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SEISMIC PERFORMANCE OF HYBRID COREWALL BUILDINGS

Bahram M. SHAHROOZ1, Patrick J. FORTNEY2, Gian Andrea RASSATI3

SUMMARY

A common structural system involves the use of reinforced concrete core walls, which are

typically formed by coupling individual wall piers, and steel perimeter frames. For low-to-moderate rise buildings up to 2530 stories, the core walls are the primary lateral load resistingsystem, and the perimeter frames are designed for gravity loads. The use of dual systems is morecommon in taller buildings, in which the perimeter frames are engaged (through outrigger beamsor story-deep outrigger trusses) with the walls/cores as a means of reducing lateral drifts. Thispaper provides an overview of a number of issues related to design of hybrid structures with

reinforced concrete central core walls and perimeter steel frames. A number of design options forsteel or steel-concrete composite coupling beams and their connections to core walls, andconnections between outrigger beams and core walls are discussed. Preliminary results from anew concept for enhancing the performance of coupling beams are also presented.

Keywords: composite construction, core walls, coupling beams, hybrid walls, outrigger beams,seismic design.

INTRODUCTION

An efficient hybrid structural system is obtained when reinforced concrete core walls are used in conjunctionwith steel perimeter frames. Core walls can effectively be formed by coupling individual wall piers with the useof reinforced concrete, steel, or hybrid coupling beams. The walls may be reinforced conventionally withlongitudinal and transverse reinforcing bars, or may include embedded structural steel boundary columns inaddition to conventional reinforcing bars. For low-to-moderate rise buildings up to 25 to 30 stories, the core canbe used to provide a majority of the lateral force resistance. For taller buildings, the use of dual systems is more

common, where the perimeter frames are engaged with the core. Outrigger beams are framed between the corewalls and columns (which may be all steel or composite) in the perimeter frame. The successful performance ofsuch hybrid structural systems depends on the adequacy of the primary individual components which are the corewalls, steel frames, and frame-core connections. The focus of this paper is to provide an overview of designissues for (a) steel or hybrid coupling beams and their connections to core walls, and (b) connections betweenoutrigger beams and core walls. Each of these two issues is discussed separately.

COUPLING BEAMS

Coupling beams are analogous to and serve the same structural role as link beams in eccentrically braced frames.Reinforced concrete coupling beams are subject to significant limitations in terms of the shear stress they have toresist, and often require impractically deep sections to carry the loads demanded of them. Steel placement for

diagonally reinforced beams is often impractical for beams having a span-to-depth ratio greater than 1.5. Finally,the displacement capacity of reinforced concrete coupling beams has been shown often be exceeded by thedemand. Viable alternatives to reinforced concrete coupling beams are steel beams or steel beams encased in

1 Professor, Department of Civil and Environmental Engineering, University of Cincinnati, 765 Baldwin Hall, Cincinnati, Ohio 45221-0071,U.S.A., e-mail: [email protected]

Doctoral Research Assistant, Department of Civil and Environmental Engineering, University of Cincinnati, 765 Baldwin Hall, Cincinnati,Ohio 45221-0071, U.S.A., e-mail:[email protected] Visiting Professor, Department of Civil and Environmental Engineering, University of Cincinnati, 765 Baldwin Hall, Cincinnati, Ohio

45221-0071, U.S.A., e-mail: [email protected]


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concrete having varying levels of longitudinal and transverse reinforcement, which are referred to as compositeor hybrid coupling beams. The advantages of steel or hybrid coupling beams become particularly apparent incases where height restrictions do not permit the use of deep reinforced concrete beams, or where the required

capacity, stiffness, or deformation capacity cannot be developed economically with a concrete beam. Steelcoupling beams have been used in a number of structures (Ferver, et al., 1974, Lehmkuhl, 2002, Paulay, 2002,Taranath, 1998,Yao, 2003), and more cases with steel coupling beams are currently in the design phase. Similar

to reinforced concrete coupling beams, energy dissipation characteristics of steel or composite coupling beamsplay an important role in the overall performance of the structure. Depending on the coupling beam length, steelor hybrid coupling beams may be detailed to dissipate a major portion of the input energy by flexure or by shear.

For most coupling beams, however, it is more advantageous to design the coupling beams as shear critical, orshear yielding members since such members exhibit a more desirable mode of energy dissipation. Such a choiceis not possible for reinforced concrete members.

