total)

MiddltJ Longitudinal Wall PantJls 12 - #9 Bars (24 total)

I ' 17'-O~ 'I \ I ~ \I 3'-0-

L '\ - '\

~Chord 80r '\1111111 ~

\4 I I I II

\lJJJL 14 Bars at 12H OC

Each Way

31'-9 1/2- 29'-9 1/2-(typIcal reInforcing In each panel)

1 Finished Floor

--'-----I- r 1 - 16 Bar 1~6~ I - -------- -----------,-

TW pvs. m

Chord bars: Lapped and welded

I 17'-O·----~

\ 1 \1 '\

'\

\ \.

\

\.. "'-t:=:t:J

2- #5 Bars

V

~ I'-- J'x7' NtJt OpenIng

t--- --------------

Bend In alternate panel horizon tal bars.

Twical PaMI/PilasttJr ConntJction OtJtall

1-----17'-0----1

\ 1 \1 "\

'\

\

\

\ \

\.

2-#5 Bars~ ...... '-'C.,

12'x14' Net OpenIng

t-- ------------- r----

Fig. 5.4 Typical panel reinforcement details - Hollister warehouse.

I--' v'> W r-.

- ~

1 ~ ~ ~ ~ IJ n n Eost Devotion

1 0 ~ ~II~ ~ r~ ~ ~ 0 ~II~ ~ ~ 1_ We'sf Elevation

North Elevation South Elevotlon

Fig. 5.5 Elevations - Hollister warehouse.

I~

13 8 o

1-

Fig. 5.6

o

Glulam Beams

Tilt-Up Wall Panels Steel Pipe Columns

-. . .

. . . . . i!J .. : .. 1. .... ~ ..... 1. ....

300'-0n

Roof Plan

Steel Pipe Columns

o o

Slab Plan

0

Elevation West Wall

·····0····· ·····0

0 0

t a I

a

9 t a

~ -I

Sensors 4 and 11 are mounted on glulam beams. Sensor 13 is mounted on the floor slab. All other sensors are mounted on the wall panels.

Locations of strong-motion instruments - Hollister warehoUse.

134

Hollister Warehouse

0.4 1984 Morgan Hill Earthquake NS

0.0

-0.4 0.4 1984 Morgan Hill Earthquake - EW

0.0

-0.4 \)) 0.4 1986 Hollister Earthquake - NS

... c::

0.0 0 "-....... \) ~ -0.4 Q) -Q) 0.4 1986 Hollister Earthquake - EW () ()

~

Q) 0.0

CI) \)

-0.4 CtJ 0.4 1989 Lorna Prieto Earthquake - NS

0.0

-0.4 0.4 1989 Loma Prieto Earthquake - EW

0.0

-0.4 Time, sec

0.0 10.0 20.0 30.0 40.0 50.0 60.0 I I I I I I I

Fig. 5.7 Measured ground accelerations - Hollister warehouse.

135

......... w 0\

Hollister Warehouse 1984 Morgan Hill Earthquake - NS

t)) O. 5 r--r---~-~-~--r-----'

~"0.4 '-...... tJ

~ 0.3 Q) l.) <J ~ 0.2

~ l.) 0.1 Q)

~

2%

I ------- 5% -"-"- 10% ----- 20%

O.O~I ----~--~--~----~--~--~ 0.0 1.0 2.0 3.0

.5.0 .S ........ ~ 4.0 Q)

E ~ 3.0 t1 -Q..

~2.0 15 ~ 1.0 <J Q)

Q.. Vj 0.0

0.0 1.0 2.0 Period, sec Period, sec

1984 Morgan Hill Earthquake - EW 0.5 .1 ---,---r-----y-----r----.------.

t))

~ 0.4 '-..... e ~ 0.3 Q) l.) tl ~ 0.2

e t 0.1 Q)

~

2%

1 ------- 5% _00_00- 10% ----- 20%

o. 0 ''--_----'' __ ''---_--'---_---1.. __ "'"-_--'

0.0 1.0 2.0 3.0

.5.0 ·S ........ ~ 4.0 Q)

E: ~ 3.0 t1 -Q..

62.0

-t)

~ 1.0 <J Q)

Q.. Vj 0.0

0.0 1.0 2.0 Period, sec Period, sec

Fig. 5.8 Linear response spectra - Hollister warehouse.

3.0

3.0

~

w .....

Hollister Warehouse 1986 Hol/ister Earthquake - NS

1.00 ..--, --.--,.----.--.-----r-i t})

.5.0~i --~----~--~--~--~----.

c: .~ 0.75 ...... e Q) -~ 0.50 u "'(

-0 .b 0.25 (j Q)

2% ------- 5% - - - 10% ----- 20%

Q

(fj O. 00 I ,--.....~ ... , T:::2 ?1 0. 0 1. 0 2. 0 3. 0

Period, sec

·S ",--,"

c: 4.0 Q)

t: Q) u 3.0 CJ Q cS 2.0 -CJ .b 1.0 u Q)

1986 Hol/ister Earthquake

1.0 2.0 3.0 Period, sec

EW 1. 00 ~I ----r--,.----~---.--___r-___,

t)) .5.0~i --~----~--~--~----~--~

c:" ~ 0.75

~ Q) ~ 0.50

"'(

2% ------- 5% -"--"- 10% ------ 20%

·S ........ c: 4.0 Q)

t: Q) u 3.0 CJ -Q

cS 2.0

e 025 y\'~ 0 t· :'\'..'\{~/\ .b 1.0 Q) '\',,,- \1 (j

Q "~~--~~'''''~'' Q) U) ~~'~~.:-.... Q 0.00 ---- <.rj 0.0' ....- iii

0. 0 1.0 2. 0 3. 0 0.0 1. 0 2. 0 3. 0 Period, sec Period, sec

Fig. 5.8 (cont.) Linear response spectra - Hollister warehouse.

th

c:: a .-~

\) ~ Q) --Q)

0 0 ~

Redlands Warehouse 1986 Palm Springs Earthquake

0.2 Channel 8 Root East Wall

0.0

-0.2 0.2 Channel 9 - Root South Wall

0.0

-0.2

0.2 r

0.0 l -0.2

Channel 10 - Roof, West Wall

.,.& ~ i ... f, no.~ , .,J. 4¥r~. A - A ... • •

0.2 Channel 11 - Ground, West Wall

0.0

-0.2 Time, sec 0.0 5.0 10.0 15.0 20.0 25.0 30.0

I ~------~!------~!--------~------~I------~~----~!

Fig. 5.26 Redlands warehouse -1986 Palm Springs earthquake.

(b) Longitudinal acceleration response.

171

Hol/ister Warehouse 1989 Loma Prieta Earthquake - NS

2.0 .25.0 t1) ·s c: 2% ..........

~ 1.5 ------- 5% c: 20.0 Q)

_,,_00- 10% E: CJ ----- 20% \... ~ 15.0 Q) -Q) t:l

~ 1.0 -Q

~ £:S1G.0 -tJ -.!:J 0.5 tJ \... 5.0 0 .......

Q) 0

~ Q)

0}-0.0 0.0 0.0 1.0 2.0 J.O 0.0 1.0 2.0 J.O

Period, sec Period, sec 1-4 w 00 1989 Loma Prieta Earthquake EW

2.0 .25.0 t1) ·S C: .. 2% ....... .. :8 1.5 ------- 5% c: 20.0

Q.)

0 _00-00- 10% E

\... ----- 20% ~ 15.0 Q) -Q) tJ

~ 1.0 -Q

~ £:S 10.0 -tJ -~ 0.5 tJ ~ 5.0 0 .......

Q) lJ

~ Q)

Q 0.0 (J) 0.0

0.0 1.0 2.0 J.O 0.0 1.0 2.0 J.O Period, sec Period, sec

Fig. 5.8 (cont.) Linear response spectra - Hollister warehouse.

C))

c:-o '-"""-' \) ~ Q) -Q) \.) \.)

~

Hollister Warehouse 1984 Morgan Hill Earthquake

0.3 Channel 2 Roof, South Wall

0.0

-0.3 0.3

Channel 3 - Roof, North Wall

0.0

-0.3 0.3 Channel 4 - Roof, Center

0.0

-0.3 0.3 Channel 5 - Roof, West Wall

0.0

-0.3 0.3 Channel 6 - Midpanel1 West Wall

0.0

-0.3 0.3

Channel 7 - Ground, West Wall

0.0

-0.3 Time, sec

0.0 5.0 10.0 15.0 20.0 25.0 30.0 I ~----~I~----~I------~I------~I------_I~----~I

Fig. 5.9 Hollister warehouse - 1984 Morgan Hill earthquake.

(a) Transverse acceleration response.

139

C))

c:: a '--i-J

C) I..... Q) --Q) C,j C,j

"'Z

Hollister Warehouse 1984 Morgan Hill Earthquake

0.3

0.0

-0.3 0.3

0.0

-0.3 0.3

0.0

-0.3 0.3

0.0

-0.3

0.0 !

Channel 10 Roof, West Wall

Channel 11 - Roof, Center

Channel 12 - Roof, East Wall

Channel 13 - Ground, North Woll

Time1 sec

5.0 10.0 15.0 20.0 25.0 ! J I I ,

Fig.S.9 (cant.) Hollister warehouse -1984 Morgan Hill earthquake.

(b) Longitudinal acceleration response.

140

30.0 ,

Hollister Warehouse - 1986 Hollister Earthquake

0.3 Channel 2 - Roof, South Wall

0.0

-0.3 0.3 Channel 3 - Roof, North Wall

0.0

-0.3 0.3 Channel 4 - Roof, Center

C))

0.0 .. c:: a ~ -0.3 ~ 0.3 Channel 5 - Roof, West Wall

--Q) ()

0.0 ()

""Z

-0.3 0.3 Cha nel 6 - Midpanel, West Wall

0.0

-0.3 0.3 Channel 7 - Ground, West Wall

0.0

-0.3 Time, sec 0.0 5.0 10.0 15.0 20.0 25.0 30.0 I ~----~I------~I------~I------~I------~I------~I

Fig. 5.10 Hollister warehouse - 1986 Hollister earthquake.

(a) Transverse acceleration response.

141

Hollister Warehouse 1986 Hollister Earthquake

0.3 Channel 10 Roof, West Wall

0.0

-0.3 0.3 Channel 11 - Roof, Center

\J)

0.0 '" c::

a :g -0.3

~ ~. ~ r. .111 •. ~~~~ne/ 12 ~ Roof, East Wall (-, u u ['LWhL'''.J...!~~WI.w..l\ I O"Unt/0 A. '" 1\'~_!1 1\, A Q 0 ~ O.a _. -v • --, ... ..,.~rl' "rv ""'7ji 1\1 QYFrf- crv' '1V v.v zoe '\F 0 0 u us v us

~

-0.3 0.3 Channel 13 - Ground, North Wall

Time, sec -0.3

0.0 5.0 10.0 15.0 20.0 25.0 I I I I , I

Fig. 5.10 (COI1t.) Hollister warehouse - 1986 Hollister earthquake.