Coupling Beam Wall Connections

The coupling beam-to-wall connection depends on whether the wall boundary element is reinforcedconventionally or contains embedded structural steel columns. Two possible connections for the latter case areshown in Fig. 1. Ifthe wall boundary

element isreinforced withlongitudinal and

transversereinforcing bars, atypical connectioninvolvesembedding thecoupling beam

into the wall andinterfacing it withthe boundaryelement, as shown

in Fig. 2. Inaddition to

boundary elementreinforcing, embedded steel members may also be providedwith welded vertical reinforcing bars, attached to theflanges through mechanical half couplers welded to theflanges, or with shear studs. The vertical bars have beenfound to assist in transfer of bearing stresses around the

flanges, and to enhance the overall stiffness of theconnection (Shahrooz et al., 1993). Adequate control of thegap that opens at the beam flanges upon load reversal mayalso be provided by (a) placing two-thirds of the requiredvertical boundary element steel within a distance of one half

the embedment length from the face of the wall, and (b)ensuring that the width of the boundary element steel not toexceed 2.5 times the coupling beam flange width (Harries et

al., 1997). It is not necessary, nor is it practical, to passboundary element reinforcing through the web of theembedded coupling beam. It is, however, necessary toprovide good confinement in the region of the embedded

web. Confinement may be accomplished using hairpins andcross ties parallel to the web as shown in Fig. 2 (section A-

A). Additionally, vertical boundary element reinforcementin this region may also be relied upon to provide significantconfinement to the embedded region.

Figure 1Two types of hybrid coupling beam connected to embedded steel columns

(a) steel coupling beam attachedto steel boundary column

moment connection

end plate

wall steel not shown

embedment providesshear resistance

(b) steel coupling beam attachedto steel erection column

shear connection

wall steel not shown

embedment providesmoment resistance

shear studs as requiredend plate

SECTION A-A

SECTION B-B

A

B

A

B

Figure 2Steel coupling beam embedment

details


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Calculation of Embedment Length

The coupling beam is embedded in the wall such that its capacity can be developed. A number of methods

(Marcakis and Mitchell, 1980; Mattock and Gaafar, 1982) may beused to calculate the necessary embedment length. These methodsassume rigid body rotation of the embedded section and calculate

the internal moment arm between bearing forces generated at eachend of the embedment as shown in Fig. 3. Proposed methodsdiffer simply by the assumed stress distribution in the embedment

concrete. For steel coupling beams, Vu is taken as the plastic shear

capacity of the steel member, i.e., wfyp tthFV )2(6.0 = where

Fy = yield strength of web steel, h = beam depth, tf = flangethickness, and tw = web thickness. To account for strain hardeningand material over strength, it is recommended that Fy be taken as1.5 times the nominal yield strength. For hybrid coupling beams,

shear capacity may be taken as )V(V.V RCsteeln += 561 in

which )t(h-tF.V fwysteel 260= and

s

dfA

dbf'.V

yvwcRC += 1660 where bw = web width of the

encasing element around steel coupling beam, d = effective depthof the encasing element, Av = total area of stirrups in the encasingelement, fy = yield strength of the stirrups, and s = spacing of the

stirrups. This method has been calibrated based on a relativelylarge number of case studies (Gong and Shahrooz, 2001). Thenominal values of Fy and fc (in MPa) are to be used because theequation has been calibrated to account for strain hardening andmaterial over strength. Additional transfer bars attached to thebeam flanges (as discussed above) can contribute to the capacity ofthe embedment. The required embedment length can be modified to account for the additional strength.

However, to ensure that the calculated embedment length is sufficiently large to avoid excessive inelastic

damage in the connection region, it is recommended that the contribution of transfer bars be neglected in theembedment capacity design.

Design and Detailing of Coupling Beams

Well established guidelines for shear links in eccentrically-braced frames may be used to design and detail steelcoupling beams (AISC, 2002). The expected coupling beam rotation angle plays an important role in therequired beam details such as the provision of stiffeners. This angle is computed with reference to the collapsemechanism shown in Fig. 4 which corresponds to the expected behavior of coupled wall systems. The value of

p is taken as Cde, where Cd is the deflection amplification factor (ICBO, 2000), and e is the elastic inter-storydrift angle computed under code level

lateral loads. Knowing the value ofp,

the shear angle, p, is calculated asindicated in Fig. 4.

Previous research on hybrid couplingbeams (Gong and Shahrooz, 2001)

indicates that nominal encasementaround steel beams provides adequateresistance against flange and webbuckling. Therefore, hybrid coupling

beams may be detailed without webstiffeners. Due to inadequate data

regarding the influence of encasement onlocal buckling, minimum flange and web

slenderness requirements similar to steelcoupling beams should still be used.