(b) Longitudinal acceleration response.

142

30.0 I

Hollister Warehouse 1989 Loma Prieto Earthquake

0.8 Channel 2 Roof, South Wall

0.0

-0.8 0.8 Channel 3 - Roof, North Wall

0.0

-0.8 0.8 Channel 4 - Roof, Center

\))

0.0 c: a :g -0.8 ~ 0.8 Channel 5 - Roof, West Wall -Q) \.)

0.0 \.)

~

-0.8 0.8 Channel 6 - Midpanel, West Wall

0.0

-0.8 0.8 Channel 7 - Ground, West Wall

-0.8 Tirne, sec

0.0 5.0 10.0 15.0 20.0 25.0 30.0 !~----~!------~I------~!------~! ______ ~!~ ____ ~!

Fig. 5.11 Hollister warehouse - 1989 Lorna Prieta earthquake.

(a) Transverse acceleration response.

143

Hollister Warehouse 1989 Lama Prieta Earthquake

c: a

0.8

-0.8 0.8

:g -0.8 ~ 0.8 --Q)

Channel 10 Roof, West Wall

Channel 11 - Roof, Center

Channel 12 - Roof, East Wall

(J (J O.O~----~~~~~~~~~~~~~~--~~--~--~ ~

-0.8 0.8

-0.8

0.0 !

Channel 13 - Ground, North Wall

Time, sec

5.0 10. 0 15. 0 20.0 25.0 ! I I t I

Fig. 5.11 (cant.) Hollister warehouse -1989 Lorna Prieta earthquake.

(b) Longitudinal acceleration response.

144

30.0 I

Cb -a :::i

..........

Hollister Warehouse - 1984 Morgan Hill Earthquake

~ 1. 0 f Channel 4 Roof, Center

~ 0.5 ~ Cb o. 0 0' --.,..-.-.~

:::i

~ ~: ~ f ,~Chann~5 ~:=t Wall

~ 0.0~_6~~~~~~~~~~~~~~~~~~--~--~--~ E

~ iif ll~b' ~: ~ f·. I! L • dCh~ o. 0 - "=ooa.J

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 B.O 9.0 10.0 Frequency, Hz

Fig. 5.12 Hollister warehouse - 1984 Morgan Hill earthquake.

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

145

Hollister Warehouse - 1984 Morgan Hill Earthquake

~:~f o.o~--~--~--~--~~~~--~~~~~~~~~~

o. 0 1. 0 2.0 3.0 4. 0 5.0 6.0 7.0 8. 0 9.0 10. 0 Frequency, Hz

Fig. 5.12 (cont.) Hollister warehouse - 1984 Morgan Hill earthquake.

(b) Normalized Fourier amplitude spectra of longitudinal acceleration response.

146

Hollister Warehouse - 1986 Hollister Earthquake

to ' z:"=zo<"'> d

~:~ f JAi, Cha;.~e~ [ - Ground, West Wall

0.0-0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0

Frequency, Hz

Fig. 5.13 Hollister warehouse - 1986 Hollister earthquake.

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

147

Hollister Warehouse - 1986 Hollister Earthquake

~:~ f .J Channel 13 - Ground, North Wall

QO~ O. 0 7. 0 2. 0 3. 0 4. 0 5. 0 6. 0 7. 0 8. 0 9. 0 1 O. 0

Frequency, Hz

Fig. 5.13 (coot) Hollister warehouse -1986 Hollister earthquake.

(~) ~ormahzed Fourier amplitude spectra of longitudinal acceleration response.

148

Hollister Warehouse - 1989 Loma Prieta Earthquake

~:~ o. 0 ~ t"c-<>'

~ ~.~ :::~~U ::::J

.......... '-! ~:~ f f, Center

Q)on~~ .~ • L./

::::J

] ~:~ st Wall

a o.o~~~~~~~~~~~~~~~~~~~~~--~

E a <::

~:~ f ,t W~/: ~ O. 0 1. 0 2. 0 3. 0 4. 0 5. 0 6. 0 7. 0 8. 0 9. 0 10. 0

Frequency, Hz

Fig. 5.14 Hollister warehouse - 1989 Lorna Prieta earthquake.

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

149

Hollister Warehouse - 7989 Loma Prieta Earthquake

d

Roof, East Wall

~: ~ f ' ~a// !

0.0 7.0 2.0 3.0 4.0 5.0 6.0 7.0 B.O 9.0 70.0 Frequency, Hz

Fig. 5.14 (cont.) Hollister warehouse - 1989 Lorna Prieta earthquake.

(b) Nonnalized Fourier amplitude spectra of longitudinal acceleration response.

150

.s .,. ..........,

c:: Q)

..... E VI ..... Q) (J

tJ --~ '-C)

Hollister Warehouse - 1984 Morgan Hill Earthquake

1,,0

0,,0

-1,,0

1,,0

0,,0

-1,,0

0.0

----T- -- - ---I

5.0

East - West !?esponse , ,

North - South Response

10.0 Time, sec

15.0 20.0

Ground ----------- Roof

25.0 30.0

Fig. 5.15 Hollister warehouse - Absolute displacement response.

. . S

... --N

c:: Q)

~ E VI tv

Q)

0 tJ --~ '-CJ

Hollister Warehouse - 1986 Hollister Earthquake

7.0

0.0

-1.0

1.0

0.0

-1.0

0.0 I

r ,---- -,

East West Response

North - South Response Ground

----------- Roof

A /\ r\J\ !'I. f\ ".M I\~J\ (;V\ ,~. ~ I vr ~ -IF ~-v-VJ ---v:-~-~~

5.0 10.0 Time, sec

15.0 20.0 25.0

Fig. 5.15 (cont.) Hollister warehouse - Absolute displacement response.

30.0

.-. VI w

10.0 .

oS 0.0 ...

-h.J

~ -10.0 E (l)

10.0 (J tJ --~ 0- 0.0

C)

-10.0

Hollister Warehouse - 1989 Lama Prieta Earthquake r----------1

East - West Response

North - South Response

Time, sec

Ground ----------- Roof

0.0 5.0 10.0 15.0 20.0 25.0 30.0

Fig. 5.15 (cont.) Hollister warehouse - Absolute displacement response.

oS ~

c: Q)

E Q) 0 tJ -~ 0-

CJ

§2 0-~

tJ -Q)

a::

Hollister Warehouse - 1984 Morgan Hill Earthquake

0.8 Channel 4 - Roof, Center (Unfiltered)

0.0

-0.8 0.8 Channel 4 - Roof, Center (Filtered)

0.0

-0.8 0.8 Channel 5 - Roof, West Wall (Unfiltered)

0.0

-0.8 0.8 Channel 5 - Roof, West Wall (Filtered)

0.0

-0.8 0.8 Channel 6 - Midpanel, West Wall (Un filtered)

0.0

-0.8 0.8 Channel 6 - Midpan el, West Wall (Filtered)

0.0

-0.8 Time, sec

0.0 10.0 20.0 30.0 40.0 I ~--------~I __________ �~ ________ ~' __________ ~'

Fig. 5.16 Hollister warehouse -1984 Morgan Hill earthquake.

(a) Transverse relative displacement response.

154

·s ~

c: Q)

E Q) l)

CJ --&5-0-

(::)

Q) ~ '--f-..I tJ --Q)

Q:

Hollister Warehouse - 1984 Morgan Hill Earthquake

0.2 Channel 2 - Roo~ South Wall (Unfiltered)

0.0

-0.2 0.2 Channel 2 - Roo~ South Wall (Filtered)

0.0

-O.2~

0.2 Channel 3 - Roo~ North Wall (Unfiltered)

0.0

-0.2 0.2 Channel 3 - Roof, North Wall (Filtered)

0.0

-0.2 Time, sec

0.0 10.0 20.0 30.0 40.0 ! ~--------~'----------~'----------~'~--------~'

Fig. 5.16 (cont.) Hollister warehouse - 1984 Morgan Hill earthquake.

(b) Transverse relative displacement response.

155

.S

~

c::: Q)

E Q) () CJ -~ '-~ Q) :::> '-~ CJ -Q)

Q::

Hollister Warehouse - 1984 Morgan Hill Earthquake

0.2 Channel 10 - Roof, West Wall (Unfiltered)

0.0

-0.2 0.2 Channel 10 - Roof, West Wall (Filtered)

0.0

-0.2 0.2 Channel 11 - Roof, Center (Unfiltered)

0.0

-0.2 0.2 Channel 11 - Roof, Center (Filtered)

0.0

-0.2 0.2 Channel 12 - Roof, East Wall (Unfiltered)

0.0

-0.2 0.2 Channel 12 - Roof, East Wall (Filtered)

0.0

-0.2 Time, sec 0.0 10.0 20.0 30.0 40.0 I ~--------~I----------~I----------~I----------~I

Fig. 5.16 (cant.) Hollister warehouse -1984 Morgan Hill earthquake.

(b) Transverse relative displacement response.

156

.S -i-J'"

c:: Q)

E Q) \.) Q -~ '-CJ

~ '--i-J

Q -Q)

Q::

Hollister Warehouse - 1986 Hollister Earthquake

0.8 Channel 4 - Roof, Center (Unfiltered)

0.0

-O.B O.B Channel 4 - Roof, Center (Filtered)

0.0

-0.8 O.B Channel 5 - Roof, West Wall (Unfiltered)

0.0

-0.8 0.8 Channel 5 - Roof, West Wall (Filtered)

0.0

-O.B 0.8 Channel 6 - Midpanel, West Wall (Unfiltered)

0.0

-0.8 0.8 Channel 6 - Midpan el, West Wall (Filtered)

0.0

-0.8 Time, sec

0.0 10.0 20.0 30.0 40.0 I ~--------~I----------~'----------~I----------~I

Fig. 5.17 Hollister warehouse - 1986 Hollister earthquake.

(a) Transverse relative displacement response.

157

.S

~ c:: Q)

E Q) \.) Q -~ '-a Cb ~ '--i-J

Q -Cb Q::

Hollister Warehouse - 1986 Hollister Earthquake

0.2 Channel 2 - Roof, South Wall (Unfiltered)

0.0

-0.2 0.2 Channel 2 - Roof, South Wall (Filtered)

0.0

-0.2 0.2 Channel 3 - Roof, North Wall (Unfiltered)

0.0

-0.2 0.2 Channel 3 - Roof, North Wall (Filtered)

0.0

-0.2 Time, sec

0.0 10.0 20.0 30.0 40.0 I ~--------~!----------~!----------~I ________ ~I

Fig. 5 .17 (cant.) Hollister warehouse - 1986 Hollister earthquake.

(a) (cant.) Transverse relative displacement response.