L

Lwall

p

p

p p=L

Lwall

PlasticHinges

Figure 4 Determination of coupling beam angle of rotation

c L/2 = a

Vu

spalled cover

concrete

CL

Le - c

Le

Cb

Cf

f= 0.003

b xfxb

assumedstrain

distribution

1/3(Le-c)

0.85fc'

Cb

3/4(Le-c)

assumedstress

distribution

Mattock &Gaafar

1xf

xb 0.85fc'

fc'assumed

stressdistribution

Marcakis &Mitchell

Figure 3 Methods for computing

embedment length


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A pair of stiffener plates (on both sides of the web) placed along the embedment length will mobilizecompression struts in the connection region as depictedschematically in Fig. 5. These stiffener plates are commonly

referred to as face bearing plates. The first face bearing plateshould be inside the confined core of wall boundary element.

The distance between the plates should be such that the angle of

the compression struts is approximately 45 degrees (hence, thedistance between the plates should be about equal to the cleardistance between the flanges). To ensure adequate contribution

of face bearing plates, the width of each face bearing plateshould be equal to the flange width on either side of the web.

Thickness of the face bearing plates can be established based onavailable guidelines for detailing of shear links in eccentrically-braced frames (AISC, 2002).

Performance of Steel or Hybrid Coupling Beams

Recent tests by Gong and Shahrooz (2001) suggest satisfactory

performance of coupling beams designed and detailed accordingto aforementioned design and detailing procedures. Thedistribution of dissipated energy in Fig. 6 shows that the majorityof the input energy is dissipated through inelastic deformations

in the beam outside of the connection region. Moreover, thedesign methodology results in connections that exhibit stablehysteretic response and satisfactory ductility (see Fig. 7).

Despite satisfactory performance in terms of energy dissipation characteristics and hysteretic responses, theconnection region around steel or hybrid coupling beams is typically damaged extensively under extreme loading.The repair will beexpensive, and will

involve expensivechipping of the wallaround the connectionand placement of newcoupling beams. In aneffort to control the

level of damage, anongoing study has

been focused on

coupling beams withfuses where the

R.C. Wall

Transfer Bar

~

~

~

Coupling Beam

Face BearingPlates

CompressionStruts

Figure 5 Face bearing plates

Figure 7 Hysteretic response of hybrid coupling

beams

-300

-200

-100

0

100

200

300

-0.15 -0.10 -0.05 0.00 0.05 0.10 0.15

Shear Angle (rad.)

Shear(kN

)

Figure 6 Distribution of dissipated energy

0

20

40

60

80

100

0 0.025 0.05 0.075 0.1

Shear Angle (rad.)

CumulativeDissipatedE

nergy(kN-m)

Input

Connection

Beam

Figure 8 Schematic drawing of fuse for coupling beams

Fuse


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inelastic deformations are concentrated. The fuses are also detailed to eliminate/limit the damage to the couplingbeams and wall piers, and to ensure ease of replacement. A schematic drawing of such a system is shown in Fig.8. The fuse should be designed to ensure shear yielding. The plastic moment capacity of the fuse is set to be

equal or greater than the corresponding value for the coupling beam, and the plastic shear capacity of the fuse istaken as a percentage (less than 100%) of the plastic shear capacity of the coupling beam. A series of

preliminary tests were recently conducted to examine this concept. In these tests, steel beams were used to

represent wall piers. The fuses performed per design, i.e., the coupling beam remained elastic while the fusedissipated a significant amount of energy through yielding of the web while undergoing large rotations (Fig. 9).

The concept of coupling beams with fuses is being further evaluated through specimens that incorporate wallpiers. Fabrication of these specimens is currently underway, and testing would be completed by early 2004.These tests will provide further data regarding performance of fuses in coupling beams, particularly relationships

between damage in the wall piers as a function of plasticity in the fuse will be established.

OUTRIGGER BEAM-WALL CONNECTION

In low-rise buildings, up to about 30 stories, the core is the primary lateral load resisting system, and theperimeter frame is designed for gravity loads, and the connection between outrigger beams and cores is generallya shear connection. A typical shear connection is shown in Fig. 10 in which a steel plate with headed studs isembedded in the core wall during casting. After casting

beyond the plate, the web of the steel beam is bolted to theshear tab which is already welded to the plate. Variationsof this detail are common. Although this connectionprovides some moment resistance, it is generally accepted

that the connection is flexible and does not develop largemoments; hence, such connections are considered to be shearconnections. In taller buildings, story-deep trusses may be

used to engage the perimeter columns as a means of reducingthe overall deformation of the structure. The connectionbetween the top and bottom chords is essentially similar tothat shown in Fig. 10.