158

·s ~ ...

c::: Q)

E Q) \) C) --~ '-CJ Q) ::> '-~ C) --Q)

Q::

Hollister Warehouse - 1986 Hollister Earthquake

0.2 Channel 10 - Roof, West Wall (Unfiltered)

0.0

-0.2 0.2 Channel 10 - Roof, West Wall (Filtered)

0.0

-0.2 0.2 Channel 11 - Roof, Cen ter (Un filtered)

0.0

-0.2 0.2 Channel 11 - Roof, Center (Filtered)

0.0

-0.2 0.2 Channel 12 - Roof, East Wall (Unfiltered)

0.0

-0.2 0.2 Channel 12 - Roof, East Wall (Filtered)

0.0

-0.2 Time, sec 0.0 10.0 20.0 30.0 40.0 , ~--------~'----------~'----------~'--------~'

Fig. 5.17 (cont.) Hollister warehouse - 1986 Hollister earthquake.

(b) Longitudinal- relative displacement response.

159

·s ,..: c: Q)

E Q) (.)

0 -~ '-a Q) ~ '-...........

0 -Q)

ct:

Hollister Warehouse 1989 Loma Prieto Earthquake

4.0 Channel 4 Roof, Cen ter (Un filtered)

-4.0 4.0

Channel 4 - Roof, Center (Filtered)

0.0

-4.0 4.0

Cannel 5 - Roof, West Wall (Unfiltered)

0.0

-4.0 4.0 Channel 5 - Roof, West Wall (Filtered)

0.0

-4.0 4.0 Channel 6 - Midpanel, West Wall (Unfiltered)

0.0

-4.0 4.0 Channel 6 - Midpanel; West Wall (Filtered)

0.0

-4.0 Time, sec

0.0 70.0 20.0 30.0 40.0 50.0 60.0 I ~----~I------~I ______ ~I ______ ~I ______ ~I~ ____ ~I

Fig. 5.18 Hollister warehouse - 1989 Loma Prieta earthquake.

(a) Transverse relative displacement response.

160

·s ~

c: CD

E CD \.) Q --~ '-~ CD > '-~ Q --CD

Q::

Hollister Warehouse 1989 Lama Prieta Earthquake

0.5 Roof, South Wall (Unfiltered)

0.0

-0.5 0.5 Channel 2 - Roof, South Wall (Filtered)

0.0

-0.5 0.5 Channel 3 - Roof, North Wall (Unfiltered)

0.0

-0.5 0.5 Channel 3 - Roof, North Wall (Filtered)

0.0

-0.5 Time, sec

0.0 10.0 20.0 30.0 40.0 50.0 60.0 ~I _________ ~I ______ ~I ______ ~I ______ ~I ______ ~I ______ ~I

Fig. 5.18 (cont.) Hollister warehouse -1989 Loma Prieta earthquake.

(a) (cont.) Transverse relative displacement response.

161

"S -i-I"'

c: CD

E CD \.) CJ --~ "-CJ Q) > "--i-I

CJ --CD 0::

Hollister Warehouse - 1989 Loma Prieto Earthquake

0.5 Channel 10 - Roof, West Wall (Unfiltered)

0.0

-0.5 0.5 Channel 10 - Roof, West Wall (Filtered)

0.0

-0.5 0.5 Channel 77 - Roof, Center (Unfiltered)

0.0

-0.5 0.5 Channel 77 - Roof, Cen ter (Filtered)

0.0

-0.5 0.5 Channel 12 - Roof, East Wall (Unfiltered)

0.0

-0.5 0.5 Channel 12 - Roof, East Wall (Filtered)

0.0

-0.5 Time; sec 0.0 70.0 20.0 30.0 40.0 50.0 60.0

I L-____ ~' ______ -L' ______ ~I ______ ~I ______ ~I ______ ~I

Fig. 5.18 (cont.) Hollister warehouse - 1989 Lorna Prieta earthquake.

(b) Longitudinal relative displacement response.

162

163

I--'

0\ ~

§N ;70

Concr~t~ Wall Panel.

Beams

V Non-bearing Stud Partition

(F1r.,wall) I Wall

12' x 14' O.H. Door \jILl --12' x 14' O.H. Door

I ' 6 Bays 0 22'-0" I' 5 Bays (rJ 20'-0" · I

I • 232'-0" · I

Fig. 5.20 Floor plan - Redlands warehouse.

• (()

J ~ @

~ III "t

It a ..I a 0)

.... 0\ Ul

r It C) -.. I ~ C\I

L West Wall

I: C)

"I I't)

W-.

~ 1/2" Plywood Sheathing 2" x 4" Rafters @ 24" O/C 4" x 14/1 Purlins @ 8'-0" O/C

- Glulam-Bea;;; - - - T Pilaster

'" .~

5" Concrete Slab

~ It C)

.. I

~

I~ 00' a" ~1 - ~ I

Fig. 5.21 Transverse cross section - Redlands warehouse.

,..... 0\ 0\

North Elevation Sou th Elevation

East Elevation

West Elevation

Fig. 5.22 Elevations - Redlands warehouse.

r Glulam Beams r Tilt-Up Wall Panels

ir;o:: 1 vFir~wall •

:~4 7 t ~

I I 8 . I J

L------~I-· __

1-1------------232'--------~·1 Roof Plan

North Wall

1

if:71 1 I 12

I Y Firewall C) ~

I J I I I

232' I Slab Plan

South Wall

Elevation - West Woll

Sensors 1, 11, and 12 are moun ted on the floor slab. Sensor 4 is moun ted on the glulam beam. All other sensors are moun ted on the wall panels.

Fig. 5.23 Locations of strong-motion instruments - Redlands warehouse.

167

C)) 0.05

c:: .~ o. 00 ~IMM~ -;.....,

t) ~

~ -0.05 ~ 0.05 () "(

()) O.OO~~~ C/) \)

Redlands Warehouse

1986 Palm Springs Earthquake - NS

1986 Palm Springs Earthquake - EW

QJ -0.05 Time, sec

0.0 70.0 20.0 30.0 40.0 50.0 60.0 I~----------~I---------------~I---------------~I---------------~I _______________ ~I _______________ ~I

Fig. 5.24 Measured ground accelerations - Redlands warehouse.

168

I-' 0\ \0

Redlands Warehouse 1986 Palm Springs Earthquake - NS

.2.0,~ --~--~--~--~--~--~ ·s

0.25 ""-1 ---r--~-_._---.---.,--__, tl)

~ .. 0.20 '--hi

e ~ 0.15

Q.) u u ~ 0.10

2% ------- 5% _ .. - .. - 10% ----- 20%

'i-!"

c: Q.) 1.5

~ u tJ Q 1.0 .~ Cl

- f'\"W V~I tJ V\.' •• \ _ I

~ 005 ... ; \\1\ e 0.5 " .... / ........ 1 .. .....- .. -..... .. --..., U. \v" '. .... -hi I .. ------ "~ .. '-Q.) '\...,"-. \- u I .. -/ ......, Q.. ,'-~,.... Q.) /-,..~ .. ~ .. ~------.-----(" ' .~ .. ~~~ ,.-- ('\' ........ -:::::::..--' v J O. 00 -=:::::----~~-.. _~~7. - - - C) O. 0 ...-~/_'" T'C ..... :. ___ -_-'--_--'-__ -'--__ "--_---I

0.0 1.0 2.0 3.0 0.0 1.0 2.0 3.0 Period, sec Period, sec

1986 Palm Springs Earthquake EW 0.25 " --y---r--..----,----,-----.

tl) . 2.0 r-I --.,.--.,-----r------,.--..,...----,

·s g 0.20

:;:; e ~ 0.15 Q.) u u ~ 0.10

2% ------- 5% - .. - .. - 10% ----- 20%

-hi"

c: Q.) 1.5

~ u tJ Q 1.0 .~ Cl -e 0.5 -hi

u Q.) Q.. Vi O. 0 1 ..eo- 1 1 I

1.0 2.0 3.0 0.0 1.0 2.0 3.0 Period, sec Period, sec

Fig. 5.25 Linear response spectra - Redlands warehouse.

Redlands Warehouse 1986 Palm Springs Earthquake

0.2 Channel 2 Midpan el, West Wall

0.0

-0.2 0.2 Channel 3 - Roof, West Wall

0.0

-0.2 0.2 Channel 4 - Roo" Center

t))

0.0 c:: 0 :g -0.2 ~ 0.2 Channel 5 - Roof, 1/4 Point West Wall

-Q) ()

0.0 ()

"C

-0.2 0.2 Channel 6 - Roof, South Wall

0.0

-0.2 0.2 Channel 12 - Ground, West Wall

0.0

-0.2 Time, sec

0.0 5.0 10.0 15.0 20.0 25.0 30.0 I ~----~'~----~'------~'------~I------~' ____ ~!

Fig. 5.26 Redlands warehouse -1986 Palm Springs earthquake.

(a) Transverse acceleration response.

170

Redlands Warehouse 1986 Palm Springs Earthquake

0.2 Channel 8 Roo~ East Wall

-0.2

0.2 f Channel 9 - Roo~ South Wall

O 0 .. , II'" W\l···~1,f oM M IW~'~!~w.~tv l~"oj "'#"' .,1 ", ". , ., . 'If' .• ,.,.. 1 c:: .,fliH·rf ,4" 1 ,T'. p~, I r r~ pi" tv i'· rv r r , iii"· Ill' writ< 'ClQW4if .. Pi

.~ a -0.2 L

~ 0.2 -Q) \..) \..)

~

-0.2 0.2

-0.2

0.0 I

Channel 10 - Roo~ West Wall

Channel 11 - Ground, West Wall

Time, sec 5.0 10.0 15.0 20.0 25.0

I I I I I

Fig. 5.26 Redlands warehouse - 1986 Palm Springs earthquake.

(b) Longitudinal acceleration response.

171

30.0 I

~ -....) N

Redlands - i-story WarehouBe (CSMIP Station 23495)

III II Ii I'll II II Ii I' II II II I: 0 1 , II I' I 1'1' i " I' , I" I, i, II ' : I I: I' I ! I I II i I I, I! --- --- - - - -.. -- -,- ---- ----r ,'- - - - - - - - --- _.. -Record 23495-COl15-92182.02

I' 1 Ground Floor - Up

~

1~\J

II Center of

" Center of North Wall - E

Center of East Wall - N

:11 Ground Floor - N I

! -~ " E

Structure Reference Orientation: N=O° I

1_. • • • - • • • • • • • • •• ••••• ••••• • - • ~ r ·~~~~~-~--~--~----~----------~--------------~-~ I b I I I '11 I I : 2' I I ':3' I I A I I I 5 i I ! I I I ' I I I I I! 0, 0 iii 10' , ; I I ~ I ' '0, I 15' !! "i, " I I ; 20;: 'I '2Q Sec..

Fig,5.27 Redlands warehouse -1992 Landers earthquake [66).

~

-.l w

Redlands - I-story Warehouse (CSMIP Station 2349S) Record 2349S-C011S-92182.02

I '1,1' II (' I I' II II I; I I; I' I' I· I I I I I I'

,.. ------ - --- --- ---- -------. - -' .- . (' Max. Accel. ~ 0.04 g

l~~~~~--~~VV~~~~~~~----~----~----------------~----------------------------I

2V\r~~;yJ\f'~A,~~'--~ 3, 0.41 ~, .

4 ., 0.31. g .