In an effort to develop design procedures for outrigger beam-

wall connections, a coordinated experimental and analyticalstudy was recently completed at the University of Cincinnati(Shahrooz et al., 2003). The main objectives of this studywere to (a) investigate effects of concrete cracking and

yielding of wall longitudinal bars on the performance of outrigger beam-wall connections, (b) examine influenceof wall boundary elements around the connection region on the connection capacity, (c) evaluate significance of

Figure 9 Fuse rotation-shear strain

-25

-20

-15

-10

-5

0

5

10

15

20

25

-32,000 -24,000 -16,000 -8,000 0 8,000 16,000 24,000 32,000Shear strain (micro strain)

R

otation(%r

ad.)

Rotation @

shear yielding

OutriggerBeam

Steel Shear Tab

R.C.

CoreW

all

High Strength Bolts

Embedded Platew/ Shear Studs

Figure 10 Outrigger beam-core wall shear

connection


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floor slab on connection behavior, and (d) develop design guidelines.

Test Specimens and Testing Method

Two -scale specimens were selected. Each specimen had a cantilever wall that represented a portion of a core

consisting of a central I-shaped and two C-shaped walls.A rectangular wall, corresponding to the web of the I-shaped wall of the core, was used in the first specimen.

The second specimen had a T-shaped wall, whichrepresented the web and a portion of the flange of the I-

shaped portion of the core wall. Each specimen had twooutrigger beam-wall connections (Fig. 11). The firstconnection was located at 610 mm above the base, whichcorresponded to the average length of the expected plastic

hinge. The wall around the first connection had boundaryelement reinforcement. The second connection was

placed at 610 mm above the first connection. At thislocation, the wall would experience moderate levels of

cracking, and the wall did not have boundary elementreinforcement. The outrigger beam-wall connection wassimilar to that shown in Fig. 10. The connections weredesigned to resist gravity shear and diaphragm axial load

(Shahrooz et al., 2002). Four 9.5-mm diameter studs wereused for the embedded stud plate, and three 19-mm-diameter A490 bolts were used to connect the beam to theembedded stud plate. The studs were 102 mm and 69 mmlong in the rectangular wall and T-shaped wall,respectively. The flange in the T-shaped wall increases

the concrete pullout capacity of the studs; hence, therequired stud length for the T-shaped wall is about onehalf of that in the rectangular wall.

Floor diaphragm was included inthe second specimen, refer to Fig.

12. The floor metal deck wasconnected to the top flange of theoutrigger beam by five 6.4 mmdiameter x 35 mm long studsspaced at 152 mm on center, withthe first stud placed at 95 mm from

the face of the wall. The metaldeck was tack welded to the topsurfaces of two 50.8x50.8x6.35mm seat angles (one on each side

of the outrigger beam) that hadbeen connected to the wall flangeby six self-tapping 6.4-mm

diameter screws with 38 mmembedment length (3 on each sideof the outrigger beam). The connection of the metal deck to the wall was intended to resist only the factoredfloor gravity loads.

The general layout of the test setup is shown in Fig. 13. A constant axial load equal to 0.1fcAg (Ag = wall cross-

sectional area, fc = concrete compressive strength) was applied to the walls. Up to reaching 2.5% drift, therectangular wall and outrigger beams were loaded simultaneously. After subjecting the wall to 2.5% drift,loading of the wall was stopped, and each connection was separately subjected to cyclic loading until failure.Due to concerns for the adequacy of the loading system and laboratory schedule, loading of the T-shaped wall

was limited to only 0.5% before individual testing of each connection to failure. A constant gravity shear equalto 26.2 kN was applied to the top connection of both specimens. The addition of gravity shear for the bottom

Seatangle on each side

Metal Deck

Anchors

Concrete

Infill

R.C. Core Wall

Figure 12 Diaphragm-core wall connection

2743

914

a

457

1295

610

610914

2ndConnection

1st

Connection

457

#[email protected]

#3

#3@102#4@95

#8@95

#5@203

b

c

a = 1143 for rectangular wall; 1270 for T-shaped wall

b = 914 for rectangular wall; 1080 for T-shaped wall

c = 114 for rectangular wall; 241 for T-shaped wall

Diaphragm is notshown for c larity.