5POC~~~ 6~ 0.13 g

j" 7~ _~ 0.12g

0.12 g

9\!!wJ~IfJ1~~~u.rVWtV1f1IMIvJ\t.Wf(V\MHv'~ 10 0.12 g

, 0.12 g

0.10 g

. . . . . . . . . .. . . . .. - . - . '~--~----~--------------~--------~------------~~ 22 I I I . I ! I ' I . '2S' . I! I ,I I i I I' I, I' ':30 !,! " ,I ' ". 35 ' ' :' " I' i . 40' , ' ., , , . 44 Sec'.

Fig. 5.27 (cant.) Redlands warehouse - 1992 Landers earthquake [66].

f-' -..) +:00

Redlands - 1-story Warehouse (CSMIP Station 23495) Record 23495-CollS-92182.05

-- --- -- -- --- - --- --- - - - - .. - -- - - - - - - -- - - - - - - ---

1 Ground Fl 00 r - II ~,~~~~~~~~-v~~~~~~AN~~~~~~ __ ~~~~~~ Max. Accel. 0.06

2 Mid-Ile t s,ht ... ~'--\.tJ"..t\ ("~:)I'y1'v~\ I\~r-v-A v"\:,F .. ,':u", ,r\"~,~t..I'~..Aw~~ , .''1\ ' .. 1\ ;' II/V ~ .

I I " ," ,\ . " \ . \ \. I '~. ~

3 Roof: Cent!.~ r-.J IV-. -', '(VV' f,tJ, 1\ Mtv...."-. I, ;-! : I /",,1\ d'... M. t-.A M f\./~~ -.I • V .,/, .... " .I" ,.~ II Y - -V \... 'I! k' \ ... V -...!. 1;1 V v v V

, ~". ,y' .\., \ 1

4 II CCI~~ rv... r--JY :',If\N~J\./·~\ri~"". :r", r\,/\ AI"\. f'\. t-v-./\rl'l f\ ~ v . V ' \.,; \ J V' \ \ 'I" I \ V '-.I' VI V V "

, I I I • '(

5 "-~~·::::;J:);;\)\'r"l\j\~~/\f'W'A/((\\N/\t fVA/\~V~~ 6 II __ f~n!:.~r of South Wall - O. 6

7 " 0.18 g

8 Roof: Center of East Wall - N

9 II Center of South Wall - IN, 'I I ' , " , .' I ~ ~ \, \ I ! ' .l .' I' , , ' 0.44 g t \ I \ } , " ~I' • , , ., "~ J". II, ' f r '\ 1\ A I" I' , ." ,

--""'N'VfI:}..I\.·~\.v.f":1v.M\'1I "'1I.~""I,IN,"',..j' I "',/" ,~I ;111/' :Ih' .; j j .I? "\1/' ; 'A,! ,,~ . ,,,,-,/,,,, ;' I ~'I"'~~l_' '1"11,, /1 /' ,\ "'/ .\.,... .. 1.,/ /''''/ 'f.d', Iv\ ,.j\j ... ~ ... l"'''''WW, t''/ ~ j , .. '{II ~ I 11A', I i II t I I 1'1 J" .. '11 ...... IIJ .', I ," "I I • ~ ,"I '" I \' i' ol t' ;'"

, I 'V I,', Y t ~ \ I 1.' of, _ ,-' I " I', , , '

,- --r--- ---r - 0.13 g 10 II Center of West Wall - ("./vv,.rV',,~r-\l-fwo~~

11 Ground Floor - N 0.13 g

12 " E 0.17 g

Structure Reference Orientation: N=O° .. . . . . - . . . . . . .. . ......... ... - . -----------------------------------------------o 2 3 4 5 10 15 20 22 Sec.

Fig.5.28 Redlands warehouse -1992 Big Bear earthquake [37].

Redlonds Warehouse - 1986 Palm Springs Earthquake

~: ~ _[ .-.:...~--<--",---=----,,,--C_h_a;.....;an ...... n--,,e_/ ...;..;.2~~M ....... id ....... p_o_n,,--e_I'---,-~,,--e_s_t .;.....,Wt ........ 0ll.-I I----"~~ 0.0-

1.0 Channel 12 Ground, West Wall

0.5. o.o~--~--~--~--~~~~--~~~~~~--~~~

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0 Frequency, Hz

Fig. 5.29 Redlands warehouse -1986 Palm Springs earthquake.

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

175

Q) \) ::1

Redlands Warehouse - 1986 Palm Springs Earthquake

~ 1.0 Channel 9 - Roof, South Wal'

! ~:~f ~ . frO~0 !.......

i ~: ~ fl L.._..L....-_--L-c_h_a_n...ll..n_e---" :......!1....L.

o

__

R

L...!..0_O_f.-.:..

' ..L-~~es....!...t--l....-,;..~--Za:.!.../,;,..../ l!---~....L.:.......!~~ Q 0.0-

E o <: ~:~f 0.0'

O. 0 1. 0 2. 0 3. 0 4.0 5. 0 6.0 7. 0 8. 0 9.0 10. 0 Frequency, Hz

Fig. 5.29 Redlands warehouse -1986 Palm Springs earthquake.

(b) Normalized Fourier amplitude spectra of longitudinal acceleration response.

176

~

-.l -.l

Redlands Warehouse - 1986 Palm Springs Earthquake ~-~

0.5 East West Response

• \ <r l:, • f \+l" ... r>. ;/">1.. i i " ••. I ~ t ,r tI~ jC .:2 c:' 0 0 ~"* \:: CJ It\:'i.-r!iIUd .... AI· 't tn ~. i • , .. c "V" r.. c ,,_ • x '- , ." ~

c: Q)

E Q) <J tJ --g-'-e)

-0.5

0.5 North South Response

O 0 ~" £... I .. ('> n. I .. ' OtJYit{\" n;. , J ) ~ 111.. "n () <, ;' 'C ) ~) (Y 4: • bC:Ji < • '" 'Iodo ~ ,""I.. u: riO 'II Y ., 4: y~ U) -.-... , _ .

-0,5 Time, sec

Ground Roof

0.0 5.0 10.0 15.0 20.0 25.0 30.0

Fig. 5.30 Redlands warehouse - Absolute displacement response.

·s ~

c: Q)

E Q) \.) \) --g-.-

Cj

<b :::> .-~

\) --<b ct

Redlands Warehouse - 1986 Palm Springs Earthquake

0.3 Channel 3 - Roof, West Wall (Unfiltered)

0.0

-0.3 0.3 Channel 3 - Roof, West Wall (Filtered)

0.0

-0.3 0.3 Channel 4 - Roof, Center (Unfiltered)

0.0

-0.3 0.3 Channel 4 - Roof, Center (Filtered)

0.0

-0.3 0.3 Channel 5 - Roof, 1/4 Point West Wall (Unfiltered)

0.0

-0.3 0.3

0.0

-0.3 Time, sec

0.0 10.0 20.0 30.0 40.0 I ~--------~I----------~I----------~'----------~I Fig. 5.31 Redlands warehouse -1986 Palm Springs earthquake.

(a) Transverse relative displacement response.

178

·S -1-->"

c:: Q)

E Q)

u tJ -~ '-CJ Q) :::::.-'--i-J

tJ -Q)

Q:

Redlands Warehouse - 1986 Palm Springs Earthquake

0.1 Channel 2 - MidpanelJ West Wall (Unfiltered)

0.0

-0.1 0.1

Channel 2 - Midpan el, West Wall (Filtered)

0.0

-0.1 0.1 Channel 6 - Roof, South Wall (Unfiltered)

0.0

-0.1 0.1 Channel 6 - Roof, South Wall (Filtered)

0.0

-0.1 0.1 Channel 7 - Roof, North Wall (Un filtered)

0.0

-0.7 0.7 Channel 7 - Roof, North Wall (Filtered)

0.0

-0.7 Time, sec 0.0 10.0 20.0 30.0 40.0 I ~-----------~I-------------~I-------------~I----------------~I

Fig. 5.31 (cant.) Redlands warehouse - 1986 Palm Springs earthquake.

(a) (cont.) Transverse relative displacement response.

179

Redlands Warehouse - 1986 Palm Springs Earthquake

0.06 Channel 8 - Roof, East Wall (Unfiltered)

0.00~~~~~~~~~~~4++++t~~~~~~~~

-0.06 0.06

~-0.06 '- 0.06 c:

Channel 8 - Roof, East Wall (Filtered)

Channel 9 - Roof, South Wall (Unfiltered)

E O. 00 ~'-h+~=I!::r:-n'Ifiii. Cl)

g-0.06 -~ 0.06 '-

Channel 9 - Roof, South Wall (Filtered)

C:l Cl) o. 0 0 ~NtlNWtNlt1N~iIH1 ~ 0--.;.....,

..2-0.06 ~ 0.06

-0.06 0.06

Channel 10 - Roof, West Wall (Unfiltered)

Channel 70 - Roof, West Wall (Filtered)

O.OO~--------~~--~~~--~--------------~

- O. 06 Tim e, sec

0.0 10.0 20.0 30.0 40.0 L' __________ ~, __________ ~, __________ ~, __________ ~,

Fig. 5.31 (cant.) Redlands warehouse - 1986 Palm Springs earthquake.

(b) umgitudinal relative displacement response.

180

181

~

00 N

® Ref. N

(N 25 W)

1---------------7 Bays 0 24'-0· - 168'-0·---------------1

6· Concr~t~ Wall Panms

CovtJr~ LoadIng Platform >-• C)

.1 ~

br- - - (1 - - J · '>-' '=;11-----..---+-12' x 12' O.H. Door 72' x 12' O.H. Door 12' x 12' Knock

Out Pan~1

[Op'" W.b SI • ., Glrd,rs

................. D················· 0········ ........ ·D················· .Q •••.••.•••.••••••• 0- •.••.•..•••••••.• D·················

·················0················· 0················· ·D················· ·D·················· 0- •.........••..••• D·················

. . . . . . . . [8 ".8 ". . 5~'. SlruCIU~/. TU: . . ~~umn.. . . .• . . . .0 . . . . . . • • . 0 . . . . . . . .

Concrete Wall Panels (t>plcal)

8-ln. thlck ~ 16-ln. thlck

A

FIrst-Story Plan - Milpitas Industrial Building

Fig. 5.33 Floor plans - Milpitas industrial building.

~ ,I

~ , C)

.1

~ 0

-; a It. II)

• C)

.1 Cl

~

· Cl ,I

~ 0

~ lX:I 'to

· Cl ,I ~

...... 00 w

ell) Ref. N

(N 25 W)

.---------------- 7 Bays tI 24'-0· - 168'-0·----------------1

~ -.r-----'L ...-----... ~ ,.---.... ,.---~

I? (I a (I (I

rmUIQm Boomo

II

I? (I 6 (I (I

II fB x 8 x 5/16 Structural Tub. Columns

. 6 c) (I (I

Concrtfttf Wall Pantfls (t;plcat)

8-ln. thIck 16-ln. thlck

l ~ ~ ---..!

Second-Story Plan - Milpitas Industrial Building

Fig. 5.33 (cont.) Floor plans - Milpitas industrial building.