Figure 11 Wall reinforcement


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connection was determined to have minimal effects on the connection capacity, and hence was not included(Shahrooz et al., 2002).

Test Results

(a) Rectangular WallAt 2.5% drift, the rectangular wall experienced significant cracking and spalling around the bottom connection,and moderate level of cracking around the top connection. The wall longitudinal reinforcing bars were strainedat least 10 times the yield strain at the base, and approximately 6 times the yield strain at the location of the

bottom connection. The strain gages on the longitudinal bars near the top connection were lost during fabricationof the specimen, but the crack patterns do not suggest significant inelastic action around the top connection. Thestuds in the bottom and top connections yielded, and were strained to 120% and 187% of the yield strain,

respectively, when the wall was pushed to 2.5% drift. The differences in the stud strains in the top and bottomconnection are mainly attributed to the presence of boundary element around the bottom connection, whichreduced the participation of the studs. The ratio of the average strain in the bottom connection studs to the

corresponding value in the top connection was 0.56 at the start of the test, and changed slightly throughoutvarious stages of testing. Events such as yielding of the wall longitudinal bars and studs did not significantlyimpact this ratio, and the average value of 0.58 remained essentially the same as the original value. The lowervalues of stud strains in the bottom connection are also apparently due to the level of cracking. The cracksaround the bottom connection were considerably more extensive than those around the top connection. Hence,widening of the cracks mostly dissipated the input energy in the bottom connection, whereas the corresponding

energy in the top connection was predominately dissipated through yielding of the studs.

Despite excessive damage and yielding of the wall longitudinal bars around the bottom connection, at theconclusion of combined wall-connection tests the connection did not fail. The top connection did not fail either,

albeit the level of damage and inelasticity around the connection was moderate as expected. When loadedindividually, the bottom connection and top connection failed due to stud fracture and stud pullout, respectively.The boundary element around the bottom connection apparently prevented pullout of the studs.

(b) T-Shaped WallAs mentioned before, combined testing of the wall and connections was stopped after subjecting the wall to alateral drift of 0.5%. At this drift, the wall longitudinal bars in the flange and web at the base were strained to

4.6 y and 2.3y, respectively, but the level of cracking in the wall was minimal. The wall longitudinal bars nearthe bottom and top connections did not yield; the maximum strain in these bars was 0.65 y near the bottomconnection. Similar to the rectangular wall specimen, the stud strains in the top connection, which was not

surrounded by heavily confined boundary element, were larger than the bottom connection stud strains. At 0.5%

drift, the studs in the top connection yielded and reached a strain of twice the yield strain, whereas the bottomconnection studs remained in the elastic range (the maximum strain was 0.43y). This difference is attributed to

0.1f Ac g

P

P/2

P/2

26.2 kN

2743mm

Footing

Loading Slab

Figure 13 Test setup and loading


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confinement by the boundary element that reduced the participation of the studs in the bottom connection. Incontrast to the first specimen, both the top and bottom connections in the specimen with T-shaped wall failed as aresult of stud fracture.

The participation of the floor diaphragm in transferring forces to the wall was negligible. For example, the

strains in the concrete deck were on the average 18% of the strains measured in the outrigger beam, and the slab

participation away from the outrigger beam was rather insignificant. The majority of the load was transferreddirectly to the wall through the outrigger beam-wall connection. Diaphragm-wall connections may, therefore, bedesigned for gravity loads only unless slab bars are interfaced with walls, or other positive means such as

collector anchors are used.

Assessment of Connection Capacity and Mode of Failure

The analytical model shown in Fig. 14 was used to compute capacity and mode of failure of the connections in both

specimens. At stud pullout failure, the concrete strain around studs is small, and concrete essentially behaveselastically. Therefore, a linear distribution of concrete

stress is assumed in the model. The maximum concrete

stress (fc) is computed from

t

u

c I

eVkdf = in which kd and

It (transformed moment of inertia) are calculated based onstandard techniques for a cracked transformed section

analysis of reinforced concrete beams. The studs areconsidered as reinforcing bars, the width of the stud plate is

used as the beam width, and for simplicity, the contributionof the studs in compression is ignored. Knowing kd, thevalue of C is 0.5bfc kd. In the model, the contribution offriction between the embedded stud plate and concrete is

also taken into account. Equation (1) is used to computethe tensile load that can be resisted by the connection.