• <:)

.1

~ . <:)

.1

~ 0

-n ~ 14)

· c .1 c ~

• c .1

~

• ~ ~

• c .1 ~

North Elevation

1.. __ -----

South Elevation

East Elevation

West Elevation

Fig. 5.34 Elevations - Milpitas industrial building.

184

1-------180'-on------l1

................. ····I·:...ll~r---·· ._""""" """---,.. ... J Roof Plan

@ ~ 6 81 1 ::: a

L12 .. I C) C\j

Ref. N .....

(N 25 W) J t 7

Second Floor Plan

I' 168'-0 11

1

9 10 1 1

2 L 13

C)

..I C)

1 C\j .....

J Slab Plan

Fig. 5.35 Locations of strong-motion instruments - Milpitas industrial building.

185

Milpitas Industrial Building

0.15 1988 Alum Rock Earthquake - NS

0.00

-0.15 \)) 0.15 1988 Alum Rock Earthquake - EW

c::: 0 0.00 '-"'i-.J

CJ ~ Q) -0.15 -Q)

0.15 () 1989 Loma Prieta Earthquake - NS

()

~

Q) 0.00 CI) CJ

en -0.15 0.15 1989 Loma Prieta Earthquake - EW

0.00

-0.15 Time, sec

0.0 10.0 20.0 30.0 40.0 50.0 60.0 I ~-------~!---------~I------~I ______ ~I ______ ~I ______ ~I

Fig. 5.36 Measured ground accelerations - Milpitas industrial building.

186

....... 00 --l

Milpitas Industrial Building 1988 Alum Rock Earthquake NS

tJ) 0.5 I' ---,----,------,-----y--.---~

~ .. 0. 4 ~

~ ~ 0.3 Q} l) l)

~ 0.2 -o -b l) 0.1 Q}

~ 0.0 1 -~"";" ~

0.0 1.0 2.0 Period, sec

2% 5% 10% 20%

3.0

1. 00 r-I ---,---r-----,----r----,----,

.~

...........

~ 0.75 ~ (j o 'Q O. 50 .~ C)

~ 0.25 ......... (j Q)

Q Vi 0.00 ....... c"-----'-__ "'---_~ _ _L_. _ __' _ ___'

0.0 1.0 2.0 3.0 Period, sec

1988 Alum Rock Earthquake EW

tJ) 0. 5 I' ---,-----,-----r-----,--..--~

§ .. 0.4

'-........

~ ~ 0.3 Q) l) (J

"'C 0.2 o

I......

-0 0.1 Q)

2% 5% 10% 20%

Q Vi I -~~~ I 0.0r=???

0.0 1.0 2.0 3.0 Period, sec

. 1.00..-. -~-~--~-~-~-~ ·S ...........

~ 0.75 ~ (J o 'Q0.50 .22 C)

e 0.25 ......... (j Q) Q.. Vi O. 00 I < I I I

0.0 1.0 2.0 3.0 Period, sec

Fig. 5.37 Linear response spectra - Milpitas industrial building.

~

00 00

Milpitas Industrial Building 1989 Loma Prieta Earthquake - NS

1. 00 ,,-----,----,----,---r-----,----, t))

C: ..

~ 0.75 e Q)

Q3 ~ 0.50

'<;( -tl .}:; 0. 25 u Q)

bt

2% ------- 5% - .. - .. - 10% ----- 20%

0.00' , , I

0.0 1.0 2.0 3.0 Period, sec

. 15.0 ·s -No, 12.5 c: Q)

~ 10.0 u tl Q 7.5 .~

(:) 5.0

e t 2.5 Q)

Q <.rj O. 0, -e-==r, , I

0.0 1.0 2.0 3.0 Period, sec

1989 Loma Prieta Earthquake EW 1. 00 ,.-----.--------r-----.---,-----,-----,

t))

c: .. ~ 0.75

~ Q) l> 0. 50 l>

'<;(

tJ .}:; 0.25 u Q)

~

2% ------- 5% - .. - .. - 10% ----- 20%

0.00' , , !

0. 0 1. 0 2. 0 3. 0 Period, sec

. 15.0 ·S -No, 12.5 c: Q)

t: 10.0 V I Q) I

U ~I tl /

Ci. 7.5 ,"" " . .~ , ..... - ',,---'../ Cl 5 0 , __ " __ ",,- .. /'/' - . '/" -,--' tl /,' .--' \..... - ,. , '" '" J." ,--t 2.5 ,./~' __ "~_/ Q) )' ,,~,--Q .~,.-'

<.rj o. 9J. 0 1. 0 2. 0 3. 0

Period, sec

Fig. 5.37 (cont.) Linear response spectra - Milpitas industrial building.

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.3 Channel 3 - Roo~ East Wall

0.0

-0.3 0.3 Channel 4 - Roof, North Wall

0.0

-0.3 0.3 Channel 5 - Roof, West Wall

\))

0.0 c: 0 :g -0.3 ~ 0.3 Channel 6 - Second Floor, East Wall -Q) \..) 0.0 \..)

"'(

-0.3 0.3 Channel 7 - Second Floor, North Wall

0.0

-0.3 0.3 Channel 9 - Ground, East Wall

0.0

-0.3 Time, sec 0.0 15.0 5.0 10.0 20.0 I ~-----------------~I-------------~--------~----------~I

Fig. 5.38 Milpitas industrial building - 1988 Alum Rock earthquake.

(a) Transverse acceleration response.

189

th

c: a "--;......,

C) !..... <b -<b () ()

~

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.3

0.0

-0.3

0.3 r 0.0 t

-0.3

0.3

0.0

-0.3

0.0 !

Channel 11 - Roof, East Wall

Channel 72 Second Floor, East Wall

Channel 13 - Ground, East Wall

5.0 I

Time, sec 70.0

I 75.0

I

Fig. :;.38 (cant.) Milpitas industrial building - 1988 Alum Rock earthquake.

(b) Longitudinal acceleration response.

190

20.0 I

Milpitas Industrial Building - 1989 Loma Prieta Earthquake

0.3 Channel 3 - Roof, East Wall

0.0

-0.3 0.3 Channel 4 - Roof, North Wall

0.0

-0.3 0.3 Channel 5 - Roof, West Wall

\))

0.0 c:"' a :g -0.3 t3 0.3 Channel 6 - Second Floor, East Wall --Q.) (,J

0.0 (,J

~

-0.3 0.3 Channel 7 - Second Floor, North Wall

0.0

-0.3 0.3 Channel 9 - Ground, East Wall

0.0

-0.3 Time, sec 0.0 5.0 10.0 15.0 20.0 25.0 30.0

! ~ ____ ~J ______ ~' __________ ~! ______ ~I ______ ~! ______ ~,

Fig. 5.39 Milpitas industrial building - 1989 Lorna Prieta earthquake.

(a) Transverse acceleration response.

191

Q)

'" c:: a '--i-.I 0 ~ Q) -Q) () ()

"'(

Milpitas Industrial Building 1989 Lama Prieta Earthquake

Channel 11 East Wall 0.3

0.0

-0.3

0.3 Channel 12 - Second Floor, East Wall

0.0

-0.3

0.3 Channel 13 - Ground, East Wall

0.0

-0.3

Time, sec 0.0 5.0 10.0 15.0 20.0 25.0 30.0

~I ______ ~I ______ ~I ______ ~I ______ ~I ______ ~I ______ ~I

Fig. 5.39 (cont.) Miipitas industrial buiiding - 1989 Lorna Prieta earthquake.

(b) Longitudinal acceleration response.

192

Q) -c:; ::J

Milpitas Industrial Building - 1988 Alum Rock Earthquake

Channel 3 - Roof, East Wall

~:~f o.o~~~-----~-----~--~~--~~~~~~----~~~~

Channel 4 - Roof, North Wall

~~f , ~ Channel 5 - Roof, West Wall

! ~~f .~ o.o~~~--~--~--~~--~--~~~--~~~~~~

::J

~ Channel 6 - Second Floor, East Wall

~~f O.OL-~~--~--~--~----~--~~~~~~~~~

1.0 f 0.5

0.0

Channel 7 - Second Floor, North Wall

AI 1.0 Channel 9 - Ground, East Wall

~~f 0.0 7.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0

Frequency, Hz

Fig. 5.40 Milpitas industrial building - 1988 Alum Rock earthquake.

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

193

Milpitas Industrial Building - 1988 Alum Rock Earthquake

~ ~~f _ Channel 11 - Roof, East Wall

Q.

E ':( Channel 12 - Second Floor, East Woll

J ~~f -0

Q) 1'-.1

~ Channel 13 - Ground, East Wall

j ~~f O.O~~~--~--~--~~~----~~~--~~~~~

0.0 1.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0 9.0 10.0 Frequency, Hz

Fig.5.40 (cant.) Milpitas industrial building -1988 Alum Rock earthquake.

(b) Normalized Fourier amplitude spectra of longitudinal acceleration response.

194

Milpitas Industrial Building - 1989 Loma Prieta Earthquake

~: ~ [ :,t :0/: 0.0 7.0 2.0 3.0 4.0 5.0 6.0 7.0 8.0

Frequency, Hz

Fig. 5.41 Milpitas industrial building - 1989 Lorna Prieta earthquake.

o

rl '

9.0 10.0

(a) Normalized Fourier amplitude spectra of transverse acceleration response.

195

"\:J Q)

.~ -

Milpitas Industrial Building - 1989 Loma Prieta Earthquake

tJ 1.0 r Channel 13 - Ground, East Wall

~ O.5[ O.O~--~--~~~~~--~~~~~~~~~~~~

O. 0 1. 0 2. 0 3. 0 4. 0 5. 0 6. 0 7. 0 8. 0 9. 0 1 O. 0 Frequency, Hz

Fig. 5.41 (cant.) Milpitas industrial building -1989 Lorna Prieta earthquake.

(b) Normalized Fourier amplitude spectra of longitudinal acceleration response.

196

...... \0 --l

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.25 .

. S 0.00 ... .........,

~ -0.25 E Q)

0.25 (J tJ --~ "- 0.00

t:::)

-0.25

0.0

----,---------- - -- I

East - West Response

North - South Response

r,

5.0 Time, sec

10.0 15.0

Ground ----------- Roof

20.0

Fig. 5.42 Milpitas industrial building - Absolute displacement response.

I-' \D 00

Milpitas Industrial Building -- 1989 Loma Prieta Earthquake ,--------------, -- -r-----

10.0 East - West Response

. ·s 0.0 -i-J"'

~ -10.0 Ground

----------- Roof E Q) {J tJ

---g-'-CJ

10.0

0.0

-10.0

0.0 5.0

North - South Response

10.0

Time, sec

15.0 20.0 25.0 30.0 L-----________ -L ______________ ~ ______________ ~ ______________ ~ ____________ ~~ ____________ ~

Fig. 5.42 (cant.) Milpitas industrial building - Absolute displacement response.

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.08 Channel 3 - Roof, East Wall (Unfiltered)

O.OO~~~~~~~~~~~~~~~~~~------~

-0.08 0.08 Channel 3 - Roof, East Wall (Filtered)

O.OO~--~~~~~~--------------------~

c:: -0.08 0- O.OB c:

Channel 4 - Roof, North Wall (Unfiltered)

~ o. 00 ~~~~~~~~~~~~~~IIt-+-II~~"""----I Q)

g -0008 --~ 0.08 0-~

'--;...... -2 -O.OB ~ 0.08

-0.08 O.OB

Channel 4 - Roof, North Wall (Filtered)

Channel 5 - Roof, West Wall (Unfiltered)

Channel 5 - Roof, West Wall (Filtered)

O.OO~--~~~~~~~~----------------------~

-0.08

0.0 !

Time; sec

5.0 10.0 15.0 20.0 ! I ! !

Fig. 5.43 Milpitas industrial building - 1988 Alum Rock earthquake,

(a) Transverse relative displacement response.

199

25.0 I

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.05 Channel 6 - Second Floor, East Wall (Unfiltered)

O.OO~~~~~~~~~--~~~~~~~~~----~

-0.05 0.05 Channel 6 - Second Floor, East Wall (Filtered)

O.OO~--~~~~~~~~--------------------~

c:: -0.05 '- 0.05 c:

Channel 7 - Second Floor, North Wall (Unfiltered)

E O.OO~~-+~~~~~~~~W+~~~~~~~----~ Q)

g -0.05 -~ 0.05 "-~ ~ 0.00

"--;.....,

~ -0.05 ~ 0.05

-0.05 0.05

Channel 7 - Second Floor, North Wall (Filtered)

Channel 8 - Second Floor, West Wall (Unfiltered)

Channel 8 - Second Floor, West Wall (Filtered)

O.OO~--~~~~~~~~----------------~--~

-0.05

0.0 I

Time, sec 5.0 10.0 15.0 20.0

I , I I

Fig. 5.43 (cant.) Milpitas industrial building -1988 Alum Rock earthquake.

(a) (cant.) Transverse relative displacement response.

200

25.0 I

·s --i-.J""

c:: Q)

E Q)

0 ~ ......, -~ '-CJ Q) ~ '---i-.J

t) -Q)

Q:

Milpitas Industrial Building - 1988 Alum Rock Earthquake

0.2

0.0

-0.2 0.2

0.0

-n ? \J'.ttC.- -

0.2 ~ n n

-::: t 0.2

0.0

-0.2

0.0 !

Channel 11 - Roof, East Wall (Unfiltered)

Channel 11 - Roof, East Wall (Filtered)

Channel 12 - Second Floor, East Wall (Unfiltered)

,..." c . Ai I .1\. • A-"".

0= V u • vw' '" d 4:0'0

Channel 12 - Second Floor, East Wall (Filtered)

Time, sec 5.0 10.0 15.0 20.0

! , , I

Fig 5.43 (cont.) Milpitas industrial building -1988 Alum Rock earthquake.

(b) Longitudinal relative displacement response.

201

25.0 I

Milpitas Industrial Building - 1989 Loma Prieto Earthquake

0.4

-0.4 0.4

Channel 3 - Roo~ East Wall (Unfiltered)

Channel 3 - Roo~ East Wall (Filtered)

o.o~----~~~~------------------------------~

-0.4 ·s 0.4

c::

Channel 4 - Roo~ North Wall (Unfiltered)

EO. 0 I-oc--::.r'---I!I--tlit-' .... -+n--+'---'~~r-:+--+-+t-+~""'r--7C""f-~-t""'-+--f-lI~-T-+-~~-r>'1 Q)

g -0.4 -~ 0.4 Channel 4 - Roo~ North Wall (Filtered) .-a

.-~

-S? -0.4 Q)

ct: 0.4 Channel 5 - Roo~ West Wall (Unfiltered)

Channel 5 - Roof, West Wall (Filtered) -0.4

0.4 ~

O.Or~-----·~·~··~~-+~~----------------------------~

-0.4 [

0.0 ,

Time, sec 10.0 20.0 30.0 40.0 50.0 , I 1 I J

Fig.5.44 Milpitas industrial building - 1989 Lorna Prieta earthquake.

(a) Transverse relative displacement response.

202

60.0 ,

Milpitas Industrial Building - 1989 Loma Prieta Earthquake

0.4 Channel 6 - Second Floor, East Wall (Un filtered)

0.0

-0.4 0.4 Channel 6 - Second Floor, East Wall (Filtered)

0.0

·s -0.4 0.4 Channel 7 - Second Floor, North Wall (Unfiltered)

~

c: Q) 0.0 E Q) ()

-0.4 CJ -~ .- 0.4 Channel 7 - Second Floor, North Woll (Filtered) CJ Q) 0.0 ~ .-

-h..I

-2 -0.4 Channel 8 - Second Floor, West Woll (Unfiltered) Q)

Ct:: 0.4

0.0

-0.4 0.4 Channel 8 - Second Floor, West Wall (Filtered)

0.0

-004 Time j sec 0.0 10.0 20.0 30.0 40.0 50.0 60.0 ! ~----~!------~!------~!------~!------~!------~!

Fig.5.44 (cant.) Milpitas industrial building - 1989 Loma Prieta earthquake.

(a) (cant.) Transverse relative displacement response.

203

·s .....; c: Q)

E Q) () C) --~ .-~

Q) :> .-......... C) --Q)

Q::

Milpitas Industrial Building - 1989 Loma Prieta Earthquake

1.0 Channel 11 - Roof, East Wall (Unfiltered)

0.0

-1.0 1.0 Channel 11 - Roof, East Wall (Filtered)

0.0

-1.0 1.0 Channel 12 - Second Floor, East Wall (Unfiltered)

0.0

-1.0 1.0 Channel 12 - Second Floor, East Wall (Filtered)

0.0

-1.0

0.0 10.0 20.0 Time, sec

30.0 40.0 50.0 60.0 I ( ( , , (

Fig.5.44 (cont.) Milpitas industrial building - 1989 Lorna Prieta earthquake.

(b) Transverse relative displacement response.

204

I

OIR~CTION OF NORTH WALL COLLAPSE

r----- --1--------

--COLLAPSED 1 i J CQU.APSEO

ROOF I 1 ROOF-

t c;

---- '----.--L _____ ..J L ____ ..J

DIRECTION OF SOUTH WALL COLLAPSE

1004'

200' 20'- 201' r 201' 201'

; 11 ; :i;1 ::::: r- .. ... .. ..

"" .. I'" ..

"" ... ~ ..

"" ~ ~r'

: i ;, :;

\.. R.R. TRACKS

"" ~ ~ ~

"" ~ I- ~

,.. l- I- ~ ,.. .. ~ ... too

-d

1 ~L.-________ ~ ________ ~ ________ ~ ________ L-______ ~

® ® 0

~

SC:AL.E ,,- FUT

Fig 6.1 Plan view of Elmendorf Warehouse indicating locations of damage [32].

205

TYPICAL ROOF FRAM I NG - EACH SECTION eo 0 eo 100

6 AT 33'-4" : 200'-0" 25' 9" I- "\' --\

ei I· 1m"'" 1 .... 1: SECTION A-A

Fig. 6.2 Typical roof framing plan and elevation of fire wall in Elmendorf Warehouse [32].

206

II

Fig. 6.3 Failure of a wood ledger beam in cross-grain bending.

207

VECTOR ELECTRONICS

Roof-Wall P eI B win Out Roof Collapse

oof Separation

R Col lapse~

..... -::;-'" -;:::-

Sulan 1 on • 0 9 /:[

Purlln,~ -- - ----..... ,.., ,., r ",.., - _

b ~

~ ,- '"' ..... ~

(:) ~

(:) \(.\

I

I 100' I

....... ... " "1'- ... - , i\ ....

~ ~ ,

.4'~ ,

"

, ~

\ \ UpaneiS

Bowing Out

b

Ll I /

~ ~ Pan~. '-:wng Ou~ Roof Collapse t- Panel Collapse

140'

240'

SAWYER CABINET, INC.

G/ulam

Purl!n

6- Steel Column

Roof Collapse

Fig. 6.4 Summary of damage in tilt-up buildings during the 1971 San Fernando earthquake [41,55].

208

IV o \0

WAREHOUSE BUILDING

I~ 276' -0" 'I

panels bowing out

panel collapse

ALL PHASE COLOR

I

I s:: <.0 I

r-n roo

~

glulam fell (probably after the wall fell)

I.. 280'-0" _I

roof collapse

wall collapse

glulam roof collapse

wall collapse

wall stressing on exterior walls wall destroyed

Fig.6.4 (cant.) Summary of damage in tilt-up buildings during the 1971 San Fernando earthquake [41,55].

1/1" • , 16'· '7..-

1~ II

---------

(a)

v

.~.~

3 OSI DES TYP.

~----r--r---'

1/1,-

!r' I.

I!

NEW 7S 7'1.7,;.1/4

-.. p .?

?

N E trJ L 3 )( 3 ~/6 XOI_~II Lt;?N& t::. .~. V4! \ - ~/511 q ExP. BOLT (4\1~MB2v)

@ '2.. '- (.? II C?c", . TY P,

Fig.6.5 Use of steel kickers to increase the strength and stability of tilt-up panels with large openings [10].

210

Ground r Motion J

----/\

/

Out-of-Plane / Component

(~/ \. Chord Force

Fig. 6.6 Skewed wall joints [35].

PANEL

PL YWOOD SHEATHING

HANGER

Chord Force

/\..'OOD PURLIN

STEEL CHANNEL LEDGER

Fig.6.7 Typical connection between wood purlin and steel ledger [10].

211

APPENDIX A

SUMMARY OF USC PROVISIONS

212

1976 UNIFORM BUILDING CODe

Anchorage of Concrete or Masonry Walls Sec. 2310. Concrete or masonry walls shall be anchored to all floors

and roofs which provide lateral support for the wall. Such anchorage shall provide a positive direct connection capable of resisting the horizontal forces specified in this Chapter or a minimum force of 200 pounds per lineal foot of wall, whichever is greater. Walls shall be designed to resist bending between anchors where the anchor spacing exceeds 4 feet. Re­quired anchors in masonry walls of hollow units or cavity walls shall be embedded in a reinforced grouted structural element of the wall. See Sec­tion 2312 U) 2D and 2312 U) 3A.

D. Diaphragms: Floor and roof diaphragms shall be designed to resist the forces set forth in Table No. 23-J. Diaphragms supporting concrete or masonry walls shall have continuous ties between diaphragm chords to distribute, into the diaphragm, the anchorage forces specified in this Chapter. Added chords may be used to form sub-diaphragms to transmit the anchorage forces to the main cross ties. Diaphragm deformations shall be considered in the design of the supported walls. See Section 2312 U) 3 A for special anchorage requirements of wood diaphragms.

3. Special requirements. A. Wood diaphragms providing lateral sup­port for concrete or masonry walls. Where wood diaphr'agms are used to laterally support concrete or masonry walls the anchorage shall conform to Section 2310. In Zones No.2, No.3 and No.4 anchorage shall not be accomplished by use of toe nails, or nails subjected to withdrawal; nor shall wood framing be used in cross grain bending or cross grain tension.