( ) 1V4

CV0.5

T

0.5TC2

c

u

2

cap.

u =

+

+ (1)

In this equation, the stud plate is assumed to behave rigidly,and accordingly the gravity shear (Vu) and applied axial

load (Tu) are divided equally between the studs. Note thatthe coefficients of 0.5 and 4 are for the test specimens in which four studs were used. For other cases, thesecoefficients have to be changed depending on the number of studs (Shahrooz et al., 2002). The coefficient of

friction () was taken as 0.4. The design shear capacity of studs, Vc, is computed from the applicable equations inPCI Handbook (PCI, 2001). The tensile capacity of the studs, Tcap., is taken as the smaller of the stud fracturecapacity and pullout capacity, which are computed based on PCI equations with the following modifications:

1. For connections in which the studs are surrounded by a boundary element, the compressive strength of concreteis increased by 20% to reflect the confining effects of the boundary element. This increase is in accordance totests on strength and stiffness of confined concrete (Saatcioglu and Razvi, 1992).

2. The effective length of studs located in a boundary element is taken as the depth of the boundary element.

3. Stud fracture strength is computed based on the ultimate tensile strength of the stud instead of the yield strength.

4. The pullout capacity of stud groups is computed by assuming that the failure cone makes an angle of 55degrees with the plane of the stud head in accordance to the failure patterns observed in the tests. This angle isdifferent from 45 degrees assumed in PCI equations, but is consistent with provisions in Appendix D of the2002 edition of ACI Building Code (ACI, 2002) and within the range of 50 to 60 degrees recommended byothers (e.g., Rehm et al., 1988). The larger angle of 55 degrees increases the projected area of cone failure by

25.8% over the value stipulated in PCI equations.

Tcap.

C

fc

e

kd

d

Vu

Tu

C

Figure 14 Analytical model


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As evident from Table 1, the analytical model accurately predicts the mode of failure. Moreover, the modelestablishes a very close estimate of the measured capacities(see Fig. 15); on the average, the experimental capacities are

2% more than the computed values (with a coefficient ofvariation of 1.3%). Note that without the aforementioned

adjustments, the computed modes of failure and capacities

would have been appreciably different from theirexperimental counterparts (Shahrooz et al., 2003b).

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

1

1.1

R-Wall:Bottom

Connection

R-Wall:Top

Connection

T-Wall:Bottom

Connection

T-Wall:Top

Connection

Measured/Co

mputed

Figure 15 Comparison of measured and computed connection capacities

SUMMARY AND CONCLUSIONS

A number of past and ongoing studies at the University of Cincinnati have examined seismic behavior of hybridcore wall structures. In particular, seismic behavior of steel and hybrid coupling beams, and connections between

outrigger beams and core walls have been investigated.

Previous research has shown the viability of steel and hybrid coupling beams, and has been instrumental indevelopment of design methods. Adequate capacity and distribution of dissipated energy for steel and hybridcoupling beams are achieved by following these design methods. Ongoing studies are aimed at developing the nextgeneration of steel and hybrid coupling beams in order to eliminate and/or limit damage to wall piers while

dissipating a significant amount of energy, and to develop an easily repairable system. Test results from a numberof preliminary specimens are promising, and suggest that inelastic deformation can be concentrated in replaceable

fuses while ensuring elastic response of the coupling beam.

An accurate design model for outrigger beam-core wall connections has been developed. This model predicts theconnection capacity and its expected mode of failure. Using this model, a capacity design approach may be

followed to prevent brittle failure modes (i.e., stud pullout failure or stud fracture) by dissipating energy throughductile yielding and eventual fracture of the shear tab connecting the outrigger beam to the embedded stud plate.

Using this approach, the shear tab will act as an effective energy dissipating fuse.

ACKNOWLEDGEMENTS

The reported research is based on investigations sponsored by the National Science Foundation under grants

BCS-9319838, CMS-9632496 and CMS-9714860, with Dr. Shih Chi Liu as the program director. These projects

were part of the fifth phase of US-Japan cooperative research program on composite and hybrid structures. Anyopinions, findings, and conclusions or recommendations expressed in this paper are of those of the authors anddo not necessarily reflect the views of the sponsors.

Table 1 Experimental and analytical modes

of failure

Connection Computed ObservedBottom Stud Fracture Stud Fracture

Top Stud Pullout Stud Pullout

Connection Computed Observed

Bottom Stud Fracture Stud Fracture

Top Stud Fracture Stud Fracture

(a) R-Wall

(b) T-Wall


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