B. Pile caps and caissons. Individual pile caps and caissons of every building or structure shall be interconnected by ties, each of which can carry by tension and compression a minimum horizontal force equal to 10 percent of the larger pile cap or caisson loading, unless it can be demonstrated that equivalent restraint can be provided by other approved methods.

C. Exterior elements. Precast, nonbearing, nonshear wall panels or similar elements which are attached to or enclose the exterior, shall ac­commodate movements of the structure resulting from lateral forces or temperature changes. The concrete panels or other elements shall be sup­ported by means of cast-in-place ~:Jncrete or by mechanical fasteners in accordance with the following provisions.

Connections and panel joints shall allow for a relative movement be­t~ een stories of not less than two times story drift caused by wind or (3,0 . AI times story drift caused by required seismic forces; or \4 inch, whlche'yer is greater.

Connections shall have sufficient ductility and rotation capacity so as to preciude fracture of the concrete or brittle failures at or near welds. Inserts In ~O:i~rete shall be attached to, or hooked around reinforcing steel, or other", Ise terminated so as to effectively transfer forces to the reinforcing steel

Connections to permit movement in the plane of the panel for story drift shall be properly designed sliding connections using slotted or oversize holes or may be connections which permit movement by bending of steel or other connections providing equivalent sliding and ductility capacity.

213

TABLE NO. 23-J-HORIZONTAL FORCE FACTOR "Cp" FOR ELEMENTS OF STRUCTURES

DIRECTION VALUEOF PART OR PORTION OF BUILDINGS OFFORCE Cp

1. Exterior bearing and non bearing walls, Normal to interior bearing walls and partitions, flat 0.20' interior nonbearing walls and partitions. surface Masonry or concrete fences

2. Cantilever parapet Normal to flat 1.00

surface

3. Exterior and interior ornamentations and Any appendages. direction 1.00

4. When connected to, part of, or housed within a building: a. Towers, tanks, towers and tanks plus

contents, chimneys, smokestacks and 0.202

penthouse

b. Storage racks with the upper storage Any level at more than 8 feet in height direction 0.202 ) plus contents

c. Equipment or machinery not required for life safety systems or for continued 0.202 •

operations of essential facilities

d. Equipment or machinery required for life safety systems or for continued 0.50' S

operation of essential facilities

5. When resting on the ground, tank plus Any effective mass of its contents. direction 0.12

6. Suspended ceiling framing systems (Ap- Any plies to Seismic Zones Nos. 2, 3 and 4 direction 0.2()6 only)

i. Floors and roofs acting as diaphragms Any 0.12' direction

8. Connections for exterior panels or for Any elements complying with Section 2312 direction 2.00 U) 3C.

9. Connections for prefabricated structural Any 0.30' elements other than walls, with force direction

applIed at center of gravity of assembly

'See also Section 2309 (b) for minimum load on deflection criteria for mterlor pa~l!lons.

:When located in the upper portion of any building where the hnlD ratio is fl\,c·to-one or greater the value shall be increased by 50 percent.

'u p for s:orage racks shall be the weight of the racks plus contents. The ':&iue 0: Cp for racks over two storage support levels in height shall be o 1 t> for :hc: levels below the tOP two levels. In lieu of the tabulated values ~:ee~ .:.:o~age racks may be designed in accordance with U.B.e. Standard

v. nc-rc: a number of storage rack units are interconnected so that there are a mt:'llmLUTl of four vertical elements in each direction on each column line deIgned 10 resist horizontal forces, the design coefficients may be as for a bUI:amg WIth K values from Table No. 23-1, CS = 0.20 for use in the for­ml>i.1 ~. = ZIKCSWand Wequal to the total dead load plus 50 percent of tht r~a rated capacity. Where the design and rack contigurations are in accordUlce with this paragraph the design provisions in U .B.e. Standard '0 ;~ ·11 do not apply.

-For ne11ble and flexibly mounted equipment and machinery, the appropriate .:l.iue. of Cp shall be determined with consideration given to both the c .-na:TlIc properties of the equipment and machinery and to the building or ~:r ... cturc: In which it is placed but shall not be less than the lrsted values. The de.lgn of the equipment and machinery and their anchorage is an in­tegral pan of the design and specification of such equipment and mactllnc-ry.

'For Essential Facilities and life safety systems, the design and detailing of equlpmen[ which must remain in place and be functional following a major earthquake shall consider drifts in accordance with Section 2312 (k). The product of IS need not exceed 1.5.

'Ceiling weight shall include all light fixtures and other equipment which are laterally supported by the ceiling. For purposes of determining the lateral force. a ceiling weight of not less than 4 pounds per square foot shall be used.

'Floors and roofs acting as diaphragms shall be designed for a minimum force resulting from a ell of 0.12 applied to w x unless a greater force results from the distribution of lateral forces in accordance with Section 2312 (e).

'The Wp shall include 25 percent of the floor live load in storage and warefiouse occupancies.

214

1991 UNIFORM BUILDING CODE

Anchorage of Concrete or Masonry Walls Sec. 2310. Concrete or masonry walls shall be anchored to all floors, roofs and

other structural elements which provide required lateral support for the wall. Such anchorage shall provide a positive direct connection capable of resisting the ?ori­zontal forces specified in this chapter or a minimum force of200 pounds per lmeal foot of wall, whichever is greater. Walls shall be designed to resist bending between anchors where the anchor spacing exceeds 4 feet. Required anchors in masonry walls of hollow units or cavity walls shall be embedded in a reinforced grouted structural element of the wall. See Sections 2336, 2337 (b) 8 and 9.

Lateral Force on Elements of Structures and Nonstructural Components Supported by Structures

Sec. 2336. (a) General. -

(b) Design for Total Lateral Force. The total design lateral seismic force, Fp ,

shall be determined from the following formula:

Fp = ZICpWp (36-1 )

The values of Z and I shall be the values used for the building. EXCEPTIONS: 1. For anchorage of machinery and equipment required for

life-safety systems, the value of I shall be taken as 1.5.

2. For the design of tanks and vessels containing sufficient quantities of highly toxic or explosive substances to be hazardous to the safety of the general public if released, the value of I shall be taken as 1.5.

3. The value of I for panel connectors for panels in Section 2337 (b) 4 C shall be 1.0 for the entire connector.

The coefficient Cp is for elements and components and for rigid and rigidly sup­ported equipment. Rigid or rigidl y supported equipment is defined as having a fun­damental period less than or equal to 0.06 second. Nonrigid or flexibly supported equipment is defined as a system having a fundamental period, including the equipment, greater than 0.06 second.

The lateral forces calculated for nonrigid or flexibly supported equipment sup­ported by a structure and located above grade shall be determined considering the dynamic properties of both the equipment and the structure which supports it, but the value shall not be less than that listed in Table No. 23-P. In the absence of an analysis or empirical data, the value of Cp for nonrigid or flexibly supported equip­ment located above grade on a structure shall be taken as twice the value listed in Table No. 23-P, but need not exceed 2.0.

EXCEPTION: Piping, ducting and conduit systems which are constructed of ductile materials and connections may use the values of Cp from Table No. 23-P.

The value of Cp for elements, components and equipment laterally self-sup­ported at or below ground level may be two thirds of the value set forth in Table No. 23-P. However, the design lateral forces for an element or component or piece of equipment shall not be less than would be obtained by treating the item as an inde­pendent structure and using the provisions of Section 2338.

The design lateral forces determined using Formula (36-1) shall be distributed in proportion to the mass distribution of the element or component.

Forces determined using Formula (36-1) shall be used to design members and connections which transfer these forces to the seismic-resisting systems.

For applicable forces in connectors for exterior panels and diaphragms, refer to Section 2337 (b) 4 and 9.

F~rces shall be applied in the horizontal directions, which result in the most cri tical ioadings for design.

215

Detailed Systems Design Requirements

Sec. 2337.

(b) Structural Framing Systems. 1. General. Four types of general building framing systems defined in Section 2333 (f) are recognized in these provisions and shown in Table No. 23-0. Each type is subdivided by the types of vertical elements used to resist lateral seismic forces. Special framing requirements are given in this section and in Chapters 24 through 27.

2. Detailing for combinations of systems. For components common to differ­ent structural systems, the more restrictive detailing requirements shall be used.

3. Connections. Connections which resist seismic forces shall be designed and detailed on the drawings.

4. Deformation compatibility. All framing elements not required by design to be part of the lateral force-resisting system shall be investigated and shown to be adequate for vertical load-carrying capacity when displaced 3 (R .... ./8) times the dis-

placements resulting from the required lateral forces. P ~ effects on such elements shall be accounted for. For designs using working stress methods, this capacity may be determined using an allowable stress increase of 1.7. The rigidity of adjoin­ing rigid and exterior elements shall be considered.as follows:

A. Adjoining rigid elements. Moment-resistant frames may be enclosed by or adjoined by more rigid elements which would tend to prevent the frame from re­sisting lateral forces where it can be shown that the action or failure of the more rigid elements will not impair the vertical and lateral load-resisting ability of the frame.

B. Exterior elements. Exterior nonbearing, nonshear wall panels or elements which are attached to or enclose the exterior shall be designed to resist the forces per Formula (36-1) and shall accommodate movements of the structure resulting from lateral forces or temperature changes. Such elements shall be supported by means of cast-in-place concrete or by mechanical connections and fasteners in ac-, cordance with the following provisions:

(i) Connections and panel joints shall allow for a relative movement between stories of not less than two times story drift caused by wind, 3(R ..... {8) times the calculated elastic story drift caused by design seismic forces, or 1 h inch, whichever is greater.

(ii) Connections to permit movement in the plane of the panel for'story drift shall be sliding connections using slotted or oversize holes, connections which permit movement by bending of steel, or other connections provid­ing equivalent sliding and ductility capacity.

(iii) Bodies of connections shall have sufficient ductility and rotation capacity so as to preclude fracture of the concrete or brittle failures at or near welds.

(iv) The body of the connection shall be designed for one and one-third times the force determined by Formula (36-1).

(v) All fasteners in the connecting system such as bolts, inserts, welds and dowels shall be designed for four times the forces determined by Formula (36-1).

(vi) Fasteners embedded in concrete shall be attached to, or hooked around, re­inforcing steel or otherwise terminated so as to effectively transfer forces to the reinforcing steel.

5. Ties and continuity. All parts of a structure shall be interconnected and the connections shall be capable of transmitting the seismic force induced by the parts being connected. As a minimum, any smaller portion of the building shall be tied to the remainder of the building with elements having at least a strength to resist ~ times the weight of the smaller portion.

3

A positive connection for resisting a horizontal force acting parallel to the mem­ber shall be provided for each beam, girder or truss. This force shall not be less than ~ times the dead plus live load.

5

6. Collector elements. Collector elements shall be provided which are capable of transferring, the seismic forces originating in other portions of the building to the element providing the resistance to those forces.

216

7. Concrete frames. Concrete frames required by design to be part of the lateral force-resisting system shall conform to the following:

A. In Seismic Zones Nos. 3 and 4 they shall be special moment-resisting frames.

B. In Seismic Zone No.2 they shall, as a minimum, be intermediate moment­resisting frames.

8. Anchorage of concrete or masonry walls. Concrete or masonry walls shall be anchored to all floors and roofs which provide lateral support for the wall. The anchorage shall provide a positive direct connection between the wall and floor or roof construction capable of resisting the horizontal forces specified in Section 2336 or Section 2310. Requirements for developing anchorage forces in dia­phragms are given in Section 2337 (b) 9 below. Diaphragm deformation shall be considered in the design of the supported walls.

9. Diaphragms.

A. The deflection in the plane of the diaphragm shall not exceed the pennissible deflection of the attached elements. Pennissible deflection shall be that deflection which will permit the attached element to maintain its structural integrity under the individual loading and continue to support the prescribed loads.

B. Floor and roof diaphragms shall be designed to resist the forces detennined in accordance with the following formula:

i=x (37-1 )

i=x

The force Fpx determined from Formula (37 -1) need not exceed 0.75 Z I wp.r, but shall not be less than 0.35 Z I wpx .

\\Then the diaphragm is required to transfer lateral forces from the vertical resist­ing elements above the diaphragm to other vertical resisting elements below the diaphragm due to offset in the placement of the elements or to changes in stiffness in the vertical elements, these forces shall be added to those determined from For­mula (37-1).

C. Diaphragms supporting concrete or masonry walls shall have continuous ties or struts between diaphragm chords to distribute the anchorage forces specified in SectIon 2337 (b) 8. Added chords may be used to form subdiaphragms to transmit the anchorage forces to the main crossties.

D, \1.rhere wood diaphragms are used to laterally support concrete or masonry walls. the anchorage shall conform to Section 2337 (b) 8 above. In Seismic Zones r-;os :.3 and 4 anchorage shall not be accomplished by use of toenails ornails sub­jeC1 to WIthdrawal, nor shall wood ledgers or framing be used in cross-grain bend­mg or cross-grain tension, and the continuous ties required by Item C above shall be :n addItIon to the diaphragm sheathing.

E. Connections of diaphragms to the vertical elements and to collectors and con­nectlons of collectors to the vertical elements in structures in Seismic Zones Nos. 3 and.!. having a plan irregularity ,of Type A, B, C or D in Table No. 23-N, shall be deSIgned without considering one-third increase usually permitted in allowable stresses for elements resisting earthquake forces.

E In structures in Seismic Zones Nos. 3 and 4 having a plan irregularity of Type B In Table No. 23-N, diaphragm chords and drag members shall be designed con­sidenng Independent movement of the projecting wings of the structure. Each of these dIaphragm elements shall be designed for the more severe of the following two assumptIons:

~otlon of the projecting wings in the same direction.

Motion of the projecting wings in opposing directions.

EXCEPTION: This requirement may be deemed satisfied if the procedures of Section 2335 in conjunction with a three-dimensional model have been used to deter­mine the lateral seismic forces for design.

217

APPENDIXB

FILTERING OF RELATIVE DISPLACEMENT HISTORIES

Corrected absolute acceleration, absolute velocity, and absolute displacement records were provided by

the California Office of Strong Motion Studies for each of the tilt-up buildings studied [37,38,59,60,61,66,

67]. Relative displacements are typically used to interpret structural response, because the displacement of

a building relative to its foundation is a measure of the distortions within the structure during the earthquake.

However, digitizing, filtering, and integrating the measured acceleration records introduces noise [68].

Therefore, reliable values of relative displacement can not be calculated simply by subtracting the absolute

displacement of the ground from the absolute displacement of the structure.

In an attempt to quantify the amplitude of the error introduced during the digitization process, Shakal and

Ragsdale [68] digitized a straight line as if it were an acceleration trace. The digitized acceleration records

were filtered and integrated to obtain absolute velocity and displacement records. The error introduced in the

acceleration, velocity, and displacement records is plotted in Fig. B.1 as a function of the long-period filter

eut-off period. The results indicate that errors introduced by the digitization process are likely to be on the

order of 2 cm/sec2 for the acceleration records. Errors in velocity and displacement waveforms depend on

the freq ueney 0 f the long-period filter cut-off and increase as the period of the filter cut-off increases. Noise

introduced by the recording instrument has been ignored in this process, and the results should be considered

to be a lower bound to the amplitude of the actual noise introduced [68].

An Ormsby filter was used during the processing of all the strong-motion data considered in this report

[38]. The shape of the filter is shown in Fig. B.2. Frequency cut-offs are summarized in Table B.1 and were

determined using an iterative procedure. Progressively shorter long-period filter cut-off periods were used

to remove as m lieh noise as possible while retaining as much signal as possible. Given this information, an

estimate of the reliability of the displacement records can be made. For example, records measured in the Hol­

lister warehouse during the Lorna Prieta earthquake have a useable bandwidth between 0.12 and 23.6 Hz

(0.042 and 8.40 sec) [38]. The long-period filtercut-off corresponds to approximately 1 em or D.4 in. of pro­

eessing noise (Fig. B.l).

The procedure used to filter all the relative displacement data will be illustrated using the transversestruc­

tural displacement data recorded in the Hollister warehouse during the Lorna Prieta earthquake. Channel 4

(located at the center of the roof) is used to represent structural displacements and Channel 7 (located at the

center of the west wail) is used to represent the ground movement. Digitized absolute displacement records

for the two channels are plotted in Fig. B.3(a) and the Fourier amplitude spectra are shown in Fig. B.3(b).

Both plots indicate that the ground motion dominates the absolute displacement response at the roof. The

structural vibrations observed between 8 and 16 sec in the structural displacement record correspond to the

218

peak in the Fourier amplitude spectra at approximately 1.1 Hz. This corresponds to the predominant frequen­

cy identified from the acceleration records (Table 5.2). The influence of the Ormsby filter may also be seen

in Fig. B.3(b) where it is evident that the long-period response (frequencies less than 1.4 Hz) has been re­

moved from both the ground and structural signals.

The relative displacement signal obtained by subtracting the ground displacement (Channel 7) from the

structural displacement (Channel 4) at each time increment is shown in Fig. B.4(a) and the corresponding

Fourier amplitude spectra is shown in Fig. B.4(b). Although the predominant frequency in the relative dis­

placement history occurs at 1.1 Hz, it is clear that a significant amount of noise from the ground displacement

is present. The long-period oscillations that occur after 30 sec in the relative displacement history also indi­

cate the presence of noise.

A high-pass filter was used to remove the portion of the signal attributable to the ground motion. The

shape of the filter is shown in Fig. B.5, and the frequency limits were selected using an iterative approach.

The cut-off frequencies were increased until the amplitude of the filtered relative displacement response his­

tory tended toward zero at the end of the record. The resulting filtered relative displacement history is shown

in Fig. B.6. Cut-off frequencies for the different earthquakes are presented in Table B.1.

As discussed in Chapter 5, the general shape of the unfiltered and filtered relative displacement records

did not change appreciably for the transverse displacements measured near the center of the buildings. How­

ever, the nature of the filtered and unfiltered longitudinal relative displacements and transverse relative dis­

placements measured at the top of the end walls were considerably different. In most cases, the maximum

amplitude of the filtered relative displacements at these locations were of the same magnitude as the ampli­

tude of the expected error shown in Fig. B.1. Therefore, only. the transverse relative displacement data mea­

sured near the center of the buildings were considered to be reliable.

219

TABLE B.1 FILTER LIMITS USED TO PROCESS STRONG-MOTION RECORDS

Limits for Building and Earthquake CSMIP Limits for Ormsby Filter High - Pass Filter

(Fig. B.2) (Fig. B.5)

fLt (Hz) fLc (Hz) fHt (Hz) fHe (Hz) h (Hz) h (Hz)

Hollister Warehouse

1984 Morgan Hill 0.08 0.16 23 25 0.20 0.50

1986 Hollister 0.10 0.20 23 25 0.20 0.50

1989 Lorna Prieta 0.07 0.14 23 25 0.20 0.50

Redlands Warehouse

1986 Palm Springs 0.25 0.50 23 25 1.10 1.25

Milpitas Industrial Building

1988 Alum Rock 0.30 0.60 23 25 2.0 2.5

1989 Lorna Prieta 0.07 0.14 23 25 2.0 2.5

220

u

'" V>

Q z § <:

....

..l

'"' u u

""' '" '" w c..

20,Or 2.0,-

T I

10.O~ It- 10

I ~ ~ l f ~ I---- ,,~ 1.0i-- .:0

C L-

t t L

~ L. i_ I I J J I I .. I I I '! 111 I I I I I 1111 .01 O~ 1.0 10.0 20.0 0.5 1.0 10.0 20.0 0.5 10.0

P::RICD(SECl ?::RIOD (SECl PERIOD (SEC)

Fig. B.l Estimate of processing noise present in corrected strong-motion records [68].

Q)

'"tJ ::J

7.0

t---- Useable Bandwidth--..-!

~ 0.7 ............................................................................................. . Q E 0.5 ~

L.... Q) ......

~ 0.0'------'---'---------------....1...-....1.----

Frequency, Hz

Fig. B.2 Ormsby filter used to process CSMIP records [38].

221

200

0.0 !

: ~

7.00 II I

Q.) \:) ~ ~ '-Q E ~

l.... Q.) '-l.... 0.50 ~

G: 'b Q.)

.~ "0 ~ 0 <:

0.00 0.0

Hollister Warehouse 1989 Loma Prieta Earthquake

Transverse, Absolute Displacemen t Response

10.0 20.0 Time. sec

30.0 40.0 50.0 ! ! ! I !

(a) Transverse displacement history.

Transverse, Absolute Displacemen t Response

0.5

Ground Displacement (Channel 7)

- - - - - Roof Displacement (Channel 4)

( ,

1.0

\ \

', .... - '- ........... -- ....

1.5 Frequency. ?

(b) Fourier amplitude spectrum.

Fig. B.3 Absolute displacement response.

222

2.0

60.0 I

2.5

0.0 !

Hollister Warehouse 1989 Loma Prieta Earthquake

10.0 !

Fig. B.4

Transverse Roof Response

20.0 !

Time, sec

30.0 !

40.0 !

(a) Transverse displacement history.

Transverse Roof Response

Frequency, Hz

(b) Fourier amplitude spectrum.

50.0 !

Unfiltered, relative displacement response at the roof.

223

60.0 !

0.0 ,

Q,) \) ::::J ....... . -Q

1.0

E 0.5 ~

l... Q,)

.......

~ o.o~--~----~--------------------------------Frequency, Hz

Fig. B.5 High-pass filter used to calculate relative displacement response.

Hollister Warehouse 1989 Loma Prieto Earthquake

1:J.O

Fig. B.6

Transverse Roof Response

20.0 I

Time, sec

30.0 I

40.0 I

50.0 I

Filtered, relative displacement response at the roof.

224

60.0 I